AN ABSTRACT OF THE THESIS OF
Nason J. McCullough for the degree of Master of Science in Civil Engineering
presented on February 23, 1998. Title: The Seismic Vulnerability of Sheet Pile Walls.
Abstract approved:
Redacted for privacy
Stephen E. Dickenson
The seismic performance of port structures has been well documented following
recent earthquakes, and indicates that port structures are highly susceptible to earthquakeinduced damages. These damages are primarily due to soil liquefaction and the associated
ground failures. Sheet pile bulkheads provide vital intermodal and lifeline transportation
links between water-side and land-side traffic, and are waterfront structures particularly
vulnerable to liquefaction-induced damages. Due to the prevalence of liquefactioninduced damages, many ports are utilizing soil improvement techniques to mitigate these
hazards.
Many port authorities have proposed utilizing performance-based design criteria to
limit potential earthquake-induced damages. The current design method for sheet pile
walls (Mononobe-Okabe) is based on simple, limit equilibrium analysis techniques,
which are poorly suited for performance-based design. Recent advancements in the
seismic design of sheet pile walls have addressed some of the limitations of the current
design methods, but are still inadequate for performing a complete, performance-based
design for locations that contain potentially liquefiable soils and/or where soil
improvement strategies have been instituted.
This study has focused on conducting an empirical investigation and numerical
modeling to determine the seismic performance of sheet pile walls, and the performance-
based benefit of soil improvement through densification. A case history validated,
nonlinear effective stress computer program was used to perform numerical parametric
studies on various design parameters (earthquake properties, depth of sheet pile
embedment, sheet pile wall stiffness, tie rod length, density of the backfill, and extent of
soil densification). The results have been presented as a performance-based design
method, and include a design chart that provides practitioners with a preliminary design
tool that may be used to estimate the seismic deformations of sheet pile walls with or
without soil improvement.
The study has demonstrated that soil densification can greatly reduce the seismically-
induced deformations, especially when the magnitude of soil improvement extends
beyond the location of the anchor. The study has also demonstrated that the use of soil
densification techniques for mitigating seismic hazards may not be adequate in limiting
deformations to allowable limits, and that other methods of soil improvement
(cementation, drainage, etc.) or structural improvements may also be required.
C Copyright by Nason J. McCullough
February 23, 1998
All Rights Reserved
THE SEISMIC VULNERABILITY OF SHEET PILE WALLS
by
Nason J. McCullough
A THESIS
submitted to
OREGON STATE UNIVERSITY
in partial fulfillment of
the requirements for the
degree of
MASTER OF SCIENCE
Presented February 23, 1998
Commencement June 1998
Master of Science thesis of Nason J. McCullough presented on February 23, 1998
APPROVED:
Redacted for privacy
Major Professor, representing Civil Engineering
Redacted for privacy
Head or Chair:4 Department of Civil, Construction, and Environmental Engineering
Redacted for privacy
Dean of Gradu
School
I understand that my thesis will become part of the permanent collection of Oregon State
University libraries. My signature below authorizes release of my thesis to any reader
upon request.
Redacted for privacy
Nason J. McCullough, Autho
ACKNOWLEDGMENT
The completion of this project would not have been possible without the support and
guidance of my major professor, Stephen Dickenson. My minor professor, Charles Sollitt,
has also provided much support and guidance during my academic pursuits.
I am indebted to my parents, who have supported and encouraged me in all of my
endeavors, my sister who has always been a friend, and finally, and most importantly my
wife Amelia, who I love tremendously, and is the source of my joy and happiness.
I wish to extend my sincere gratitude to Dr. Susumu Iai (Director) and Mr. Koji Ichii
(Technical Official), both with the Geotechnical Earthquake Engineering Laboratory, Port and
Harbour Research Institute of the Japan Ministry of Transport. Many aspects of this research
would not have been possible without their assistance with the case studies and valuable
insights on Japanese design and construction of anchored bulkheads.
This work was supported by grant no. NA36RG0451 (project no. R/CP-33) from the
National Oceanic and Atmospheric Administration to the Oregon State University Sea Grant
College Program and by appropriations made by the Oregon State legislature. The views
expressed herein are those of the author and do not necessarily reflect the views of NOAA or
any of its subagencies. The author gratefully acknowledges the support of NOAA and Oregon
Sea Grant.
ii
TABLE OF CONTENTS
Page
1
INTRODUCTION
1.1
Background
1.2
Statement of Objectives and Scope of Work
1.2.1
1.2.2
Objectives
Report Organization
2 LIQUEFACTION ASSESSMENT OF WATERFRONT SOILS
10
10
12
13
Introduction
13
2.2
Cyclic Shear Method of Liquefaction Assessment
14
2.2.2
2.2.3
2.2.4
4
1
2.1
2.2.1
3
1
Determination of the Earthquake Induced Cyclic Shear Stresses
Cyclic Shear Stress Required to Initiate Liquefaction
Evaluation of Initiation of Liquefaction
Effects of Liquefaction
MITIGATION OF LIQUEFACTION HAZARDS
14
17
24
24
28
3.1
Introduction
28
3.2
Techniques for Mitigating Liquefaction Hazards
28
3.3
Design of Soil Mitigation
30
3.4
Design for the Area of Soil Mitigation
33
SEISMIC ANALYSIS AND DESIGN OF SHEET PILE BULKHEADS
35
4.1
Introduction
35
4.2
Pseudo-Static Design
36
4.2.1
4.2.2
Non-Liquefiable Soil
Liquefiable Soil
40
43
iii
4.3
Recent Advancements in Flexible Wall Design
4.3.1
4.3.2
4.4
5
Non-Liquefiable Soils
Liquefiable Soils
Discussions of Sheet Pile Bulkhead Design Methods
45
45
48
50
NUMERICAL MODELING
52
5.1
Constitutive Soil Model
53
5.2
Pore Pressure Generation
55
5.3
General Modeling Parameters
57
5.3.1
5.3.2
5.3.3
5.3.4
5.3.5
5.4
Modeling of Soil Elements
Modeling of Structural Elements
Modeling of the Earthquake Motion
Modeling of the Water
Boundary Conditions
Validation of Numerical Model
5.4.1
5.4.2
5.4.3
5.4.4
Akita Port
Ishinomaki Port
Kushiro Port
Discussion of Numerical Validation Results
6 PARAMETRIC STUDY ON THE SEISMIC PERFORMANCE OF SHEET
PILE BULKHEADS
57
59
60
60
61
61
62
72
79
91
93
6.1
Introduction
93
6.2
Depth of Sheet Pile Embedment
96
6.3
Tie Rod Length
96
6.4
Sheet pile Stiffness
97
6.5
Penetration Resistance of the Backfill
99
6.6
Soil Improvement
101
iv
Comparison of the Parametric Study with an Existing Deformation-Based
Design Method
102
6.8
Development of a New Seismic Design Method
103
6.9
Recommendation Design Procedures
107
6.7
6.10 Example Design Problem
108
7 SUMMARY AND CONCLUSIONS
112
BIBLIOGRAPHY
115
APPENDICES
122
APPENDIX A (Earthquake Motions Used in the Parametric Study)
APPENDIX B (Table of Parametric Study Results)
123
125
LIST OF FIGURES
Page
Figure
Lateral Displacements at: a) Shiohama Wharf, Akita Port following the
1983 Nihonkai Chubu Earthquake (PHRI, 1997) and b) Kushiro Port
following the 1993 Kushiro Oki Earthquake (PHRI, 1993)
2
Settlements at: a) Ohama Wharf, Akita Port following the 1983 Nihonkai
Chubu Earthquake (PHRI, 1997) and b) U.S. Navy Port Facilities, Guam
following the 1993 Guam Earthquake (Vandani et al., 1994)
3
Disruption of Crane Operations at Guam, following the 1993 Guam
Earthquake (Swan and Harris, 1993)
4
1.4
Typical Anchored Sheet Pile Bulkhead
5
1.5
Geotechnical Failures of Anchored Sheet Pile Walls; a) embedment
failure; b) anchor failure; and c) global failure
7
Structural Failures of Anchored Sheet Pile Walls; a) anchor, tie rod,
and/or whale system failure; and b) sheet pile wall failure
7
1.1
1.2
1.3
1.6
2.1
Range of rd Values for Different Soil Profiles (after Seed and Idriss,
1982)
15
2.2
Earthquake Induced Shear Stresses (Seed and Idriss, 1982)
16
2.3
Typical Results of Cyclic Shear Tests (PHRI, 1997)
18
2.4
Empirical Relationship Between the Cyclic Stress Ratio Initiating
Liquefaction and (N/)60 Values for Silty Sands in M=7.5 Earthquakes
(after Seed et al., 1979)
21
2.5
Relationship between a and Ka (Seed & Harder, 1990)
22
2.6
Relationship between the Effective Vertical Stress and Ka (Seed &
Harder, 1990)
22
Proposed CPT-based liquefaction curves based on field CPT and
liquefaction data (Stark & Olson, 1995)
23
2.8
Graphical Plot on the Evaluation of the Initiation of Liquefaction
24
2.9
Relationship Between Residual Excess Pore Pressure and Factor of
Safety Against Liquefaction for Level-Ground Sites (after Marcuson and
Hynes, 1990)
25
Estimation of Post-Liquefaction Volumetric Strain for Clean Sands (after
Ishihara and Yoshimine, 1992)
26
2.7
2.10
vi
Undrained Critical Strength Ratio vs. Equivalent Clean Sand Blow
Count (Stark and Mesri, 1992)
27
3.1
Example Remediation Design for a Sheet Pile Wall (PHRI, 1997)
32
3.2
Improvement area for sheet pile walls (PHRI, 1997)
33
3.3
Improvement Area for Sheet Pile Wall Anchors (PHRI, 1997)
34
4.1
Relation between Seismic Coefficient and Ground Acceleration (Noda et
al., 1975)
37
Seismic Coefficient Estimated from Damaged Quaywall (after Nozu et
al., 1997)
37
4.3
Definition Sketch for Pseudo-Static Design Parameters
38
4.4
Definition Sketch for the Location of the Resultant Dynamic Active
Earth Pressure
42
4.5
Definition of Effective Anchor Index, EAI (after Gazetas et al., 1990)
47
4.6
The Developed Seismic Design Chart (Gazetas et al., 1990)
48
5.1
Basic Explicit Calculation Cycle
53
5.2
Modeled Liquefaction Resistance Curve
55
5.3
Plan of Akita Port, Japan after the 1983 M 7.7 Earthquake (Iai &
Kameoka, 1993)
63
5.4
Akita Ohama Wharf 1 Wall Profile (Iai & Kameoka, 1993)
64
5.5
Akita Ohama Wharf 2 Wall Profile (Iai & Kameoka, 1993)
64
5.6
Deformed Cross-Sections for Akita Ohama Wharf 2
65
5.7
Soil Profile and Grain Size for Ohama Wharf 1 (Iai & Kameoka, 1993)
65
5.8
Soil Profile and Grain Size for Ohama Wharf 2 (Iai & Kameoka, 1993)
66
5.9
Shear Wave Velocity Profile for Ohama Wharf 3 (Iai & Kameoka, 1993)
67
5.10
Soil and Velocity Profile at the Akita Strong Motion Site (Iai &
Kameoka, 1993)
68
Recorded Time Histories for the 1983 M 7.7 Nihonkai Chubu
Earthquake at Akita Port
69
5.12
Input Time History for Akita Ohama Wharf 1
69
5.13
Input Time History for Akita Ohama Wharf 2
70
5.14
Deformed Grid for Akita Ohama Wharf 1 (deformations to scale)
70
5.15
Time Displacement Plot for Akita Ohama Wharf 1
71
5.16
Deformed Grid for Akita Ohama Wharf 2 (deformations to scale)
72
2.11
4.2
5.11
vii
5.17
Time Displacement Plot for Akita Ohama Wharf 2
72
5.18
Measured a) Horizontal and b) Vertical Displacements along the Face of
the Bulkhead (Iai & Kameoka, 1993)
73
Layout of Ishinomaki Port during the M 7.4 Miyagi-Ken-Oki Earthquake
(Iai et al, 1985)
74
5.20
Shiomi (-4.5m, Site D) Bulkhead Profile (Iai et al., 1985)
74
5.21
Boring logs at Shiomi Wharf (Iai et al., 1985)
75
5.22
Recorded Time Histories at Ishinomaki Port
77
5.23
Input Time History for Ishinomaki Wharf
77
5.24
Deformed Grid for Ishinomaki, Shiomi Wharf (deformations to scale)
78
5.25
Time Displacement Plot of Ishinomaki, Shiomi Wharf
79
5.26
Measured Horizontal Deformations at Shiomi Wharf (-4.5 m),
displacements in mm (Iai et al., 1985)
80
5.19
5.27
Measured Vertical Displacements at Shiomi Wharf (-4.5 m) (Iai et al.,
1985)
80
5.28
Location of Kushiro Port (Iai et al., 1994)
81
5.29
Layout of Kushiro Port (Iai et al., 1994)
81
5.30
Geometry of Kushiro Port Site C Wharf (Iai et al., 1994)
82
5.31
Geometry of Kushiro Port Site D Wharf (Iai et al., 1994)
82
5.32
Soil Profile at Kushiro Port Site C (Iai et al., 1994)
83
5.33
Soil Profile at Kushiro Port Site D (Iai et al., 1994)
84
5.34
Soil Profile at the Kushiro Port Strong Motion Site (Iai et al., 1994)
86
5.35
Earthquake Motions Recorded at the Ground Surface at Kushiro Port on
January 15, 1993
87
Earthquake Motions Recorded at a depth of 77 meters at Kushiro Port on
January 15, 1993
88
5.37
Input Time History for Kushiro Port Site C and Site D
88
5.38
Kushiro Site C Deformed Grid (deformations to scale)
89
5.39
Time Displacements for Kushiro Site C
89
5.40
Kushiro Site D Deformed Grid (deformations to scale)
90
5.41
Time Displacements for Kushiro Site D
90
6.1
Top of Wall Displacements for the Depth of Embedment Study
97
5.36
viii
6.2
Top of Wall Displacements for the Tie Rod Length Study
98
6.3
Top of Wall Displacements for the Sheet Pile Stiffness Study
99
6.4
Top of Wall Displacements for the Penetration Resistance of the Backfill
Study
100
6.5
Definition of Soil Improvement Variables
101
6.6
Horizontal Displacments at the Top of the Wall for Variations in the
Zone of Soil Improvement
102
The Developed Seismic Design Chart (Gazetas et al., 1990), Including
Data from the Parametric Study for Models with Improved Soils (solid
stars)
104
6.8
Design Chart for Lateral Displacements
106
6.9
Anchored Sheet Pile Bulkhead Design Problem
108
6.10
Lateral Displacements for the Design Problem
111
6.7
ix
LIST OF TABLES
Table
1.1
Page
Selected Historical Earthquake Damage to Anchored Sheet Pile
Bulkheads
6
2.1
Values of CSR Correction Factor, c,. (after Seed, 1979)
18
2.2
Magnitude Scaling Factors, MSF (Arango, 1996)
21
2.3
Recommended Fines Correction for Estimation of Residual Undrained
Strength (Stark & Mesri, 1992)
27
3.1
Liquefaction Remediation Measures (after Ferritto, 1996)
29
4.1
Qualitative and Quantitative Description of the Reported Degrees of
Damage (after Kitajima and Uwabe, 1979)
47
5.1
Comparison of FLAC and Finite Element Numerical Programs
54
5.2
Model Soil Properties for Ohama Wharf 1 and 2
66
5.3
Model Structural Properties for Ohama Wharf 1 and 2
68
5.4
Model Soil Properties for Ishinomaki Port
75
5.5
Model Structural Properties for Ishinomaki Port
76
5.6
Model Soil Properties for Kushiro Port
84
5.7
Model Structural Properties for Kushiro Port
85
5.8
Summary of Case Study Results
92
6.1
Bulkhead Properties for Parametric Study
94
6.2
Parametric Study Earthquake Motions
95
6.3
Plotted Case History Validations
106
6.4
Magnitude Scaling Factors for Design Example
109
6.5
Lateral Displacements for Design Example
110
THE SEISMIC VULNERABILITY OF SHEET PILE WALLS
1
1.1
INTRODUCTION
BACKGROUND
Experience at ports has demonstrated that waterfront earth-retention structures are
highly susceptible to earthquake-induced damage. A significant percentage of the damage
is due to the liquefaction of adjacent soils. Liquefaction susceptible soils consist of loose,
saturated, non-cohesive soils, and are frequently found in the port environment. These
soils are prevalent in the port environment due to hydraulic backfilling of waterfront
retaining structures and/or loose sediments deposited in alluvial
and marine
environments. Earthquake-induced bulkhead failures due only to large seismic inertia
forces have also been noted in locations that were reported as not experiencing
liquefaction (Werner and Hung, 1982), although it has also been noted that the occurrence
of liquefaction at depth may be masked on the ground surface by large wall deformations
(Towhata et al., 1996).
Earthquake-induced damage to waterfront retaining structures is usually manifested
as excessive lateral displacements, and/or settlements, with several recent examples
presented in Figure 1.1 and Figure 1.2. There are also many cases of widespread damage
to backland structures (e.g., gantry cranes, cargo handling components, buildings, bridges,
pavements, etc.) that are a result of the lateral displacements and settlements, as seen in
Figure 1.3 with the spreading of the gantry crane rails. The economic losses associated
with damaged port structures and the suspension of port operations due to earthquake
induced damages has been substantial. For example, the Port of Oakland following the
1989 Mw 6.9 Loma Prieta Earthquake, had repair costs estimated at $75 million (Werner,
1990). The U.S. Navy Facilities suffered a total of approximately $275 million in direct
2
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It('tt1::::"4;1121.1',7o"'g
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13)
Figure 1.1: Lateral Displacements at: a) Shiohama Wharf, Akita
Port following the 1983 Nihonkai Chubu Earthquake
(PHRI, 1997) and b) Kushiro Port following the 1993
Kushiro Oki Earthquake (PHRI, 1993)
1
3
a)
b)
Figure 1.2: Settlements at: a) Ohama Wharf, Akita Port following
the 1983 Nihonkai Chubu Earthquake (PHRI, 1997)
and b) U.S. Navy Port Facilities, Guam following the
1993 Guam Earthquake (Vandani et al., 1994)
4
Figure 1.3: Disruption of Crane Operations at Guam, following
the 1993 Guam Earthquake (Swan and Harris, 1993)
losses during the Loma Prieta Earthquake and the 1993 Mw 8.1 Guam Earthquake
(Ferritto, 1997). The U.S. Navy damage from both earthquakes was noted to be due
primarily to liquefaction (Ferritto, 1997). Another example of commercial port damage is
to the Port of Kobe following the 1995 Mw 6.9 Hyogoken Nanbu Earthquake, with repair
costs estimated at $5.5 billion (CKPHB, 1997), and the costs due to port downtime during
the first nine months following the earthquake estimated at $6 billion (PHRI, 1996).
Anchored sheet pile bulkheads (Figure 1.4) are particularly vulnerable to earthquake-
and liquefaction-induced damage. Seismic damage to anchored sheet pile bulkheads
following numerous international earthquakes has been well documented by Kitajima and
Uwabe (1979), Werner and Hung (1982), and Werner (1998). Table 1.1 provides a
cursory overview of earthquake damage to anchored sheet pile bulkheads, and also
includes a note on the observed occurrence of liquefaction. It is readily apparent from the
table that for the majority of anchored sheet pile bulkheads subjected to medium- to high-
5
intensity earthquake motions, liquefaction occurred, and was likely the primary cause of
the reported damage.
Al
H
=03
dredge line
3
wale
.).3z.
tie rod
anchor
backfill soil
Figure 1.4: Typical Anchored Sheet Pile Bulkhead
The damaging effects of earthquakes (for cases with or without liquefaction) include
increased active pressures due to the loss of soil strength and the seismic inertia of the
backfill, and the loss of passive soil resistance adjacent to the toe of the wall beneath the
dredge line and in front of the anchor. The increase in active pressures and decrease in
passive pressures leads to possible geotechnical failures including: a) the loss of
embedment resistance, b) the loss of anchor resistance, and/or c) a deep seated global
failure for bulkheads situated on weak foundation soils (Figure 1.5). Possible structural
failures of sheet pile walls include exceeding the yield capacity of the a) tie rod, tie rod
connections, wale system; or b) sheet pile wall (Figure 1.6). A structural interlock failure
between the sheet pile sections is also possible. The geotechnical and/or structural
failures can produce lateral deformations and/or settlements in the backland.
6
Table 1.1:
Selected Historical Earthquake Damage to Anchored
Sheet Pile Bulkheads
Lateral
Movement
Liquefaction
Noted?
Nagoya
Osaka
4m
3m
Yes
Yes
8.1
Nagoya
Osaka
4m
3m
Yes
Yes
May 22, 1960
8.4
Puerto Montt
1m
Yes
Alaska, USA
Mar 27, 1964
8.4
Whittier
Niigata, Japan
Jun 16, 1964
7.5
Niigata
+2 m
Yes
Tokachi-Oki, Japan
May 15,1968
7.8
Hachinohe
Hakodate
0.9 m
0.6 m
No
Yes
Nemuro-Hanto-Oki,
Japan
Jun 17, 1973
7.4
Hanasaki
Kiritappu
2m
negligible
Yes
Yes
Miyagi-Ken-Oki,
Japan
Jun 12, 1978
7.4
Ishinomaki
Yuriage
Sendai
1.2 m
1.2 m
negligible
Yes
Yes
Yes
Nihonkai- Chubu,
Japan
May 26, 1983
7.7
Akita
1.8 m
Yes
Kushiro-Oki, Japan
Jan 15, 1993
7.8
Kushiro
0.6 m
Yes
Guam
Aug 8, 1993
8.1
Cabras Island
Apra Harbor
0.6 m
0.6 m
Yes
Yes
Earthquake
Date
Magnitude
Tonankai, Japan
Dec 7, 1944
8.3
Nankai, Japan
Dec 21, 1946
Chile
Port
Yes
Limiting earthquake-induced deformations of waterfront retaining walls is a primary
seismic design issue at many ports, and subsequently, many ports have adopted
deformation-based seismic performance requirements. For example, the following
guidelines have been proposed for U.S. Naval facilities (Ferritto, 1997);
Design of anchored sheet pile retaining walls shall limit permanent displacement at the top
of the sheet pile to the following;
1) Less than 2.5 cm for a Level 1 earthquake (50% probability of exceedance in 50 years)
2) Less than 10 cm for a Level 2 earthquake (10% probability of exceedance in 50 years)
These standards are intended to insure that following a Level
1
(operating level)
earthquake motion, the earthquake-induced damages will be negligible and non-structural
7
b)
a)
c)
Figure 1.5: Geotechnical Failures of Anchored Sheet Pile Walls; a)
embedment failure; b) anchor failure; and c) global
failure
a)
b)
Figure 1.6: Structural Failures of Anchored Sheet Pile Walls; a)
anchor, tie rod, and/or whale system failure; and b)
sheet pile wall failure
and that the port operations will be unimpeded. They are also intended to insure that
following a larger, Level 2 (contingency level) earthquake motion, the damages will be
non-catastrophic and repairable.
8
In response to the liquefaction-induced damage that has occurred during recent
earthquakes, and the development of performance-based design requirements, many ports
are instigating programs to mitigate liquefaction hazards in waterfront areas. Common
remediation objectives for increasing the liquefaction resistance of soils include
densification, increased strength/stiffness, and/or improved soil drainage. These
improvement objectives are accomplished through many methods, including deep
dynamic compaction, vibro-compaction, stone columns, soil mixing, and many others.
Although the use of soil improvement methods is increasing, there are currently very few
tools available for designing the extent of ground treatment necessary to limit the
earthquake-induced damage of waterfront retaining structures. The only comprehensive
reference known to the author is provided by the Japanese Port and Harbour Research
Institute (PHRI, 1997), which provides guidance on soil improvement strategies at
waterfront facilities. The recommendations provided in this useful reference are largely
based on limit state analysis and model testing. The guidelines provided do not address
the wall deformations associated with varying design-level ground motions, a primary
concern in performance-based design.
Current "standard of practice" seismic design for anchored sheet pile bulkheads
involves using pseudo-static, limit equilibrium mechanics, developed for rigid retaining
walls by Okabe (1926) and Mononobe and Matsuo (1929). This method, which is based
on the Coulomb earth pressure theory, is frequently referred to as the Mononobe-Okabe
method. The seismic portion of the design is controlled by empirically determined
seismic coefficients, which are functions of the maximum ground accelerations. The
coefficients are used to estimate the seismically-induced inertial forces. Kitajima and
Uwabe (1979) have noted that even with increasing values of the design seismic
coefficients during the last fifty years, the percentage of earthquake-damaged sheet pile
bulkheads has not decreased. This could be due to several design factors, including:
inadequate maximum ground accelerations used in the wall design,
the use of maximum ground accelerations fails to account for the frequency and
duration of actual earthquake ground motions,
9
the relationships used to estimate the seismic coefficients from the maximum
accelerations may be deficient,
sheet pile walls are commonly designed using limit-equilibrium methods, but
most seismic failure criterion is based on permanent deformations,
sheet pile walls are flexible, but designed using rigid body mechanics, and
excess pore pressure generation and liquefaction is usually not accounted for,
therefore the actual active earth pressures may be much larger and the passive
earth pressures much smaller than the earth pressures used in design.
The Mononobe-Okabe method can be used to account for the presence of potentially
liquefiable soils, but only in a simplistic manner by decreasing the soil strength or
increasing the active earth pressures. The Mononobe-Okabe method is also restricted by
being limit-equilibrium based, and therefore not applicable for deformation-based
analysis.
There have been several recent additions to the Mononobe-Okabe seismic design
method. These methods may be divided into two categories; 1) procedures which include
the possibility of liquefaction, and 2) those that do not include the possibility of
liquefaction. The methods that do not include the possibility of liquefaction include
Dennehy (1985) and Gazetas et al. (1990); Neelakantan et al. (1992); and Steedman and
Zeng (1990). A method that attempts to account for liquefaction has been developed by
Towhata and Islam (1987). These recent methods have contributed to the knowledge of
seismic behavior and design of sheet pile bulkheads, but they are limited in their use as
design methods, as discussed in Section 4.4.
In summary, the prevalent issues for the deformation-based, seismic analysis of sheet
pile bulkheads includes the need to estimate lateral deformations for:
non-liquefiable soils,
potentially liquefiable soils,
varying design-level ground motions, and
bulkheads with partial soil improvement.
10
1.2
STATEMENT OF OBJECTIVES AND SCOPE OF WORK
Objectives
1.2.1
Primary objectives of this project were to determine the seismic vulnerability and
necessary remediation methods for flexible, waterfront retaining structures. Several
objectives were determined at the onset of this project that included; a) review the
extensive technical literature and establish a data base of case histories, b) identify
common failure modes and evaluate the applicability of current "standard of practice"
methods for the seismic design of anchored sheet pile bulkheads, c) perform numerical
soil-structure interaction studies, and d) develop design recommendations for the
performance-based seismic design of anchored sheet pile walls. The scope of work for
each of these objectives is outlined below.
1.2.1.1
Establish a Data Base of Case Histories
An extensive technical literature search was conducted to collect case histories on the
seismic performance of sheet pile bulkheads. The search was conducted intensely during
the first six months of the project, and to a lesser extent during the remainder of the
research. A review of the literature was used to evaluate the performance of sheet pile
bulkheads, and to determine the controlling variables used in design. Five of the case
histories collected during the literature search contained the necessary data for use in
validating the numerical model.
1.2.1.2 Evaluation of Current Methods
An extensive literature search was also conducted to determine the current "standard
of practice" design method, and the applicability of the design method against the failure
modes identified from the case histories. The search resulted in a literature collection of
11
the traditional methods used in the seismic design of sheet pile bulkheads, and recent
additions that account for limitations of the "standard of practice" design method.
1.2.1.3
Perform Analytical Soil-Structure Interaction Studies
Analytical soil-structure interaction studies were performed to model the behavior of
sheet pile retaining walls in liquefiable and improved soils. This objective utilized the
geomechanical-modeling program FLAC (Itasca Consulting Group, 1995). There were
two phases of this portion of the study, the first involved validating FLAC by modeling
case histories to assess the capabilities and accuracy of FLAC. Phase two utilized the
validated FLAC model to perform parametric studies of the seismic performance of sheet
pile bulkheads. The parametric studies were used to determine the influence of several
design and remediation factors, including:
embedment depth of the sheet pile wall,
length of the tie rod,
stiffness of the sheet pile wall,
penetration resistance of the backfill,
zone of soil densification, and
ground motion characteristics.
1.2.1.4 Development of Design Recommendations
The results of the analytical numerical analyses were used to develop an improved
seismic design procedure that incorporates all of the parametric study data, and includes
field performance data incorporated from the literature review. Design recommendations
were also developed that highlight the usage of a simple design chart and, importantly,
the assumptions and limitations of using the chart for design purposes.
12
1.2.2
Report Organization
The remaining chapters are organized as follows. Chapter 2 highlights the necessary
aspects of performing a liquefaction assessment for waterfront structures, and includes
aspects on field exploration, laboratory testing and liquefaction assessment. Chapter 3
highlights the current methods available and commonly used in soil improvement for
liquefaction remediation. Chapter 4 details the current design method for sheet pile
bulkheads, and also includes summaries of recent contributions to the design of sheet pile
bulkheads. Chapter 5 deals with the aspects of the numerical model used in the analytical
studies, and includes the results of the validation case histories. Chapter 6 presents the
results of the numerical parametric study and the recommended design procedures.
Chapter 7 provides a summary and conclusion of the research project and
recommendations for future work.
13
2
2.1
LIQUEFACTION ASSESSMENT OF WATERFRONT SOILS
INTRODUCTION
There have been several extensive reports indicating that the most significant source
of earthquake damage to waterfront structures has been due to the liquefaction of loose to
medium-dense saturated sands (Kitajima and Uwabe, 1979; Werner and Hung, 1982;
Werner, 1998). Liquefaction is the result of cyclic shearing of loose to medium-dense,
saturated granular soils. To determine the liquefaction susceptibility of soils, a
liquefaction assessment needs to be conducted. The common simplified method used to
perform a liquefaction assessment in the United States is the cyclic shear method, and is
described in the following sections of this chapter. Prior to the liquefaction assessment, it
is necessary to first complete the following:
1) Perform a Geotechnical/Geological Exploration
to provide an assessment
of the soil types, stratigraphy, location of the water table, site profile and
geometry, and hydraulic conditions. It may also necessary, depending on the
liquefaction assessment analysis, to perform in-situ field testing (SPT, CPT, shear
and pressure wave velocities, etc.) and collect samples for laboratory testing.
2) Determine the Soil Engineering Properties
through laboratory tests and
stress calculations. Laboratory tests for liquefaction assessment are conducted to
determine one or more of the following; density, cyclic shear resistance, damping
characteristics, soil structure, dynamic shear modulus, and grain size. It is also
necessary to determine the static soil stresses (vertical, horizontal and shear) with
depth. The sample disturbance that takes place during sample collecting,
transportation, and laboratory testing needs to be acknowledged and accounted
for in the interpretation of the test results. Descriptions on the determination of
soil engineering properties can be found in USACE (1984), Terzaghi et al.
(1996), and PHRI (1997).
14
3) Determine the Earthquake Motion(s)
using either simple estimation
procedures that are prescribed by regulatory bodies (e.g. FEMA, UBC, etc.), or
perform a site-specific deterministic or probabilistic seismic hazard analysis. A
complete description of these methods can be found in Dickenson et al. (1998)
and Kramer (1996).
2.2
CYCLIC SHEAR METHOD OF LIQUEFACTION ASSESSMENT
Several steps are involved in using the cyclic shear method for liquefaction
assessment; 1) determine the earthquake induced cyclic shear stresses, 2) determine the
cyclic shear stress required to initiate liquefaction, and 3) compare the results of steps 1
and 2 to assess the liquefaction susceptibility of the soil. These steps are described in
Sections 2.2.1 to 2.2.3. A short discussion on the estimation of liquefaction-induced
effects (residual excess pore pressure, post-liquefaction volumetric strain, and undrained
residual shear strength) is also provided in Section 2.2.4.
Determination of the Earthquake Induced Cyclic Shear Stresses
2.2.1
The shear stresses developed at any location within a soil deposit during an
earthquake appear to be due primarily to the vertical propagation of shear waves. There
are two common methods for estimating the earthquake induced cyclic shear stresses;
1) perform a simple empirical analysis, or 2) perform a site specific ground response
analysis.
2.2.1.1
Simplified Empirical Analysis
The simplified procedure developed by Seed and Idriss (1971, 1982) uses the concept
of vertically propagating shear waves producing cyclic shear stresses within the soil
column. The first step involves estimating the maximum acceleration within the soil
column at depth z (am.,z) from the maximum ground surface acceleration (am,), using
Equation 2.1 and the data plotted in Figure 2.1.
15
Amax z = amax rd
2.1
a max,z
rd
0.1
0.2
0.3
0.4
a max
0.5
0.6
0.7
0.8
0.9
1.0
Figure 2.1: Range of rd Values for Different Soil Profiles (after
Seed and Idriss, 1982)
The empirical data plotted in Figure 2.1 indicates a wide range between the upper
and lower bound values, especially below a depth of approximately 12 m. It
is
recommended that the average values (or for a conservative analysis, the upper bound
values) be used for depths between 0 and 12 m, and that a site specific dynamic ground
response analysis be conducted to estimate the maximum acceleration for depths
exceeding 12 m.
16
After the maximum acceleration values with depth are estimated, the earthquake-
induced cyclic shear stresses at the depth of interest (1-,,,,,) can be calculated using
Equation 2.2, where g is the acceleration due to gravity and o
is the total overburden
stress. The value of rm,Z is then converted to an equivalent uniform cyclic shear stress
(ray) to account for the transient time history behavior of the shear stresses. Seed and
Idriss recommend reducing z-,, by 65% to account for the irregular loading (Figure 2.2).
The value of rav,z is then estimated using Equation 2.3. The value of ra,,, is then
normalized by the effective overburden stress (o, ') to produce the cyclic stress ratio
induced by the earthquake (CSReq), which is given in Equation 2.4.
a max, z
MaXZ
2.2
VO
g
T
T
max
av
0.65 T
max
TIME
Figure 2.2: Earthquake Induced Shear Stresses (Seed and Idriss,
1982)
1- QV, Z
CSR
= 0 65 z- MaXZ
= av'
a vo
z
2.3
2.4
17
2.2.1.2 Site Specific Analysis
A more in-depth method of estimating the CSReq involves calculating anza,z or
i-max,z
directly using a dynamic ground response analysis. A ground response analysis is
conducted using a numerical computer model, such as the equivalent linear program
SHAKE (Schnabel et al., 1972) or a fully nonlinear dynamic program, such as DESRA2C (Lee and Finn, 1991) or SUMDES (Li et al., 1992). The advantages and disadvantages
of each program should be thoroughly examined before use.
2.2.2
Cyclic Shear Stress Required to Initiate Liquefaction
The cyclic shear stress ratio required to initiate liquefaction (CSRfidd) can be
determined from either laboratory or field tests. Each method is discussed in the
following sections, along with the advantages and disadvantages of each method.
2.2.2.1
Laboratory Tests
Laboratory tests are used to determine the number of loading cycles necessary to
produce liquefaction failure (N1) for various amplitudes of shear stress. Liquefaction
failure is usually defined as the point at which liquefaction is initiated, or a set amount of
strain is exceeded. It should be noted that the denser the material, the larger the values of
N1. Typical results of cyclic shear tests are shown in Figure 2.3. There are two different
laboratory tests commonly conducted to determine the cyclic stress ratio; the cyclic
simple shear and the cyclic triaxial shear. The cyclic shear strengths obtained from either
the simple shear or triaxial shear tests are normalized by the effective overburden
pressure to produce CSRSS and CSR,x, respectively. Because the cyclic simple shear and
the cyclic triaxial shear tests impose different loadings, the CSRSS and CSR,x are not
equivalent, but are related by the following equation;
CSRSS = Cr CSRIx
2.5
18
where recommended values of cr have been compiled in Table 2.1 as a function of the
static coefficient of earth pressure (K0).
0.4
CC
6,
w,.
0.3
cr,
-J
uj 0.2
UI
>- (1)
0 ISHINOMAKI PORT SAND Dr
0.
50.0%
HANASAIG PORT SAND Dr = 38.5%
1
0 KUSHIRO PORT SAND Di
51.3%
0
3
30
10
300
100
NUMBER OF WAVES NI
Figure 2.3: Typical Results of Cyclic Shear Tests (PHRI, 1997)
Table 2.1:
Values of CSR Correction Factor, c (after Seed, 1979)
Cr
Reference
Finn et al.
(1971)
Equation
Cr =
Ko = 0.4
Ko = 1.0
0.7
1.0
1+ K
2
varies
Seed and Peacock (1971)
Castro (1975)
cr =
0.55
0.72
1.0
2(1+ 21C0)
315
0.69
1.15
The cyclic simple shear and cyclic triaxial shear produce shear stresses in only one
direction, in contrast to the actual earthquake motions, which produce shear stresses in
different directions simultaneously. Boulanger and Seed (1995) conclude from laboratory
tests that the CSR for unidirectional simple shear tests was approximately 5 to 25% less
than multidirectional earthquake shaking. This is consistent with the recommendation of
19
Seed (1979), that the CSR for unidirectional simple shear tests be approximated as 10%
less than that for multidirectional shaking. Utilizing the work of these researchers, it is
concluded that multidirectional shaking reduces the liquefaction resistance of soils by an
average of approximately 10%. Therefore, the recommended laboratory CSR values are
related to the field CSR values by;
CSR field = 0.9 . CSR,s = 0.9.c, CSRt,
2.6
Cyclic laboratory tests are normally conducted on re-constituted samples, due to the
extreme difficulty in obtaining relatively undisturbed samples of cohesionless soils. The
re-constituted cyclic laboratory tests are able to reproduce the in-situ density and effective
confining pressures, but due to the depositional and historical environment of many
waterfront areas, there are several conditions that cyclic tests are unable to reproduce
accurately. These conditions include the soil fabric, past earthquake loading, overconsolidation ratio, lateral earth pressure coefficient, and the duration of sustained
pressure before testing. Relatively undisturbed sampling (e.g. in-situ ground freezing) is
necessary to confidently perform a liquefaction assessment using cyclic laboratory tests,
otherwise, caution should be taken in using the results of re-constituted laboratory
specimens.
2.2.2.2 Field Tests
The recommended method of characterizing the liquefaction resistance of a soil
deposit is based on the results of in-situ tests, due to the disturbance inherent in the
sampling and laboratory testing of cohesionless soils. The standard penetration test (SPT)
has historically been used for liquefaction assessments. Another method that is becoming
more common for liquefaction assessments is the cone penetration test (CPT). These
methods are discussed in detail in the following sections. Information on correlations
between other field test results (e.g. shear wave velocity, dilatometer) and the cyclic stress
ratio can be found in Kramer (1996).
20
Standard Penetration Test (SPT). Seed and Idriss (1982) noted that the specifics of
the SPT testing method have a very large influence on the measured results. Therefore, it
is necessary to standardize the measured results of the SPT test (N blows/30 cm). Seed et
al. (1985) provide a method for standardizing SPT results that has been used extensively
since it was first introduced. The resulting standardized SPT results are presented as
(A/1)60, in which the subscript 1 represents being normalized for an overburden pressure of
1 kg/cm2, and the subscript 60 represents being normalized for an applied energy of 60%
of the theoretical maximum free-fall energy.
Seed et al. (1979) plotted standardized SPT values versus the cyclic shear stress
ratios for earthquakes of magnitude 7.5 (CSRm=75) for locations with and without the
occurrence of liquefaction (Figure 2.4). This plot provides a very practical method of
estimating the cyclic stress ratio necessary to initiate liquefaction using normalized SPT
values and the percentage fines of the sandy soil. It should be noted that the CSRM=7.5
values in the figure are for earthquake magnitudes of 7.5 only, and for level sites that
have relatively shallow liquefaction susceptible soils. Because variations in earthquake
magnitudes lead to an increased number of earthquake induced shear cycles, the CSRA1=75
should be modified to account for different earthquake magnitudes. The CSRm=75 should
also be modified to account for the presence of large overburden pressures and for static
shear stresses (my). The CSRfieid can be calculated from the CSRm=75 with the following
equation;
CSR field =
Ka Ka MSF CSRA1,7.5
2.7
where Ka is the correction for the static shear stresses on horizontal planes in the soil
deposit (Figure 2.5), K0- is the correction for large overburden pressures (Figure 2.6), and
MSF is the correction for different magnitude earthquakes (Table 2.2). The MSF factor is
used to account for the variations in the duration (and corresponding cycles of loading)
between earthquake motions of different magnitude.
21
0.6
0
Percent fines = 35
<5
15
0.5
I
J
A
0.4
CSRm = 7.5
r
A
0.3
0.2
Fines content 25%
Modified Chinese code proposal (clay content
0.1
Liquefaction
Marginal
Liquefaction
0
0
A
Pan-American data
Japanese data
A
Chinese data
0
10
0
20
30
40
(NOso
Figure 2.4: Empirical Relationship Between the Cyclic Stress
Ratio Initiating Liquefaction and (A11)60 Values for
Silty Sands in M=7.5 Earthquakes (after Seed et al.,
1979)
Table 2.2:
Magnitude Scaling Factors, MSF (Arango, 1996)
Earthquake Magnitude
MSF
5%)
No
Liquefaction
5.50
6.00
7.00
7.50
8.00
8.25
3.00
2.00
1.25
1.00
0.75
0.63
50
22
rho
I
oC
Figure 2.5: Relationship between a and KG, (Seed & Harder, 1990)
2.0
1.2
3.0
4.0
6.0
5.0
8.0
7.0
°
\
0
\0
1.0
0
0.8
0
... N.
*X".....
6 °o
..4.1
......
V .....
K,
,
.....:,
.......
0.6
0
V
C1
0.4
A
V
0
0
0.2
0
0
0
0
SHEFFIELD DAM SHELL
UPPER SAN LEANDRO DAM SHELL
LONER SAN FERNANDO DAM SHELL
UPPER SAN FERNANDO DAM SHELL
LOS ANGELES DAM SHELL
ERRiS DAM SHELL,
AC
o
0
H.,..
N.. ......
FAIRMONT DAM
LAKE ARROWHEAD DAY
V
0
..,.
...H.,
"...
NH., ......
......,
V
95,100%
SARDIS DAM SHELL
SARDIS DAM FOUNDATION
THERIAAL110 AFTERRAY DAY FOUNDATION
THERMALITO FOREMAY DAM FOUNDATION
ANTELOPE DAM IMPERVIOUS MATERIAL
FORT PECK DAM SHELL
SR, NO, TO, PO%
SACRAMENTO RIVER SANG, D.
MONTEREY 0 SAND, 0. SO%
V
REID BEDFORD SAND, Dr.. 40, O%
NEw JERSEY 'MCNEILL, FPI RC OS%
1.0
40
60
5.0
2.0
3.0
EFFECTIVE CONFINING PRESSURE (tsf ) or (kW
7.0
Figure 2.6: Relationship between the Effective Vertical Stress and
K, (Seed & Harder, 1990)
8.0
23
Cone Penetration Test (CPT). The use of the CPT has increased dramatically in
recent years, due to several advantages over the SPT, including rapid and inexpensive
testing, and it also provides a continuous soil profile with depth. It is also possible to
enhance the CPT with the addition of pore pressure transducers and instruments to
measure the seismic wave velocity. The cone penetration resistance (qc) and the shaft
friction resistance (fs) are measured continuously during the testing. Stark and Olson
(1995) have proposed a relationship between liquefaction field data and CPT penetration
resistance (Figure 2.7). The seismic shear stress ratio is equivalent to CSRA,f=7.5. The value
of qa is the qc value has been normalized by a vertical effective overburden stress of
approximately 100 kPa. The CSRm=75 value can then be substituted into Equation 2.7 to
estimate the CSRfierd
0.6
0.5
0.4
0.3
02
0.1
0
0
5
10
15
20
25
30
CORRECTED CPT TIP RESISTANCE, cio (MPa)
Figure 2.7: Proposed CPT-Based Liquefaction Curves Based on
Field CPT and Liquefaction Data (Stark & Olson,
1995)
24
2.2.3
Evaluation of Initiation of Liquefaction
After the cyclic stress ratio caused by the earthquake (CSReq) and the cyclic stress
ratio necessary to induce liquefaction (CSRfi ) are determined at the depths of interest,
the potential for liquefaction can be evaluated. A graphical representation is the simplest
method of determining the liquefaction potential, and is accomplished by plotting the
CSReq and CSRfierd versus depth (Figure 2.8). Liquefaction is likely to occur at any depth
where the CSReq exceeds the CSRfield A factor of safety against liquefaction (FS1) can also
be evaluated at specific depths with the following equation;
FS i=
CSR field
2.8
CSReq
CSR
Zone of
Liquefaction
0a)
CSRfiwd
CSReq
Figure 2.8: Graphical Plot on the Evaluation of the Initiation of
Liquefaction
2.2.4 Effects of Liquefaction
The liquefaction of soils is caused by the generation of excess pore pressures, which
result in decreased effective stresses, soil stiffness and soil strength. Volumetric strains
due to the dissipation of generated excess pore pressures and consolidation of the
25
densified soil are also an effect of liquefaction. It should be noted that excess pore
pressures are still generated even with FS1 values in excess of one.
Marcuson and Hynes (1990) developed a figure relating the excess pore pressure
ratio (ru) to the factor of safety against liquefaction for both gravel and sand (Figure 2.9).
The excess pore pressure ratio is the relationship between the residual excess pore
pressure (uexcess) and the effective overburden pressure (cry° ') given by;
ru =
U excess
VO
2.9
'
1.0
;=
0.8
Gravel
Sand
0.6
0.4
0.2
0
10
1.2
1:1:"Zzi
2.2
11
2.1.0
1.6
FACTOR OF SAFETY At:MIST L1OUEFAACTION. FSL
1.4
sie,
2.4
'6
Figure 2.9: Relationship Between Residual Excess Pore Pressure
and Factor of Safety Against Liquefaction for LevelGround Sites (after Marcuson and Hynes, 1990)
Several researchers have evaluated the post-liquefaction volumetric strain due to the
dissipation of excess pore pressures. An approach that utilizes the FS/ was developed by
26
Ishihara and Yoshimine (1992), and relates the factor of safety to the post-liquefaction
volumetric strain for clean sands of various relative densities (Dr), given in Figure 2.10.
2.0
1.8
1.6
1.4
1.2
.3
FS,
1.0 -3.5%
4%
0.8
D, =40
0, 60
0.6
0.4
6%
( N,
(NI = 20,
?max = 10%
N, = 6
=
q
qct = 45
Dr= 30
N, = 3 .
qc, =
= 60
= 80 )
/
8%
D, = 50
13, =
gc, = no)
80
= 25,00 = 147)
0.2
Dr = 90%
N, = 30
200kg cm2
1,0
20
30
40
50
Post-liquefication volumetric strain, Ev(%)
Figure 2.10: Estimation of Post-Liquefaction Volumetric Strain for
Clean Sands (after Ishihara and Yoshimine, 1992).
The most common method of estimating the residual undrained shear strength was
developed by Stark and Mesri (1992) and utilizes SPT results corrected for the percentage
of fines contained in the soil. The relationship between the clean sand SPT ((Aid 6o-cs) and
the undrained residual shear strength (normalized by the effective overburden pressure
27
and plotted as critical) is given in Figure 2.11. The clean sand SPT is calculated using the
data from Table 2.3 and the following equation;
(N1)60-CS = (N1)60
Ncorr
2.10
Recommended Fines Correction for Estimation of
Residual Undrained Strength (Stark & Mesri, 1992)
Table 2.3:
Percent Fines
Ncorr (blows/30 cm)
0
0.0
2.5
4.0
5.0
6.0
6.5
7.0
10
15
20
25
30
35'
0.6
0.5
FIELD CASE HISTORIES
O CYCUC TRIAXIAL CRITICAL STRENGTH
0 CYCUC TRIAXIAL CRITICAL STRENGTH AT 100 CYCLES
CYCLIC TRIAXIAL AND BLOWCOUNT FROM EON. (6)
v SIMPLE SHEAR AND TORSIONAL SHEAR TESTS
0.4
Su (YIELD. MOB)
0.3
CVO
0.011 (N
1
)6o_cS
M7.5
Su (CRITICAL)
(:)"
0.2
z
0.1
0
0.0055 04
1
1 60-CS
VO
:
4
8
12
16
20
24
28
32
EQUIVALENT CLEAN SAND SPT BLOWCOUNT, (N1)60
36
-CS
Figure 2.11: Undrained Critical Strength Ratio vs. Equivalent
Clean Sand Blow Count (Stark and Mesri, 1992)
40
28
3
3.1
MITIGATION OF LIQUEFACTION HAZARDS
INTRODUCTION
If a liquefaction assessment (Chapter 2) demonstrates that liquefaction is likely to
occur in the waterfront soils during the specified design earthquake, and the liquefaction-
induced damage exceeds the performance criteria, mitigation strategies should be
evaluated. Mitigation strategies should be designed to limit the earthquake-induced soil
displacements and settlements within acceptable levels. There are two categories of
mitigation, soil improvement and structural enhancement. Only the soil improvement
techniques are addressed herein.
3.2
TECHNIQUES FOR MITIGATING LIQUEFACTION HAZARDS
Common remediation objectives of soil improvement are to increase the liquefaction
resistance of soils through densification, increased strength, and/or improved soil
drainage. Table 3.1 presents the most common remediation methods, and the principle,
suitable soil conditions, effective treatment depth and relative costs for the tabulated
methods.
Many of the tabulated remediation methods have limited the occurrence of
liquefaction during recent earthquakes. Sand compaction piles, vibro-compaction, sand
drains and surcharge soil mitigation techniques were successful used in preventing
liquefaction in Kobe Port during the 1995 Hyogoken Nanbu Earthquake (Yasuda et al.,
1996). Sand compaction piles and gravel drains were successfully used in the Port of
Kushiro prior to the 1993 Kushiro Oki Earthquake (Iai et al., 1994). Hayden and
Boulanger (1995) describe the use of compaction grouting of liquefiable soils at a number
of different sites. Baez and Martin (1992) provide an evaluation on the use of stone
column (vibro-compaction) techniques for liquefaction mitigation. Egan and others
(1992) describe the installation of stone columns at the Port of Oakland for the mitigation
of liquefaction hazards.
29
Table 3.1:
Liquefaction Remediation Measures (after Ferritto,
1996)
1)
Most Suitable
Soil Conditions
or Types
Maximum
Effective
Treatment Depth
Densification by vibration;
liquefaction-induced settlement
and settlement in dry soil under
overburden to produce a higher
density.
Saturated or dry
clean sand; sand.
20 m routinely
(ineffective above 34 m depth); > 30 m
sometimes;
vibrowing, 40 m.
Moderate
Densification by vibration and
compaction of backfill material
of sand or gravel.
Cohesion less
soils with less
than 20% fines.
> 20 m
Low to
moderate
Method
Principle
Vibratory
Probe
a) Terraprobe
b) Vibrorods
Relative
Costs
c) Vibrowing
2)
Vibrocompaction
a) Vibrofloat
b) VibroComposer
system.
3)
Compaction
Piles
Densification by displacement of
pile volume and by vibration
during driving, increase in
lateral effective earth pressure.
Loose sandy soil;
partly saturated
clayey soil; loess.
> 20 m
Moderate
to high
4)
Heavy tamping
(dynamic
compaction)
Repeated application of highintensity impacts at surface.
Cohesionless
soils best, other
types can also be
improved.
30 m (possibly
deeper)
Low
5)
Displacement
(compaction
grout)
Highly viscous grout acts as
radial hydraulic jack when
pumped in under high pressure.
All soils.
Unlimited
Low to
moderate
6)
Surcharge or
buttress
The weight of a
surcharge/buttress increases the
liquefaction resistance by
increasing the effective
confining pressures in the
foundation.
Can be placed on
any soil surface.
Dependent on size
Moderate
if vertical
drains are
used
Relief of excess pore water
pressure to prevent liquefaction.
(Wick drains have comparable
permeability to sand drains).
Primarily gravel drains;
sand/wick may supplement
gravel drain or relieve existing
excess pore water pressure.
Permanent dewatering with
pumps.
Sand, silt, clay.
Gravel and sand >
30 m; depth limited
by vibratory
equipment; wick, >
45 m
Moderate
to high
Penetration grouting-fill soil
pores with soil, cement, and/or
Medium to coarse
sand and gravel.
Unlimited
Lowest of
grout
methods
7)
Drains
a) Gravel
b) Sand
c) Wick
d) Wells (for
permanent
dewatering)
8)
Particulate
grouting
clay.
of
surcharge/buttress
30
Table 3.1
(Continued)
Most Suitable
Soil Conditions
or Types
Maximum
Effective
Treatment Depth
Relative
Costs
Solutions of two or more
chemicals react in soil pores to
form a gel or a solid precipitate.
Medium silts and
coarser.
Unlimited
High
Penetration grouting-fill soil
pores with lime
Medium to coarse
sand and gravel.
Unlimited
Low
Stabilizing chemical moved into
and fills soil pores by electroosmosis or colloids in to pores
by electrphoresis.
Saturated sands,
silts, silty clays.
Unknown
Expensive
12) Jet grouting
High-speed jets at depth
excavate, inject, and mix a
stabilizer with soil to form
columns or panels.
Sands, silts, clays.
Unknown
High
13) Mix-in-place
piles and walls
Lime, cement or asphalt
introduced through rotating
auger or special in-place mixer.
Sand, silts, clays,
all soft or loose
inorganic soils.
> 20 m (60 m
obtained in Japan)
High
14) Vibro-
Hole jetted into fine-grained soil
and backfilled with densely
compacted gravel or sand hole
formed in cohesionless soils by
vibro techniques and compaction
of backfilled gravel or sand. For
grouted columns, voids filled
with a grout.
Sands, silts, clays.
> 30 m (limited by
vibratory
equipment)
Moderate
Small-diameter inclusions used
to carry tension, shear, and
compression.
All soils.
Unknown
Moderate
to high
Shock waves and vibrations
cause limited liquefaction,
displacement, remolding, and
settlement to higher density.
Saturated, clean
sand; partly
saturated sands
and silts after
flooding.
> 40 m
Low
9)
Method
Principle
Chemical
grouting
10) Pressure
injected lime
11) Electrokinetic
injection
replacement
stone and sand
columns
a) Grouted
b) Not grouted
15) Root piles, soil
nailing
16) Blasting
3.3
DESIGN OF SOIL MITIGATION
The design of soil mitigation strategies involves investigating the cost/benefit ratio,
the performance, and the effect on adjacent structures of the mitigation technique(s). The
mitigation methods listed in Table 3.1 may be placed into one of three different
categories; compaction, drainage, and cementation. The design for each of these
31
remediation categories is briefly discussed below. A more in-depth discussion on the
design of soil mitigation strategies and procedures may be found in PHRI (1997).
Compaction. Compaction remediation methods mitigate liquefaction hazards through
soil densification with vibration or impact (examples include methods 1, 2, 3, 4 and 14
from Table 3.1). Compaction methods are more suitable for use in saturated, cohesionless
soils with a limited percentage of fines. Compaction remediation methods cause noise
and vibration during installation, and also increase horizontal earth pressures against
adjacent structures. The increase of horizontal earth pressures is the major disadvantage
of compaction methods in close proximity to retaining walls, and must be accounted for
during the design of a soil improvement program. The major advantage of compaction
methods for soil improvement is the relatively low cost/benefit ratio. The degree of
compaction that is necessary can be evaluated using penetration resistances that have
been back-calculated from an acceptable factor of safety against liquefaction (Chapter 2).
Drainage. Drainage remediation methods mitigate liquefaction hazards by enhancing
the rate of excess pore pressure dissipation. The most common methods of drainage
remediation are through the use of gravel, sand or wick drains. Drains are suitable for use
in sands, silts or clays. One of the greatest advantages of drains is that they induce
relatively small horizontal earth pressures during installation. Therefore, they are suitable
for use adjacent to sensitive existing structures. In the design of drains, it is necessary to
select a suitable drain material that has a coefficient of permeability substantially larger
than the in-situ soils.
Cementation. Cementation remediation methods mitigate liquefaction hazards
through increased soil strength. The soil strength is increased with the addition of a
cementatious material (i.e. cement, grout, lime, chemicals, asphalt). Cementation
techniques (methods 5, 8, 9, 10, 12, and 14 from Table 3.1) can be used with any type of
soil. Cementation methods are advantageous because the installation methods are
relatively quiet and induce relatively small vibrations compared with compaction
methods. The induced horizontal earth pressures are also less with cementation methods
than with compaction methods, but are greater than induced pressures from drainage
32
methods. The disadvantage of cementation methods is their relatively high cost/benefit
compared with compaction and drainage methods.
The relative performance of a specific improvement method is also of concern in the
design of a mitigation program. Experience has demonstrated that compaction and
cementation techniques reduce the liquefaction susceptibility of soils to a larger extent
than drainage methods.
The design of a mitigation strategy usually includes the combination of two or more
improvement techniques. An example is provided in Figure 3.1, in which a drainage
method was used adjacent to the sheet pile wall to limit the induced horizontal earth
pressures, and compaction techniques were used in front of the wall and behind the
anchor, likely to reduce the economic costs and for the better performance offered by
compaction methods in increasing the liquefaction resistance of the soil.
17m
20m
25m
+4.00
H W.L* 2.00m
L.W1,+ 0.00m
STEEL PIPE SHEET PILE
0711. /
SEABED
-corn
5
20.00m
23m
-21 .5m
Figure 3.1: Example Remediation Design for a Sheet Pile Wall
(PHRI, 1997)
The influence on existing structures during soil improvement is a primary design
consideration. The construction of mitigation methods may lead to increased horizontal
earth pressures, which can result in increased bending stresses in sheet pile walls.
Mitigation methods may also induce excess pore pressures and vibration, which will also
33
affect the sheet pile wall and adjacent structures. Compaction techniques have the largest
influence on adjacent structures, followed by cementation and drainage methods, which
have the least amount of influence.
3.4
DESIGN FOR THE AREA OF SOIL MITIGATION
The Japan Port and Harbour Research Institute (PHRI, 1997) has produced one of the
few design guidelines that exist for specifying the extent of soil improvement adjacent to
sheet pile bulkheads. This recommended extent of treated soil is shown in Figure 3.2 for a
sheet pile wall and in Figure 3.3 for sheet pile wall anchors. The area of soil improvement
indicated was developed primarily to limit the effect of the propagation of excess pore
pressures. It was noted by PHRI (1997) that during shaking table tests, the area enclosed
by the 30° triangle exhibited unstable characteristics, and therefore led to the
recommended design guidelines.
IMPROVEMENT AREA
LIQUEFACTION
LAYER
NONLIQUEFACTION LAYER
Figure 3.2: Improvement area for sheet pile walls (PHRI, 1997)
There is much uncertainty in the effectiveness of varying ground treatment strategies
for limiting lateral deformations of sheet pile bulkheads due to the lack of data on the
performance of improved soil sites during design-level earthquakes. A limited amount of
research has been performed in evaluating seismically-induced lateral deformations of
34
improved soil sites. The guidelines provided by PHRI, though not deformation-based,
provide an applicable method of estimating the necessary extent of ground treatment that
is based on "state-of-the-art" numerical and laboratory modeling.
IMPROVEMENT AREA
AS I V E
FAILURE
SURFACE
DURING
EARTHQUAKE
ACTIVE FAILURE. SURFACE
DURING EARTHQUAKE
LIQUEFACTION
LAYER
NONLIQUEFACTION LAYER
(a) FOR ANCHOR PLATE
IMPROVEMENT AREA
2.,13
211111111.1111illai'lliV
30°
PASSIVE FAILURE
SURFACE
DURING
EARTHQUAKE
30.
NONLIQUEFACTION LAYER
LIQUEFACTION
LAYER
1.1: POINT OF CONTRAFLEXURE
(b) FOR ANCHOR PILE
IMPROVEMENT AREA
LIQUEFACTION
LAYER
AV
NONLIQUEFACTION LAYER
(c) FOR BATTER ANCHOR PILES
Figure 3.3: Improvement Area for Sheet Pile Wall Anchors
(PHRI, 1997)
35
4
4.1
SEISMIC ANALYSIS AND DESIGN OF SHEET PILE BULKHEADS
INTRODUCTION
The static design of flexible retaining walls involves using limit-equilibrium methods
that have been developed for rigid walls. An addition to the rigid-body design procedures
has been made to account for the reduced moment observed in flexible walls (Rowe,
1952). The static design procedures are very well established and will not be discussed
herein. For a complete description of static design, the reader is referred to Terzaghi
(1954), United States Steel (1975), Arbed (1991), Ebeling and Morrison (1993), or
USACE (1994).
The "standard of practice" design method for incorporating earthquake forces on
flexible retaining walls utilizes pseudo-static earth pressures. Pseudo-static pressures are
a function of the estimated maximum horizontal acceleration, and an addition to the static
Coulomb earth pressures. The pseudo-static design of flexible walls also uses Rowe's
(1952) flexible wall moment reduction, and for potentially liquefiable soils, empirical
estimations of pore pressure generation are also included.
There are several limitations of pseudo-static design, which includes the assumptions
inherent in Coulomb's earth pressure theory (e.g. planar failure surface, soil is isotropic
and homogeneous, the failure wedge behaves as an elastic-perfectly plastic rigid body,
etc.); 1) the soil profiles must be simple (only one soil layer) and uniform into the
backland; 2) the transient seismic motion is modeled with a single parameter; 3) very
simplified inclusion of pore pressure generation; and, 4) pseudo-static design is not
deformation-based. The "standard of practice" pseudo-static design procedures are
summarized in Section 4.2. There have been several recent modifications, or different
approaches to the pseudo-static design of flexible walls that can be categorized into either
design methods for potentially liquefiable or non-liquefiable soils. Several of these
methods are summarized in Section 4.3
36
4.2
PSEUDO-STATIC DESIGN
The standard of practice for earthquake design of flexible retaining walls utilizes
pseudo-static earth pressures calculated from a modified representation of the Coulomb
earth pressure theory developed by Okabe (1926) and Mononobe and Matsuo (1929).
This method is commonly referred to as the Mononobe-Okabe method. The pseudo-static
earth pressures are represented as static earth pressures increased by the earthquake
induced inertia of the soil. The inertial forces are estimated using the horizontal (kh) and
vertical (10 seismic coefficients. The seismic coefficients are empirical functions of the
estimated maximum acceleration at the bulkhead. One method commonly used to
estimate the horizontal seismic coefficient is to take 65% of the maximum acceleration
(Ebeling and Morrison, 1993). Another common method for estimating kh was developed
by Noda et al. (1975), and is given by Equation 4.1, utilizing the data of Figure 4.1.
kh=a
0.2g
g
4.1
1(a) 3
3
g
a> 0.2g
where a is the maximum earthquake acceleration and g is the acceleration due to gravity.
The figure developed by Noda and others (1975) has been updated by Nozu et al. (1997),
and is shown in Figure 4.2, and includes a more recent assessment of the seismic
performance of retaining structures.
There has been very little research into the effect of 1c, on the pseudo-static earth
pressures, even though the vertical acceleration values can equal or exceed the horizontal
values, especially in epicentral regions (Ebeling and Morrison, 1993). The difference
between the upward and downward vertical accelerations also needs to also be addressed
during design. Ebeling and Morrison recommend that a vertical seismic coefficient be
used for anchored sheet pile walls when the horizontal seismic coefficient exceeds 0.05g.
37
0.3
th-a/s
=-1-(g
)I
3
t23
26
/7(19
15
I
8(19
3::11 2
' 9:134 3
?
2
Al
17.(1964)
...".
it t 1 Pt.Mont t
33 1 /1'24t5(1973)
731,1
,.
0.1
e
4 i
.*
A !..3.'...29..
.121
"...14
..._
"
''
27
t (195?) )'
0.2
24
28(1935)
21'1)1,13.5a/8
2
719
123
716
18
28(1930)
s Gravitational Acceleration
Ground Accel
d Acceleration
OD
0
500
400
300
200
100
Ground Accelerationgals
Figure 4.1: Relation between Seismic Coefficient and Ground
Acceleration (Noda et al., 1975)
035
I
0.30
0.25
i
0.20
I
VV
!
5
i
e
AvA
0
0
.
AI
1 v°
I
U)
i
9
a)
A
!
I
i
1
$
i
i
el
It. °e
161,1
A
cI
tibia j 'qv
A
1
I
3
i
bti
,
flk!
0.1
h=( ag )3
1
Ica
AZ: Ai 1
v1
---t-
I
i
i
I
1
1
1
0.0
0.
Al
A
I
Upper and lower limits of seismic coefficient estimated from
stability analysis of damaged quaywalls without liquefaction.
Lower limits of seismic coefficient of 7 quaywalls estimated from
stability analysis of damaged quaywalls.
i
100
ii
1
1
1
I
1
I
200
300
400
500
Peak Ground Acceleration for SMAC-type Accelerograph (gal)
Figure 4.2: Seismic Coefficient Estimated from Damaged
Quaywall (after Nozu et al., 1997)
600
38
They do not, however, recommend any relationships or values for kv. The vertical seismic
coefficient is often neglected in design.
The Mononobe-Okabe method is a modification of the static Coulomb earth pressure
theory and the resultant forces are the combined static and dynamic forces. The resulting
dynamic earth pressure forces and angles are defined in Figure 4.3 and given in Equations
4.2 through 4.8 for the passive and active cases of dry, uniform soils.
Pee = Ka,
Ppe = Kpe
where:
1
r
[y t (1 1 c v)1H2
4.2
1
v1(1-k)]H2
4.3
P = dynamic plus static earth pressure force
K=
lc,
=
=
H=
coefficient of dynamic earth pressure
total unit weight of the soil
vertical seismic coefficient
height of the wall
Figure 4.3: Definition Sketch for Pseudo-Static Design Parameters
39
The coefficients of earth pressures are given by the following;
ae
cos2 (0
=
cos coo +
0[1
4.4
2
sin(q$
+ Ilsin(fb +
99)
cos(8 + (o)cosp
cos2 (0
Kpe
V)
4.5
2
cos co cos(8
where:
V)
0[1 ilsn°
4-
sin(Ch
fi
CI))
cos(8 + go)cosfl
K=
active and passive dynamic earth pressure coefficient
angle of internal friction for the soil
seismic inertia angle
= angle of friction between the soil and wall
slope of the backfill
/6
The seismic inertia angle is given by;
= arctan
[
4.6
khk, )1
The angle of the failure plane to the horizontal is given by the following;
a=0
co + arctan[
tan(0 co /6)+
1 + [tan(8 + yokan((
P)+ cot(0 P)A1
ape = 0 + p arctank
tan(fb
[tan(8 + gokan(0
/6)+ C2
+ /6)+ cot(93
4.7
4.8
P))]]
40
where;
= Vtan(0
/3)[tan(cb
c2 = Vtan(0
p +13){tan(0
co M+ cot(q
+ tan(8 + Ocot(g)
co)]
co + ,(3)+ cot(q$ 41+ tan(8 + yo)cot(0 (0)]
These equations are applicable for dry backfill only. If the backfill is saturated, it is
necessary to modify the above equations (as described in the following sections),
depending on whether the backfill soils are potentially non-liquefiable or liquefiable.
Non-Liquefiable Soil
4.2.1
The following equations are used for non-liquefiable soils, and they are dependent on
the behavior of the pore water relative to the soil matrix. If the hydraulic conductivity of
the soil is less than 1x10-3 cm/sec, the pore water is assumed to be restrained and if the
"3
hydraulic conductivity is greater than 1x10 cm/sec, the pore water is assumed to be free.
The term restrained is used to define the condition where the pore water moves with the
soil skeleton and there is no fluid flow, and the term free is the condition where the pore
water moves relative to the surrounding soil during shaking. These conditions do not
affect the generation of excess pore pressures, and are only used to approximately
describe the pore water behavior during an earthquake.
4.2.1.1
Restrained Water Case
For submerged backfill with restrained water and without excess pore pressure
generation the following equation is commonly used to determine an effective horizontal
seismic coefficient;
71 7,
nh'=--mh
Yb
4.9
41
where:
total unit weight of the soil
Yb = buoyant unit weight of the soil
y,
=
The effective horizontal seismic coefficient (kh') should then be substituted for kh in
Equation 4.6. It is also necessary to use yb in Equations 4.2 and 4.3, and to include the
hydrostatic water forces on the wall.
4.2.1.2 Free Water Case
For the case of free pore water, Equations 4.10 and 4.11 should be used to estimate
the effective seismic coefficient and hydrodynamic force.
k h ' = 21 kh
4.10
Yb
where:
yd = dry unit weight of the soil
Yb = buoyant unit weight of the soil
If Equation 4.10 is used to calculate Equation 4.6, it is necessary to use yb in
calculating Pa, and Ppe. Because the water is now treated as acting independently from the
soil, it is necessary to include the inertia force of the water within the backfill. The inertia
forces from the free water are estimated using the following relationship developed by
Westergaard (1931). The resultant P,, is assumed to act at 0.4(H+D) above the base of the
wall.
P, = ukhr (H + D)2
where:
Pi = inertial water pressure force
= horizontal seismic coefficient
y, = unit weight of the water
H = wall height
D = depth of embedment
kh
4.11
42
The location of the resultant earth pressure forces from the Mononobe-Okabe method
cannot be determined from the above relationships. To determine the point of action for
the resultant forces, the following procedures developed by Seed and Whitman (1970) are
commonly used. Equation 4.12 (Seed and Whitman, 1970) provides the dynamic active
earth pressure force (Pae) as the sum of the static active earth pressure force (Pa) and the
dynamic active earth pressure force increment (AP,e);
4.12
Pae = Pa + APae
The static analysis for a uniform soil deposit, gives Pa acting at one third the total
wall height above the bottom of the wall. Seed and Whitman (1970) present APae as
acting at 0.6 times the total wall height above the bottom of the wall. Utilizing the known
points of action and Equation 4.12, Equation 4.13 was derived by Seed and Whitman to
determine the point of action for Pae above the bottom of the wall. A definition sketch is
given in Figure 4.4.
(H +
Y=
pa
APae[0.6(H + D)]
3
4.13
Pae
APae
g
0.6(H+D)
1/3(H+D)
Figure 4.4: Definition Sketch for the Location of the Resultant
Dynamic Active Earth Pressure
43
9.2.2 Liquefiable Soil
A potentially liquefiable soil is defined as a soil having the possibility of excess pore
pressure generation during cyclic shaking. The common method for representing excess
pore pressure generation in design was given by Equation 2.8 (ru = uexcessl crvo), where ru
represents the excess pore pressure ratio. Full liquefaction is defined as the state when the
excess pore pressures equal the effective overburden pressure, or rt, is equal to unity.
Potentially liquefiable soils will behave in one of three ways during dynamic excitation;
1) no excess pore pressure generation will occur (ru= 0),
2) partial excess pore pressures will be generated, but full liquefaction will not
occur (0 < ru < 1), or
3) full liquefaction will occur (ru = 1).
There is also the question of whether the pore water will flow freely or be restrained
during the cyclic loading, similar to the non-liquefaction case. The different procedures
are described below.
4.2.2.1
Restrained Water Case
There are two ways to analyze the restrained water case with excess pore pressure
generation using simple analytical procedures. The first is to have the excess pore
pressures reduce the unit weight of the soil and increase the unit weight of the water, until
full liquefaction, where the unit weight of the water equals the summation of the unit
weights of the soil and water. The effective unit weight of the soil and water are given by;
7 6' = 7 b(1
4.14
ru)
Twt= 7w + I b
ru
4.15
44
It is then necessary to compute an effective horizontal seismic coefficient (Equation
4.16) and corresponding seismic inertial angle (Equation 4.6).
'ch.= 11--, kh
4.16
Yb
The effective unit weights should be substituted in all preceding calculations, in
place of the soil and water unit weights. As rz, approaches unity, m' approaches zero, and
7w' becomes the total weight of the soil and water combined. Therefore, during full
liquefaction the soil-water matrix is treated as a heavy fluid. When full liquefaction
occurs, the unit weight of the heavy fluid (70 should be used with Equation 4.11 to
determine the dynamic fluid pressure.
The second method of including pore pressure generation involves decreasing the
effective angle of internal friction of the soil, as given in Equation 4.17. Ebeling and
Morrison (1993) describe this procedure in detail, and based on their calculations, it is not
recommended that this procedure be used because it over predicts Pae, notably when kh
approaches zero.
tan 01,q = (1
ru) tan 0'
4.17
4.2.2.2 Free Water Case
If the water acts independently from the soil, the pressures are due to the thrust from
the soil, and the thrust from the water. The thrust from the soil is estimated by using an
effective horizontal seismic coefficient (Equation 4.18), calculated from the effective soil
unit weight (Equation 4.14) and substituted into Equation 4.6.
=L-1 ,kh
kh'
b
4.18
45
The thrust from the water is estimated using Equation 4.11 with 7w, and the static
water pressure using 7w' from Equation 4.15.
4.3
RECENT ADVANCEMENTS IN FLEXIBLE WALL DESIGN
There have been numerous recent advancements in the design of flexible walls that
either enhance the pseudo-static design methods that propose alternative methods of
analysis. These methods are used for two different potential cases, liquefiable and nonliquefiable soils. The following sections summarize these recent advancements for both
cases, and the sub-sections are differentiated by the lead researchers.
4.3.1
Non-Liquefiable Soils
Neelakantan et al. Neelakantan et al. (1992) developed a balanced seismic design
method for anchored bulkheads without liquefaction, and validated the design method
with large shaking table tests. The method determines a factor of safety against wall
failure by dividing the driving moment (active seismic earth pressures) by the resisting
moment (passive seismic earth pressures). A balanced seismic design method was also
developed for use in anchor design. The researchers note that the balanced seismic design
method for anchored sheet pile walls enhances the seismic resistance as obtained from the
Mononobe-Okabe method, based on the results of the large shaking table tests.
Steedman and Zeng. Steedman and Zeng (1990) performed centrifuge model studies
and developed an analytical design method for anchored walls. The effects of varying the
shear modulus with depth, ground motion amplification with depth, the hydrodynamic
water pressure, and the natural frequency of the wall were analyzed. The study indicates
that their results are in good agreement with the Mononobe-Okabe design method. It was
also noted that in certain situations, the actual hydrodynamic water pressure on the wall is
larger than that predicted by Westergaard's (1931) formulation. It was also noted that the
response of a flexible wall during earthquake ground motions is highly dependent on the
natural frequency of the bulkhead.
46
Dennehy, Gazetas, and Dakoulas. In a modification of the "standard of practice"
pseudo-static design method to include deformation based analysis, Dennehy (1985) and
Gazetas et al. (1990), developed an empirical relationship between sheet pile damage and
sheet pile wall geometry for locations that were reported as not experiencing liquefaction
flow failures during the specific earthquake. Two non-dimensional factors, the effective
anchor index (EAI) and the embedment participation index (EPI) were related to the
degree of damage recorded for the bulkheads. Referring to Figure 4.5, EAI is used to
quantify the amount of available anchor capacity with;
EAI
=d
H
4.19
The EPI provides the contribution of the wall embedment as;
EPI =
P
e
Pae
4.20
f+H
For uniform soils, EPI can be approximated as;
EPI = KPe
Kae
f ) (1 f+H
f+H
4.21
The relationships between the EAI and EPI factors were plotted for 75 Japanese case
histories (Figure 4.6), with the descriptions of the degrees of damage given in Table 4.1.
Although this chart has greatly enhanced the deformation based design of sheet pile
bulkheads, several limitations are noted;
1) the design chart applies only to non-liquefiable soils,
2) the criteria for the occurrence of liquefaction is based on surface evidence
(whereas the lateral displacements of retaining structures masks the evidence of
47
liquefaction at depth), it is therefore questionable whether liquefaction did not
occur at all of the 75 case histories,
3) there is no direct contribution from the wall stiffness or earthquake characteristics
(i.e. intensity, frequency, duration) to the deformations, and
4) the EAI and EPI factors are indirectly a function of the seismic coefficients used
in design, which has been noted by Kitajima and Uwabe (1979) to be
unsatisfactory in relation to seismically induced deformations.
active failure surface
effective "point"
of rotation
Figure 4.5: Definition of Effective Anchor Index, EAI (after
Gazetas et al., 1990)
Table 4.1:
Degree of
Damage
Qualitative and Quantitative Description of the
Reported Degrees of Damage (after Kitajima and
Uwabe, 1979)
Description of Damage
2
no damage
negligible damage to the wall itself; noticeable damage
to related structures (i.e. concrete apron)
noticeable damage to the wall itself
3
general shape of anchored sheet pile preserved, but
0
1
Permanent Displacement
at Top of Sheet pile (cm)
<2
10
30
60
significantly damaged
4
complete destruction, no recognizable shape of wall
remaining
120
48
2
1.5
1
0.5
0
-0.5
0.5
0
1
Embedment participation index (EPI)
Figure 4.6: The Developed Seismic Design Chart (Gazetas et al.,
1990)
The design chart provides a useful method of estimating deformations for flexible
bulkheads embedded in non-liquefiable soils. For cases where liquefaction is expected,
this method is unsatisfactory for deformation-based design.
4.3.2 Liquefiable Soils
Towhata and Islam. Towhata and Islam (1987) developed a method of estimating the
critical seismic coefficient necessary to initiate displacements. Their approach is similar
to the method developed by Newmark (1965) to estimate lateral deformations of rigid
49
retaining structures. Towhata and Islam used the simplified expressions of the dynamic
earth pressure coefficients proposed by Seed and Whitman (1970), given as;
3
Kae = Ka + AK = Ka +-4kh
Kpe = Kp + AKpe
4.22
=K p 178 kh
4.23
and equilibrating the forces in the vertical and
Neglecting wall friction and
horizontal, Towhata and Islam developed the following equation for estimating the
critical horizontal seismic coefficient (kcr).
a tanaae
kcr =
btan(cb
aae)(1+btanaae)
1+ ctanaae
where:
m7; + P +
1
P
a=
W,
tan0+ AU sinaae
yw(hw
b=2
c=
23mn7;
1
=
tang,
W.
Wra
Wm
,(h, + D) 2 +
8(Kp
Ka)
17 Pp)/ t
7 ywhw2
8Kpr b
12
r
-i[rt(h.-FD)2
±rd(H h,)(H + 2D + hi,)1
4.24
50
where: m = 0 for no anchor capacity, 1 for full anchor capacity (dependant on the
amount of excess pore pressure generation around the anchor)
T, = the static anchor force
AU= excess pore water pressure due to cyclic shearing
n =1 when the anchor block is above the water table, and y/yb when the
anchor is completely submerged
Using Equation 4.24 it is possible to determine the limiting kh for lateral
deformations to occur. Towhata and Islam also performed numerical studies on the
effects of the anchor capacity and depth of embedment for a specific case. No generalized
guidelines are provided.
Byrne et al. Byrne et al. (1994) developed a rational method of estimating the seismic
displacements of earth dams using post-liquefaction stress-strain relations and energy
concepts. Even though this method was not developed specifically for the seismic design
of sheet pile bulkheads, it may be applied (with considerable judgement) to a
performance-based design of sheet pile bulkheads.
4.4
DISCUSSIONS OF SHEET PILE BULKHEAD DESIGN METHODS
The seismic performance of anchored sheet pile walls in non-liquefiable soils
designed with Mononobe-Okabe has been mixed. The cases of non-liquefaction failures
have been due primarily to inadequate depth of embedment, anchor design and/or the
partial generation of excess pore pressures. These deficiencies have been addressed to a
degree in the current design methods (Ebeling and Morrison, 1993) and by the work of
Dennehy (1985) and Gazetas et al. (1990). Neelakantan and his co-workers developed a
straightforward balanced seismic design method that also accounts for the past seismic
failures due to inadequate anchor design. Steedman and Zeng (1990) have highlighted
other limitations, such as the need to include the dynamic amplification of ground
motions due to the response of the backfill and phase effects on the behavior of flexible
bulkheads. These methods, though useful for non-liquefaction design (competent or
improved soils), are severely limited in that the majority of anchored bulkhead failures
51
have been caused by the liquefaction of backfill soils (which is only addressed in a
simplified manner by the Mononobe-Okabe method and its derivatives). The majority of
these methods also lack the ability to evaluate deformations (except for the method by
Dennehy and his co-workers). The only design methods that account for liquefaction are
Mononobe-Okabe (in a simplified method) and Towhata and Islam, neither method
allowing for deformation estimations. A serious limit to all of these methods is that the
transient earthquake motions are approximated by single seismic coefficients that do not
account for the duration or frequency content of actual ground motion time histories.
52
5
NUMERICAL MODELING
The database of case histories on the seismic performance of sheet pile bulkheads is
very limited. In order to supplement the case history data, a numerical modeling study
was conducted to analyze the seismic performance of sheet pile bulkheads. The numerical
model is advantageous for this study because numerous scenarios can be analyzed, and
the various design parameters can easily be adjusted to determine their influence on the
seismic performance of the bulkhead. The major concern in applying a numerical model
to soil-structure interaction problems is the numerical uncertainty, which is limited in this
study by performing a series of validation studies using the available case history data on
the seismic performance of sheet pile bulkheads.
The numerical modeling was accomplished utilizing a commercial finite difference
computer program entitled Fast Lagrangian Analysis of Continua (FLAC) version 3.30
(Itasca Consulting Group, 1995). FLAC is a non-linear, two-dimensional finite difference
program capable of modeling both static and dynamic situations. Elements or zones
represent the materials (structural and soil), with all of the elements and zones
constituting the grid (mesh). The numerical formulation of FLAC utilizes a timemarching scheme, where during each timestep the following procedures take place within
FLAC (Figure 5.1); 1) nodal velocities and displacements are calculated from stresses and
forces using the equations of motion, 2) a constitutive model is then used to calculate
strain rates from the velocities and stresses from the strain rates. These two procedures
are then repeated until the computed unbalanced forces within the mesh are within a user
specified limit. FLAC uses an explicit method, where the calculation timestep is very
short compared with the time necessary for information (acceleration, velocity,
displacement) to physically pass from one element to another. FLAC also utilizes a
Lagrangian formulation, in which the incremental displacements are added to the
coordinates at each timestep so that the grid moves and deforms with the material that it
represents.
53
There are several advantages and disadvantages in using FLAC, compared to implicit
finite element programs as outlined in Table 5.1. Because of the explicit Lagrangian
methodology and the explicit use of the equations of motion, FLAC is advantageous in
modeling nonlinear, large strain, physically unstable situations (Itasca, 1995).
Equilibrium Equation
(Equation of Motion)
new
stresses or
forces
new velocities
and
displacements
Stress/Strain Relationship
(Constitutive Model)
Figure 5.1: Basic Explicit Calculation Cycle
This chapter presents the aspects of the numerical modeling program used in this
study. Section 5.1 describes the constitutive soil model, Section 5.2 describes the pore
pressure generation scheme, Section 5.3 describes the general modeling parameters (for
soils, structures, water, earthquake, and the boundary conditions), and Section 5.4
describes the extensive program used to validate FLAC for use in modeling seismicallyinduced liquefaction and the performance sheet pile bulkheads.
5.1
CONSTITUTIVE SOIL MODEL
An effective stress Mohr-Coulomb constitutive model was used for this study. The
constitutive model is able to model plastic deformations utilizing a plastic flow rule. The
54
Table 5.1:
Comparison of FLAC and Finite Element Numerical
Programs
FLAC (explicit)
Timestep must be smaller than a critical value for
stability.
Finite Element (implicit)
Timestep can be arbitrarily
unconditionally stable schemes.
large,
with
Small amount of computation effort per timestep.
Large amount of computational
timestep.
No significant numerical damping introduced for
dynamic solution.
Numerical damping dependent on timestep
present with unconditionally stable schemes.
No iterations necessary to follow nonlinear
constitutive law.
Iterative procedure necessary to follow nonlinear
constitutive law.
effort per
Provided that the timestep criterion
is always Always necessary to demonstrate that the above
satisfied, nonlinear laws are always followed in a mentioned procedure is a) stable, and b) follows
the physically correct path (for path-sensitive
valid physical way.
problems).
Matrices are never formed. Memory requirements
are always
limitations.
at a
minimum. No band-width
Stiffness matrices must be stored. Ways must be
found to overcome associated problems such as
band-width. Memory requirements tend to be
large.
Since matrices are never formed, large
displacements and strains are accommodated
without additional computing effort.
Additional computing effort needed to follow
large displacements and strains.
elastic behavior of the soil is defined by the bulk and shear modulus, and the strength is
defined by the angle of friction and cohesion. This significantly simplifies the dynamic
soil behavior and is not capable of accounting for the strain dependent dynamic properties
such as damping and shear modulus. Despite the use of this simplified constitutive soil
model, it has been demonstrated to yield satisfactory displacement results for a variety of
applications involving seismically-induced deformations of earth structures and retaining
walls (e.g. Roth et al., 1986; Roth & Inel, 1993; Dickenson & Yang, 1998). The elastic
and strength properties of the soil were estimated from established correlations with
normalized SPT values ((Nd60), unless otherwise noted.
55
5.2
PORE PRESSURE GENERATION
The pore pressure generation scheme utilized in these analyses uses several empirical
procedures developed by Seed and others over the last 25 years (Martin et al., 1975, Seed
et al., 1976, Seed 1979). During each timestep of the dynamic analysis, the effective
stresses decrease and pore pressures gradually increase in liquefiable soils, until a state of
full liquefaction is reached. Liquefaction resistance curves are developed using the cyclic
stress ratio (CSRfi ) and number of cycles to liquefaction (Nijq) at 3 and 30 cycles (Figure
5.2). The values of 3 and 30 are used because they cover the range of number of shear
cycles for earthquake magnitudes in the range of engineering interest (M = 5 to 8). The
linear relationship provides a fairly good representation of the majority of CSRfied curves
(Figure 2.3).
CSR @
3 cycles
CSR
CSR @
30 cycles
3
30
Number of Cycles Required for Liquefaction (Nhz)
Figure 5.2: Modeled Liquefaction Resistance Curve
There are two methods of inputting the CSRfidd data; 1) directly from the results of
cyclic testing (e.g. simple shear, triaxial), or 2) utilize in-situ penetration resistances (e.g.
SPT, CPT) and empirical relationships. Method 1 is straightforward, as long as the
CSRfieid is corrected for the testing method, in-situ stresses and various earthquake
magnitudes, as described in Section 2.2.2. For method 2, the CSRfidd is corrected for
initial static shear stresses and initial overburden stresses (Section 2.2.2), and the values
56
of CSR at 3 and 30 cycles are determined by multiplying CSRfidd value by 1.5 and 0.89,
respectively. The multiplication factors were determined from Arango (1996) for various
numbers of shear cycles.
Cyclic stress ratios caused by the earthquake motion (CSReq) are monitored during
each timestep, and if CSReq exceeds CSRfield, a numerical curve fitting is conducted to
determine Nhq at CSRfieid. The earthquake-induced damage (a function of the excess pore
pressure generation) is monitored in the FLAC model using a damage parameter (D). The
damage parameter is updated at each shear cycle, and is defined as the summation of the
inverse number of cycles to liquefaction (Equation 5.1). After each cycle, D is related to
the pore pressure by an empirical function. The simplest function
was used in the
current study, based on the satisfactory results of Roth and Inel (1993) for liquefaction
studies.
D =E
1
5.1
Niiq
A simplification was made in the models in which pore pressures were allowed to
generate using the above formulation, but were not allowed to dissipate. This
simplification was used because the results from validation case studies in which the
dynamic motion and groundwater dissipation where coupled proved to be unsatisfactory.
The reasons for the unsatisfactory modeling of dynamic motion and groundwater
dissipation are unclear. The simplification that was made can be somewhat justified by
noting that expected flow in sand during the 10 to 40 seconds of earthquake motion
would be very small (-0.1 to 0.4 cm3/cm2). It should also be noted that the largest effect
of pore pressure dissipation is ground surface settlement, which was calculated following
the analyses using the proposed post-liquefaction volumetric strain relationship by
Ishihara and Yoshimine (1992). Therefore, for this study, the neglect of pore pressure
dissipation is an acceptable simplification.
57
5.3
GENERAL MODELING PARAMETERS
The numerical modeling involved two solutions; static and dynamic. After
generating the model geometry, a static solution was sought to equilibrate the initial
stresses within the soil. After the static solution, the dynamic modeling was performed.
The modeling for the validation case histories and the parametric studies (Chapter 6)
followed the same procedures, as outlined in the following sections.
5.3.1
Modeling of Soil Elements
The known site properties for most of the analysis included the uncorrected standard
penetration test (SPT) results, and the elevation of the soil layers. In order to conduct a
FLAC analysis, the soil density (p), angle of internal friction (0), cohesion (c), bulk
modulus (K), shear modulus (G), Poisson's ratio ( v), corrected penetration resistance
(Wd60), relative density (DR) and undrained residual shear strength of the liquefied soil
(Sr) are also needed. The soil density and angle of internal friction were estimated from
the SPT results using well-established correlations, when the values were not provided in
the references. The low-strain shear modulus (Gmar) was calculated from the shear wave
velocity (Vs) and total density of the soil (Equation 5.2). The shear wave velocity was
estimated from the SPT, when not provided in the references. Several different
established methods were used to estimate the shear wave velocity of the soils, depending
on the specific site profile and soil conditions.
Gmax = Vs2 Ptotal
5.2
Once the shear modulus was determined, Equation 5.3 was used to estimate the bulk
modulus of the soil, using Poisson's ratio.
K=
2G(1+ 0)
3(1 2v)
5.3
58
The corrected penetration resistance was calculated using Equation 5.4. The initial
static analysis provided the effective overburden pressures (o-') in the model, which was
then used to determine CN using Equation 5.5 (Liao & Whitman, 1986). Because all of
the case histories were from Japanese cases, the efficiency (Er) of the SPT method was
assumed to be 72% of the maximum, which is the average of the free-fall and throw
release type hammers, as given by Seed et al. (1985).
(N1)60
CN =
Er
N 'LN 60
95760 Pa
5.4
5.5
avo
If the shear wave velocity was unknown, it was estimated from SPT data using the
relationship given in Equation 5.6, formulated by Imai et al. (1975) from Japanese case
history data.
Vs = 89.8N0341
5.6
The relative density of the soil was estimated using Equation 5.7 (Yoshida et al.,
1988), with the following; SPT values uncorrected for overburden pressure (N60), the
effective overburden pressure (cr), and the best fit values of Co = 25, Cl = 0.12 and
C2
= 0.46. Equation 5.7 was used to estimate relative density because it was formulated
from Japanese case history data, and all of the validation case studies were from Japanese
locations.
5.7
59
The undrained residual shear strength was estimated using Equation 5.8 (from Figure
2.11), developed by Stark and Mesri (1992) for SPT results corrected for fines content.
S,. = 0.0055 avo '*(N1)60-cs
5.8
The constitutive model used during the numerical analysis followed the MohrCoulomb failure criteria (r=cr'tan0), and as the pore pressures increase during cyclic
shearing, the effective stresses decrease, until the limiting shear stress value (undrained
residual shear strength) is reached. The strength of the soil at full liquefaction was then as
the undrained residual shear strength.
5.3.2
Modeling of Structural Elements
The sheet pile walls and anchors were modeled using FLAC pile elements. These
elements allow relative movement between the soil and pile elements. The relationship
between the pile and soil elements is simulated with springs that have a stiffness and
yield value for both the normal and shear directions. The stiffness was calculated as the
minimum value of Equation 5.9 for all soil elements adjacent to the pile elements. The
same calculated stiffness value was used for both the shear and normal directions. The
yield value for the normal direction was set very large to prevent the soil moving through
the wall, whereas the yield value for the shear direction was set equal to zero, assuming
that there was no static friction between the soil and pile. A frictional shear resistance
angle (b) was input for the dynamic (not earthquake) friction between the wall and soil. If
the wall was continuous (sheet pile wall, or a continuous anchor wall), the modulus of
elasticity was increased to account for the effect of plane strain using Equation 5.10
(Itasca, 1995) to determine the new modulus. If the tie rods or anchor were not
continuous, the modulus of the tie rod and anchor were divided by the spacing of the tie
rod or anchor to provide a modulus value per unit depth. The tie rods were modeled as
cable elements, which are able to transfer tension loads, but unable to transmit moments.
60
Stiffness =
+yG
K
3
62min
5.9
E
E plane strain
1
5.3.3
5.10
02
Modeling of the Earthquake Motion
The input earthquake motions were recorded accelerograms that had been corrected
for the recording instrument. For simplification, the vertical motion was not used, and the
horizontal motions were vectorally summed to the motion perpendicular to the face of the
wall. The motions were input at the base of each model as acceleration time histories. The
damping of the earthquake motions for numerical stability utilized the Rayleigh damping
method within FLAC. The acceptable Rayleigh damping used in the models was
determined from the validation studies to be 5% at 5 Hz.
There is an inherent displacement in recorded earthquake motions due to various
factors
(e.g.
movement
of the
recording
instrument,
permanent
earthquake
displacements), and is termed baseline shift. The baseline shift has been removed from all
of the presented FLAC displacements.
It should also be noted that the applied acceleration time histories do not always
correspond to the same time scale as the recorded motions. This is because some of the
small amplitude portions of the recorded motions were removed to decrease the solution
time for the numerical runs.
5.3.4 Modeling of the Water
The water in front of the wall was modeled indirectly by including the resulting water
pressures along the boundaries. This allows a simple modeling of the water, but does not
account for the dynamic interaction of the wall and water. The water within the soil was
modeled directly, and was allowed to flow during the static solutions. During the dynamic
61
solutions excess pore pressures were allowed to generate, but the dissipation of these pore
pressures was not modeled (as discussed in Section 5.2).
5.3.5
Boundary Conditions
The boundary conditions for the static solutions consisted of the bottom boundary
being fixed in both the horizontal and vertical directions, while the sides of the model
were treated as rollers (by fixing only the horizontal direction). During the dynamic
analysis the bottom boundary was freed in the horizontal direction to allow application of
the horizontal acceleration, and the sidewalls were treated as an infinite medium (free
field), having the same properties as the adjacent model perimeter zones.
The lateral dimensions of the model were determined as optimum from the validation
case studies as seven times the total wall height (H+D) in the backland and three times
the total wall height in front of the wall.
5.4
VALIDATION OF NUMERICAL MODEL
The use of FLAC for the seismic modeling of generic sheet pile walls in ports
required that the numerical model be validated and calibrated. The validation of FLAC
for this study, involved modeling well-documented case histories and comparing the
calculated and measured displacements. Five case histories were chosen, including: two
wharves at Akita Port during the MJMA 7.7 1983 Nihonkai Chubu Earthquake, a wharf at
Ishinomaki Port during the MJMA 7.4 1973 Miyagi Ken Oki Earthquake, and two wharves
at Kushiro Port during the MJMA 7.8 1993 Kushiro Oki Earthquake. These were the most
detailed case histories available for typical sheet pile bulkheads. The modeled geometries
were the same as given in the references, except for the noted differences and
simplifications.
Several
notes on the calculated displacements include; a)
the horizontal
displacements were measured at the top of the sheet pile wall. The vertical displacements
were measured at the top of the soil column, adjacent to the wall, therefore the reported
62
vertical displacements take into account the relative vertical movement between the wall
and soil. b) the vertical time history displacement plots are a result of the volume change
due to horizontal movements only, and do not account for the volumetric strain due to
pore pressure dissipation. The volumetric strain due to pore pressure dissipation was
calculated following the analysis using the proposed post-liquefaction volumetric strain
relationship by Ishihara and Yoshimine (1992). The post-liquefaction vertical
displacements are noted in the text.
Akita Port
5.4.1
The 1983 Mjmik 7.7 Nihonkai Chubu Earthquake struck Akita Port, Japan, causing
severe liquefaction damage in some locations, while other locations were left undamaged.
Two case histories were chosen from this earthquake. One site suffered severe damage,
while the other suffered negligible damage. There is considerable data available on the
earthquake, site conditions, and structure properties for both case histories (Iai &
Kameoka, 1993). The layout of the port (Figure 5.3) includes the location of the strong
motion observation station where the modeled earthquake acceleration time histories
were recorded. Ohama Wharf 1 and 2 bulkheads are almost identical. The differences
include the soil properties of the backfill, depth of wall penetration, and structural details
of the anchors. Ohama Wharf 1 and 2 bulkhead geometries are show in Figure 5.4 and
Figure 5.5, respectively.
5.4.1.1
Soil Conditions
The soil conditions, including standard penetration and sieve analysis results, for
Wharf 1 and 2 are shown in Figure 5.7 and Figure 5.8, respectively. The numerical
modeling utilized the following soil layers and conditions. For Wharf 1, the soil profile
was simplified into two layers for the numerical analysis, the top layer, a hydraulically
backfilled sand, extends to a depth of approximately 22.5 m below the ground surface.
The second layer is also a sand layer, and extends to the base of the model (24 m). For
Wharf 2, four soil layers were modeled. The hydraulically backfilled sand layer, which
63
has a sloping base (Figure 5.5), is 12 m thick adjacent to the sheet pile wall. The backfill
is underlain by 6 m of naturally deposited sandy soil. The third layer, a medium stiff clay,
is 4.5 m thick, and underlain by another sand layer to the base of the model (24 m). The
soil properties used in the numerical analysis are presented in Table 5.2. The strength
properties were based on the laboratory test results by Iai and Kameoka (1993).
I30°E
135°
14.°
awn,
Akita
0
Japan Sea
okyo
(a) Location of Akita
Closer View
(b)
Ohoma No.l Wharf
Shinio River
Ohamo No. 2 Wharf
Akita part51-
/
,
'Japan Sea
,
o
.14
o'
Observation
Sto ion
ca
4
,
Contours of Base
i.,k'
;--o
Liquefied Area
;
/ S,trong,-Motion
,
,
p 7 .,,e
0
.
...yo
,.-Jorn
(c) Plan of Akita Port
Figure 5.3: Plan of Akita Port, Japan after the 1983 M 7.7
Earthquake (Iai & Kameoka, 1993)
64
20.00
12.00
0.00_
" SEMI HIGH TENSILE
STRENGTH STEEL
TIE ROD
0
0:
-10.00
-10.10
Layer 2 ( Sand)
-15.50
Unit in Meters
20.50
Loyer 4 ( Sand)
Figure 5.4: Akita Ohama Wharf 1 Wall Profile (Iai & Kameoka,
1993)
20.00
t 2.00
_zr I 0 00
10
HIGH STRENGTH STEEL TIE ROD
11
_11
-11
Layer
CRA C
I
Bockfill Sand)
O
W
-10.00
I
Layer 2 ( Sand)
a.
-14.50
-20.50
Layer 3 ( Cloy)
Layer 4 (Sand)
- BEFORE THE EARTHQUAKE
AFTER THE EARTHQUAKE
UNIT IN METERS
Figure 5.5: Akita Ohama Wharf 2 Wall Profile (Iai & Kameoka,
1993)
65
3
2 r-1
2
(m)
(m)
(m)
(m)
0
0
2 EI
1
IL
0
I
I
r
I
0
1
0
5
\-110
10
m)
(m)
1-110
(m)
10
(m)
(m)
Figure 5.6: Deformed Cross-Sections for Akita Ohama Wharf 2
Cr
100
90
60 -(No.58-22)
U" X 70 _
(9 60
Z,W 50
030 -
W
0-
.----
Iiv
--- -7'DEPTH
/
,
...,,/
10 .-_
001
01
I
CC
100
L
70
7
1
03 20
°0001
I, 3m
---6
mm)
I
10.0
10
i
500
DEPTH
7 9m-1
---- 10-16
5.g
y a0
20
10
0
0001
0.01
01
1.0
10.0
50.0
InIrn1
Figure 5.7: Soil Profile and Grain Size for Ohama Wharf 1 (Iai &
Kameoka, 1993)
The SPT values were input exactly as shown in the figures. Shear wave velocity (V3)
measurements were obtained from an exploratory borehole at a neighboring wharf,
Ohama Wharf 3. The measured shear wave velocity and soil profiles at Ohama Wharf 3
are given in Figure 5.9. Because the SPT data for the soil layers at Ohama Wharves 1, 2
and 3 are similar (except for the backfill), the shear wave velocities were assumed the
same between the three sites.
66
100
LC5
Elev- Soil
otion Type
S PT N -VALUE
10
(m)
20 30 40
500.01
Z
( mm )
D50
0.1
LT:
1.0
x
w
0
+0.33m
''''' i
Wf`.
Wy
7/'
90 -(No 58-2 )
80
70
60 ORIGINAL GROUND
-
LI
BACKFILL SAND
1/
10
0 11mm)
0.01
100
/
7 _
8:
,71 60
50
10
<,
IO
10.0
f----.---
90 -(No.48-10)
..
12
40
30 o
0.001
ix
2m
4-10
V
SOIL
W 20
,
DEPTH
500
I DEPTH
16m-
...--
18
,
20
%
40
to 3
-a
2
10
-15
o
---------
0
v
0.001
__- ..---
0.01
01
1.0
10.0
50.0
tk
-25
-30
o-o 48 5
58 - 2
Figure 5.8: Soil Profile and Grain Size for Ohama Wharf 2 (Iai &
Kameoka, 1993)
Table 5.2:
Model Soil Properties for Ohama Wharf 1 and 2
Wharf 1
Layer 1
Layer 2
Density (kg/m3)
Friction Angle (0)
Cohesion (kPa)
Poisson's Ratio
Porosity
Layer 1
Wharf 2
Layer 2
Layer 3
1500
43
1500
44
1500
37
1500
0
0.3
0.4
0
0.3
0.4
0
0
0.3
0.4
0.3
0.4
41
1500
39
48
0.3
0.4
Layer 4
1600
44
0
0.3
0.4
5.4.1.2 Structural Properties
Ohama Wharf 1 and Wharf 2 were constructed with the same wall section, FSP 6L,
manufactured by Nippon Steel. The properties of the modeled structural elements are
presented in Figure 5.9. The anchor and tie rod spacing was 2 m for both Wharf 1 and
Wharf 2 (Iai & Ichii, 1997). The friction between the sheet pile wall and the soil was
modeled as 14 degrees for Wharf 1, where limited pore pressure generation was expected,
and 0 degrees for Wharf 2, where significant pore pressure generation was expected.
67
5.4.1.3
Earthquake Motion
The acceleration time histories recorded at the strong motion site (Figure 5.3) were
vectorally summed to yield the acceleration time history normal to the wall. The motion
normal to the wall was then deconvolved to bedrock using the equivalent linear dynamic
soil response model, SHAKE91 (Idriss and Sun, 1992). The soil profile at the strong
motion site is shown in Figure 5.10. The computed bedrock motion was then propagated
to the base of the models (24 m below the ground surface) for Ohama Wharves 1 and 2
using SHAKE91. The earthquake time histories recorded at the strong motion site are
shown in Figure 5.11. The transformed time histories for Wharf 1 and 2 are shown in
Figure 5.12 and Figure 5.13, respectively.
x
1w
0
tai
0.
,I-
SPT
N-VALUE
(1.3 ---- Vp (m/sec)
1090
2000
Vs
.7,1,
(m) 0
cr)
lo
j
,
( m / sec I
10203040500
400
13
4
_
1200
L
-I
11,
0
(90
1900
15
290
20
lI
4.
11
25
vt
li
1:
1;..
I
210
1760
30
Figure 5.9: Shear Wave Velocity Profile for Ohama Wharf 3 (Iai
& Kameoka, 1993)
68
Table 5.3:
Model Structural Properties for Ohama Wharf 1
and 2
Property
Sheet Pile Wall:
Type
Moment of Inertia, I (m4)
Modulus of Elasticity, E (kPa)
Cross-Sectional Area (m2)
Wharf 1
Wharf 2
FSP6L
8.60e-4
2.20e11
0.0306
FSP6L
8.60e-4
2.20e 1 1
0.0306
Anchor:
Type
Diameter (mm)
Thickness (mm)
Moment of Inertia, I (m4)
Modulus of Elasticity, E (kPa)
Cross-Sectional Area (m2)
Cable:
Modulus of Elasticity, E (kPa)
Radius (m)
SOIL TYPE
DE
(m1
12
1.66e-3
1.00el1
0.0236
7.84e-4
1.00el1
0.0207
1.00el1
0.10
1.00el1
0.10
SPY
N -VALUE
S WAVE
P WAVE
m/s ) 0 (m/s )
CCIARSE SAND
900
'
5
550
10
I 203040500 200 400 ECO I
0 ec.
'
2 Identical Pipe Piles
Pipe Pile
750
E=TIE
1.132ii
Si
GRAVEL MIXED
WITH SAND
I0
SANDY GRAVEL
.
SAND MIXED
SANDY GRAVEL
15
I
.5/
II
GRAYEL MIX 0
WITH SAND
0
20
SAND STONE
570
25
Figure 5.10: Soil and Velocity Profile at the Akita Strong Motion
Site (Iai & Kameoka, 1993)
69
North
3.0
M
2.0
1.0
2
0.0
e
-1.0
IMIRIPPIIIMPIPPIMITSITANIIIITINATIMIF
7/
§ -2.0'
3.0
East
3.0
'a
C
2.0
s
IMIIMPIPIIMIT111113WIT1111.1111INFRNIfillTrir
0.0
ti
11141111,1,1111MIMEALMia
-1.0
-2.0
-3.0
Vertical
3.0
2.0
1.0
0.0
-1.0
-2.0
-3.0
0
10
5
20
15
25
30
35
40
Time (seconds)
Figure 5.11: Recorded Time Histories for the 1983 M 7.7 Nihonkai
Chubu Earthquake at Akita Port
1.0
0.5
0.0
-0.5
I MINI& 111111111% MAI
A
"w
0111111111111111Wall
Al
r
A
A
11"
-1.0
5
10
15
20
25
Time (seconds)
Figure 5.12: Input Time History for Akita Ohama Wharf 1
30
35
70
1.0
0.5
C
0.0
WilliglillEMIEMIFIVI AI
A
-0.5
-1.0
5
10
15
20
25
30
Time (seconds)
Figure 5.13: Input Time History for Akita Ohama Wharf 2
5.4.1.4 Results of Ohama Wharf 1
The deformed grid for Wharf 1 is shown in Figure 5.14, with calculated horizontal
and vertical displacements of 0.03 m 0.055 m, respectively. The calculated vertical
displacement, including the post-liquefaction volumetric strain, was 0.055 m. The time
histories of the displacements (not including the post-volumetric strain) are shown in
Figure 5.15. The maximum measured horizontal displacement was 0.05 m. The measured
vertical displacements were not available. The results for Ohama Wharf 1 are consistent
with the measured displacements.
Figure 5.14: Deformed Grid for Akita Ohama Wharf 1
(deformations to scale)
71
0.01
- Horizontal Displacement
0.00
Vertical Displacement
-0.01
E
t -0.02
E
es
a. -0.03
1\\*NOVO
-0.04
-0.05
-0.06
0
5
10
15
20
25
30
Time (seconds)
Figure 5.15: Time Displacement Plot for Akita Ohama Wharf 1
5.4.1.5
Results of Ohama Wharf 2
The deformed grid for Wharf 2 is shown in Figure 5.16, with the horizontal and
vertical time displacements are given in Figure 5.17. The measured displacements are
presented in Figure 5.6 and Figure 5.18. The maximum horizontal displacement
calculated in FLAC was 1.30 m at the top of the wall, where the measured displacements
had a maximum of about 1.6 m and an average of about 1.1 m. The calculated vertical
displacements in the backfill adjacent to the wall, including and not including postliquefaction volumetric strain, were 0.78 and 0.60 m, respectively. The measured vertical
displacements ranged between an average of 0.5 and a maximum of 1.4 m. It should be
noted that the formation of a plastic hinge was evident during post-earthquake inspections
of the bulkhead (Figure 5.6). A plastic hinge developed only to a small degree in the
numerical model. The difference between the measured and calculated profiles could be
contributed to several reasons; 1) inadequacies in the numerical model, 2) lack of
structural integrity in the field due to corrosion, physical imperfections, etc.
72
10 meters
Figure 5.16: Deformed Grid for Akita Ohama Wharf 2
(deformations to scale)
0.20
-Horizontal Displacement
0.00
Vertical Displacement
-0.20
-0.40
0
-0.60
0.
-0.80
5
10
15
20
25
30
35
Time (seconds)
Figure 5.17: Time Displacement Plot for Akita Ohama Wharf 2
5.4.2
Ishinomaki Port
The 1978 MimA 7.4 Miyagi-Ken-Oki Earthquake caused severe liquefaction damage
in the Ishinomaki Port. The case study utilizes the data from Shiomi Wharf (-4.5 m) that
was presented by Iai, Tsuchida and Finn (1985). The general location of the city, strong
73
APPROACH
SECTION A
SECTION B
105m
80m
Horizontal Displacement at the Top of Sheet Pile
APPROACH
20m
30m
2
I
Cm>
a)
(m)
I
1
1
Elevation of Sheet Pile
Design Elevation
+1
b)
(m)
(m)
+I
+0.5
+0.5
-0.5
-0.5
Figure 5.18: Measured a) Horizontal and b) Vertical Displacements
along the Face of the Bulkhead (Iai & Kameoka, 1993)
motion site and Shiomi Wharf (-4.5 m, Site D), are shown in Figure 5.19. The bulkhead
profile is presented in Figure 5.20.
5.4.2.1
Soil Conditions
Four borings were conducted along the wharf, one previous to (CB-10 dashed), and
three after (CB-2, CB-6, CB-10 solid) the earthquake (Figure 5.21). The boring before the
earthquake was used to determine the soil properties to approximately 14 m below the
ground surface, where the combination of the other three borings were used for the deeper
soil profile. The generalized profile used in the FLAC model consisted of two soil layers,
a medium to fine sand backfill extending from the ground surface to a depth of 15 m, and
underlain by a sandy silt that extends to the base of the model (23 m). A rubble mound is
located directly behind the wall and in front of the anchor. The mound behind the wall is
small and was not modeled, but the mound in front of the anchor was incorporated into
the FLAC model. The uncorrected SPT values were input directly from the boring logs.
The modeled soil properties are given in Table 5.4. The shear wave velocities were
estimated from the SPT values.
74
141°E Ofunato 142°
39° N
Strong-motion
I
Fault plane
Ishinomaki
Accelerogroph Site
Send 2i
Xj
hiogama
38°
Kaihoku Bridge
Epicenter
Pacific Ocean
mi
ISHINOMAKI CITY
ISHINOMAKI PORT.
Site-C(Non-Liquefaction)
Site -D
(Liquefaction
Ishinomaki Fisting Port
000ctoo
orn000cloocIT
Site- B(Liquefaction )
Site-F(Liquefaction
Site -A(Liquefaction) Site -E(Non Liquefaction
ISHINOMAKI BAY 9
icrr/
Figure 5.19: Layout of Ishinomaki Port during the M 7.4 MiyagiKen-Oki Earthquake (Iai et al, 1985)
15.00
14.20
0
K.P+3.00
K.P +1.50
020.0(TP-0.8745)
01.0.0(KP-0.025)
li
emu Hugh Tensile Stren th Steel
t :20.80m
Ao:.
400
TO
Steel sheet pile anchorage
1) type 400x 10x 10.5
K.P :4.50
+3.50
+250
+328
325
=4.0m
-1.50
Steel sheet pile
IIA type 400x 150x 13.1 C: 11.5m
-10.00
Figure 5.20: Shiomi (-4.5m, Site D) Bulkhead Profile (Iai et al.,
1985)
75
'
ii
Soil Tyne
'N -Value
0
ra
N - value
Soil Type
1 0 203040 50
20 3040 50
Asphalt
Block
Gravel
Gravel Stone
___w___'
rk Groy
Medium
10 20 30 40 50
93
Asphalt
O'ir
CB-I0
N - volu4
au-value(kgt/cmt)-00.6 09 1.0 1.2 14 1.6
CB-2
2( 03m4,N4;
Brown
0.94
I
20m N.4
N.2
Stone
San;ii
111111111111
11/111111111
''
11M11111111
ffIr41333:22
11E111111111
111111111111
Sandy
111111111111
'''''''''d1 111111111111
11 111
'')
(Medium
c'.
1
I
8I0
N.13 I
)1(2,4,2)
/
'1111
-506
m IN = 16
1110m N.)
(3,4,4)
IIIII
N.I6 10.0m N.15
(4,5,6)11
(4,4,5)
11111
(5,7,4)
12.0m N.18
N
1
I
4.5m N.I5
(5,4,6)1
;''' (F'n.sad) 111111111111
Dark grey
(
N
18
5.
1
Fine sand
-1506
lue gray
Measured in 97
18.0m N =S
11
1
1
1
1
-14
IIIII1
11
2t/Om N.6
20.3m N 6
Silt mixed
with sells
1
Measured112967.
181m N.10
( 2,2,6)
v
1
140m N.16
(4,6,6
III I
8m WI
6.2m N.8
1111111111
(a ) CB -2
1111111
o Very fine xnd
(20w N =16
,,;:..0::,.,
I
%,6.0m NOB
100m N =13
1.:
1mo
''.1 Scm soli 111111111111111111111111
(Script
"
1
(5,5,6
Fine to
Medium Sand
11M 11111111
111111111111
111111111111
1aiCloy
1
(3,5,5)
111111111111
:,
1
I
(2,3,2)
- Fine sand
60m N =13
11111111E11
--4
II
4.0rn N.7
2., Medium send
40m N.6
:..4.-,
1
Sandy Silt
2.3m N.5
220m
III III
24.0m N.5
24.0m
Medium sand
mixed with
shells
iBlue gr ay
2825m N.7
Tirr
with shells
30.0m N 8
--e-Dark gray
(1,1 1.11
(b) CB-6
Cc) CB -10
Figure 5.21: Boring logs at Shiomi Wharf (Iai et al., 1985)
Model Soil Properties for Ishinomaki Port
Shionzi Wharf (-4.5m, Site D)
Layer 1
Layer 2
Rubble
Density (kg/m3)
Friction Angle (0)
1500
36
Poisson's Ratio
Porosity
0.30
0.40
1500
36
0.30
0.40
4
Dark gray
-12
Table 5.4:
/4
1800
42
0.20
0.40
N.8
27.5m N-9
(4,23)
11111
29.0 m N.4
76
5.4.2.2 Structural Properties
The wall section was constructed from Nippon FSP3A sheet pile sections, and the
anchor was constructed from Nippon FSP2 sheet piles sections. The total height of the
sheet pile wall is 13 m, with 5.5 m embedment, the anchor embedment is 4 m, with a tie
rod length of 20 m. The spacing of the tie rod was unknown, and assumed to be one
meter. This introduces an error into the calculated displacements because of the
uncertainty in the elastic behavior of the tie rod system. Because the backfill within the
model was expected to liquefy, the friction between the sheet pile sections and soil was
assumed to be zero. The modeled structural properties are presented in Table 5.5.
Table 5.5:
Model Structural Properties for Ishinomaki Port
Property
Sheet Pile Wall:
Type
Moment of Inertia, I (m4)
Modulus of Elasticity, E (kPa)
Cross-Sectional Area (m2)
Shiomi Wharf (-4.5m, Site D)
FSP3A
2.281e-4
2.20e1 l
0.0186
Anchor:
Type
Moment of Inertia, I (m4)
Modulus of Elasticity, E (kPa)
Cross-Sectional Area (m2)
FSP2
8.740e-5
2.20e1 l
0.0153
Cable:
Modulus of Elasticity, E (kPa)
Radius (m)
5.4.2.3
2.00e11
0.10
Earthquake Motion
The strong motion time histories were recorded on bedrock, and were assumed to be
the same as the bedrock motions beneath the site. The transformed time history was
propagated from bedrock to the base of the model (23 m below the ground surface) using
SHAKE91. The earthquake time histories recorded at the strong motion site (Kaihoku
77
Bridge) are shown in Figure 5.22. The transformed time history for Shiomi wharf is given
in Figure 5.23.
S42W
3.0
a
2.0
3 -2.0
-3.0
W42N
3.0
2.0
1.0
1111111111,1111111111MITIFI1111,
0.0
IN 111
-1.0
WAIN "Nil 'UM
-2.0
3.0
0
10
14
12
16
20
18
Time (seconds)
Figure 5.22: Recorded Time Histories at Ishinomaki Port
1.5
1.0
0.5
0.0
-0.5
111111 7
.1 Ail
Ai.
11
r.
V
-1.0
-1.5
8
2
Time (seconds)
Figure 5.23: Input Time History for Ishinomaki Wharf
10
12
78
5.4.2.4 Results of Ishinomaki Shiomi (-4.5 m) Wharf
The deformed grid at 12 seconds is given in Figure 5.24. The calculated horizontal
and vertical time displacements are shown in Figure 5.25. The time displacement plot
extends for 50 seconds, where the earthquake motion was only 12 seconds long, and it
should be noted that the displacements continued to increase after 12 seconds at a steady
rate. This is theorized to be due to the modeling limitation of not allowing the excess pore
pressures to dissipate. If excess pore pressures were allowed to dissipate, the
displacements would likely taper off after 12 seconds. Therefore, the calculated
displacements are from 12 seconds. The calculated horizontal displacement is 0.60 m and
the vertical displacements, including and not including post-liquefaction volumetric strain
are 0.30 and 0.25 m, respectively.
The measured results are presented in Figure 5.26 and Figure 5.27, and indicate a
large variation in the horizontal deflections along the top of the bulkhead. This is
theorized to be partly due to the end effects of the bulkhead. It was also noted by Iai et al.
(1985) that a landslide of unknown extent was noted at the site, and may have also caused
the large variations in the displacements. The maximum measured displacement of the
wall was 1.16 m, with an average horizontal displacement of approximately 0.60 m. The
measured vertical displacements ranged between an average of 0.05 and a maximum of
0.10 m.
Figure 5.24: Deformed Grid for Ishinomaki, Shiomi Wharf
(deformations to scale)
79
0.20
Horizontal Displacement
0.00
Vertical Displacement
-e- -0.20
-0.40
E
tv
ct.) o -0.60
a.
co
6 -0.80
-1.00
-1.20
0
10
20
30
40
50
Time (seconds)
Figure 5.25: Time Displacement Plot of Ishinomaki, Shiomi Wharf
5.4.3
Kushiro Port
The MRAA 7.8 Kushiro Earthquake occurred on January 15, 1993, and caused severe
liquefaction damage to the Port of Kushiro at Kushiro City. The majority of the data for
this case study was taken from Iai et al. (1994) and Ueda et al. (1993). The location of
Kushiro City is shown in Figure 5.28. The case study used two wharves from Kushiro
Port, indicated as Site C and Site D in Figure 5.29. The location of the strong motion site
is also indicated in Figure 5.29. The geometry of Site C is shown in Figure 5.30, and
consists of a sheet pile wall, backed by a rubble mound, and supported by a battered pile
anchor, located 20 m into the backland. The geometry of Site D is shown in Figure 5.31,
and consists of a steel pipe wall, backed by a rubble mound, and supported by a
continuous sheet pile wall, 30 m into the backland. Site D was also an area of soil
improvement prior to the earthquake, including both sand compaction piles and gravel
drains.
80
N
No.16
6 No.15
23
No.1 4
40
No.l 3
37
No.12
Boring CB-10
E
O
o
co
to
O
O
N
o
Figure 5.26: Measured Horizontal Deformations at Shiomi Wharf
(-4.5 m), displacements in mm (Iai et al., 1985)
(2'
za
z6
za
za
zd
za
z6
6
z
za
z6
II3
z
6
z
z6
o
z
z6
ID
6
ID
zei
Design I levation
M
+3.000
4.2.900
+2.800
__
4-2.700
+2_600
Figure 5.27: Measured Vertical Displacements at Shiomi Wharf (4.5 m) (Iai et al., 1985)
81
x Epicenter
1993. 1.15
(M = 7.8)
Pacific Ocean
(a) Location of Kushiro
Closer View
Figure 5.28: Location of Kushiro Port (Iai et al., 1994)
KUSHIRO PORT
f
OrI
/EOSI Port
0
500
I000m
Site A
Site C
Site D
West Port
Site S
Pt,
ormer Beach Line
Figure 5.29: Layout of Kushiro Port (Iai et al., 1994)
82
20.0
*2.7
L.W.
Unit (ml
0
%
t. 0.0
2.9
*2.2
.1.0
Tie rod SS 41 065 L22.8
.
.
Rubble
E
tft
of
Filled Sand
Steel Pipe
-7.5
ISheet Pile
in
O
Original Ground
N
o.
before eorthquoke
otter eorthquoke
-14.5
co
-10.3
-14.3
Figure 5.30: Geometry of Kushiro Port Site C Wharf (Iai et al.,
1994)
30.0
3.0
4
20.0
2%
3.4
2.0
H.W.L1.5
L
0.0
Tie- Rod ctcl 98
0
-J
Grovel ()rain
# 400mrr
u ,-
Sond Cornpoct ion Pi le 2.
# .700mm
-12.0
4.rakraiorriAa
22.7
Unit(m)
Figure 5.31: Geometry of Kushiro Port Site D Wharf (Iai et al.,
1994)
.&
83
5.4.3.1
Soil Conditions
The boring logs for Site C and Site D are shown in Figure 5.32 and Figure 5.33,
respectively. The simplified soil conditions for the numerical analysis at Site C consisted
of 10 m of hydraulically backfilled sand, underlain by the naturally deposited sandy soil
to the depth of the model (26 m). At Site D, the soil compaction was modeled as
extending infinitely into the backland. The soil layers where simplified as a coarse sand
extending to a depth of 6 m below the surface, underlain by 12 m of fine sand, which is
underlain by another layer of coarse sand to the depth of the model (32 m). The area of
soil compaction was modeled as a uniformly compacted zone, individual gravel drains
and sand compaction piles were not modeled. The rubble mounds were modeled in both
Site C and Site D as indicated in the wharf geometries. Model soil properties used in the
analyses are given in Table 5.6. The SPT values were input as shown in the boring logs,
with the shear wave velocities estimated from the SPT values.
De p th
ml
Soil Type
SPT-N Value
0 10 20304050
'Elevotiol
s2.34rn
-5
Sand
Gravelly
-10
Sand
Fine
Sand
Gravelly
Sond
Fine
-15
Sand
Course
Sand
Sand
and
-20
Grave
.
Figure 5.32: Soil Profile at Kushiro Port Site C (Iai et al., 1994)
84
Depth
cm)
Soil Type
0
I SPT - N Volue
01020304050
levationl
2.94m J
", ?"'ZI4
Coarse Sond
-
Coarse Sond
mixed with
r,
Gravel
Fine Sand
-10 :t Fine Sand
Silt
Fine Sond
mixed with
and
_
-1-
"":
:
Shells
-15
Shelly Sond
-
Coarse Sand
mixed with
e. Grovel
Coarse Sond
before
compaction__:
:
Figure 5.33: Soil Profile at Kushiro Port Site D (Iai et al., 1994)
Table 5.6:
Model Soil Properties for Kushiro Port
Layer 1
Density (kg/m3)
Friction Angle (q)
1300
32
Poisson's Ratio
Porosity
0.30
0.40
Site C
Layer 2
1500
42
0.30
0.40
Site D
Layer 2
Layer 3
Rubble
Layer 1
1800
42
1500
45
1400
40
1500
45
1800
42
0.20
0.40
.0.30
0.40
0.30
0.40
0.30
0.40
0.20
0.40
Rubble
5.4.3.2 Structural Properties
The model structural geometry of Site C consisted of a sheet pile wall (Nippon SPZ25) anchored with battered pipe piles. The wall is 17.2 m tall, with 7 m embedment, and
the left and right anchors extend to a depth of 13 and 17 m below the ground surface,
respectively. The tie rod and anchor spacing were unknown, and assumed to be 1 m. The
model wall for Site D consisted of a steel pipe pile wall, anchored with a continuous sheet
pile wall (Nippon FSP4). The tie rod spacing is 1.98 m. The radius of the tie rod was
85
unknown for Site D, and conservatively assumed to be 1 m. The model structural
properties are given in Table 5.7.
Table 5.7:
Model Structural Properties for Kushiro Port
Property
Sheet Pile Wall:
Type
Diameter (mm)
Thickness (mm)
Moment of Inertia, I (m4)
Modulus of Elasticity, E (kPa)
Cross-Sectional Area (m2)
Site C
Site D
FSPZ25
Pipe Pile
914.4
---
16
3.824e-4
2.20e11
0.0236
4.804E-3
2.20e11
0.0460
Pipe Pile
400
FSP4
Front Anchor:
Type
Diameter (mm)
Thickness (mm)
Moment of Inertia, I (m4)
Modulus of Elasticity, E pcPa)
Cross-Sectional Area (m)
9
---
2.262e-4
1.00el1
0.0113
3.86e-4
2.20e11
0.0240
Pipe Pile
700mm
-------
Back Anchor:
Type
Diameter (mm)
Thickness (mm)
Moment of Inertia, I (m4)
Modulus of Elasticity, E (kPa)
Cross- Sectional Area (m2)
12mm
1.616e-3
1.00e 11
0.0264
Cable:
Modulus of Elasticity, E (kPa)
Radius (m)
5.4.3.3
1.00ell
0.10
_
__
1.00e 11
1.0
Earthquake Motion
The strong motion recording station included a surface and downhole seismograph
array. The downhole record was recorded at a depth of 77 m. The depth to bedrock at the
strong motion site was unknown, and assumed to be 77 m. The soil profile at Site C and
Site D was unknown for depths greater than approximately 20 m, but because the strong
motion site and wharf sites are in close proximity, the same profiles were used for depths
exceeding 20 m. The soil profile at the strong motion site is shown in Figure 5.34. The
86
Density
Elevo- Soil 1 SPT
( 1 /m3)
tionfrn) Type N-volue
5
15 2.0
0 K) 203340
is.
36
In Groun.
Surface
41.6
Seismograph
0-
1
i
5
r=--.
i..
I0
r4.
11111
1111
oorse
Sand
11
ine
and
15
111
i.
ravel ly
0
Sand
Silt
30
Ali
qualg
6.,
40
11 Ii
:...:
.1
4
--qv
- 45 -
:
M.
-
d.
....ili
bud
r
1
4
-70
Oownhole
Seismograph
.- 73.
Figure 5.34: Soil Profile at the Kushiro Port Strong Motion Site
(Iai et al., 1994)
87
recorded time histories are given in Figure 5.35 and Figure 5.36 for the surface and
downhole array, respectively. It is evident from the North record at the surface, that at
approximately 30 seconds, the predominant period of the ground motion increases. It has
been noted by Iai et al. (1995) that this is probably due to the dilatancy behavior of the
sandy soil. Regardless, the surface histories are unsatisfactory for use in the numerical
model, therefore the downhole histories were used. The recorded downhole acceleration
time histories were propagated through the soil to the depth of the model (26 and 32 m
below the ground surface for Sites C and D, respectively) using SHAKE91. Because the
orientations of Site C and Site D are almost identical, the same transformed time history
was used for both sites. The transformed record is given in Figure 5.37
North
5.0
t7,
3.0
1.0
0
-1.0
-3.0
5.0
East
5.0
fi; 3.0
E
1.0
0
=7NIZ,1:1C:!,
-1.0
-3.0
5.0
Vertical
5.0
3.0
1.0
19111.1i
i
i
i
-1.0
I, 1
,1.1
1 71!
I, ILL
771rril
JJahl
-3.0
5.0
10
20
30
40
50
Time (seconds)
Figure 5.35: Earthquake Motions Recorded at the Ground Surface
at Kushiro Port on January 15, 1993
60
88
North
5.0
a 3.0
1.0
e
1.0
-3.0
5.0
East
5.0
3.0
1.0
0
1.0
8
3.0
.0
5.0
Vertical
5.0
3.0
1.0
-1.0
-3.0
5.0
0
10
20
30
40
50
60
Time (seconds)
Figure 5.36: Earthquake Motions Recorded at a depth of 77 meters
at Kushiro Port on January 15, 1993
2.0
1.0
0.0
-1.0
-2.0
10
15
20
25
Time (seconds)
Figure 5.37: Input Time History for Kushiro Port Site C and Site D
30
89
5.4.3.4 Results of Site C
The deformed grid is shown in Figure 5.38. The calculated vertical and horizontal
time displacements at the top of the wall are shown in Figure 5.39. The measured
horizontal and vertical displacements at the top of the wall were approximately 0.6 and
0.4 m, respectively. The calculated horizontal and vertical displacements are 0.45 and
0.25 m, respectively. The post-liquefaction volumetric strain was negligible, due to
limited pore pressure generation directly adjacent to the sheet pile wall.
Figure 5.38: Kushiro Site C Deformed Grid (deformations to scale)
0.10
- Horizontal Displacement
0.00
Vertical Displacement
A
-0.10
i..-
a
A.
-0.30
V.
V
-0.40
-0.50
0
5
10
15
20
Time (seconds)
Figure 5.39: Time Displacements for Kushiro Site C
25
30
90
5.4.3.5 Results of Site D
The deformed grid is shown in Figure 5.40. There were no reported deformations at
Site D, but it was noted that no damage was observed. The calculated horizontal and
vertical time displacements are shown in Figure 5.41. The horizontal and vertical
displacements were calculated as 0.070 and 0.085 m, respectively. The post-liquefaction
volumetric strain was negligible, due to limited excess pore pressure generation in the
improved backfill soils.
Figure 5.40: Kushiro Site D Deformed Grid (deformations to scale)
0.01
0.00
Horizontal Displacement
Vertical Displacement
-0.01
-0.02
g -0.03
m -0.04
0
al -0.05
a
o
-0.06
-0.07
-0.08
-0.09
-0.10
0
5
10
15
20
Time (seconds)
Figure 5.41: Time Displacements for Kushiro Site D
25
30
91
5.4.4 Discussion of Numerical Validation Results
The results of the calculated permanent horizontal displacements at the top of the
wall and the vertical displacements in the backfill adjacent to the sheet pile wall are
tabulated in Table 5.8. Despite the numerical uncertainties and limitations noted in the
previous sections, the calculated and measured results compare rather well. The
calculated horizontal displacement for Akita Wharf 1 compares well with the measured
value, and the calculated vertical displacement appears to be acceptable, even though
there are no measured results. The horizontal and vertical displacements for Akita Wharf
2 fall within the range of measured values. The horizontal displacement for Ishinomaki
compares well with the measured value, while the vertical displacement is over-predicted.
The horizontal and vertical displacements for Kushiro Site C are both under-predicted,
while the calculated displacements for Kushiro Site D are fairly well predicted.
The horizontal displacements are on average under-predicted by approximately 23%
(as compared to the average of the maximum and average measured displacements). The
vertical displacements are on average under-predicted by approximately 29% (as
compared to the average of the maximum and average measured displacements), except
for Ishinomaki, which is largely over-predicted.
Several possible reasons for the variations in the calculated and measured values
include:
1)
soil properties, the majority of which were determined from correlations with
standard penetration test results,
2) constitutive model, which is Mohr-Coulomb based with the addition of a plastic
flow rule,
3) input earthquake motion, which was determined using the simplified dynamic
soil analysis program SHAKE91, and
4) pore pressure generation, which was a numerical variation on the simplified
method develop by Seed and his co-workers.
92
There are uncertainties in the numerical model, but it is also noted that there are
numerous uncertainties in the field case histories. There is also much variability in the
lateral and vertical deformations observed in the field (maximum displacements were
between 50 to over 100% greater than the average displacements), therefore the
numerical model is judged to provide useful, representative results for the dynamic
behavior of anchored sheet pile bulkheads. Due to the satisfactory modeling of the
validation case histories, the numerical model was used to perform parametric studies on
sheet pile bulkheads, as described in the following chapter.
Table 5.8:
Summary of Case Study Results
Calculated
Displacements (m)
Measured .,
Displacements (m)
Horizontal
Vertical*
Horizontal
Vertical
Akita Wharf 1
0.03
0.06
0.05
not reported
Akita Wharf 2
1.20
0.78
1.1 to 1.6
0.5 to 1.4
Ishinomaki Shiomi
0.60
0.30
0.60 to 1.16
0.1
Kushiro Site C
0.45
0.25
0.6
0.4
Kushiro Site D
0.07
0.09
no damage
(-0.05)
not reported
the vertical displacements include post-liquefaction volumetric strain
the range of measured displacements represent displacementaverage to displacement.
93
6
6.1
PARAMETRIC STUDY ON THE SEISMIC PERFORMANCE OF SHEET
PILE BULKHEADS
INTRODUCTION
After modeling the field case histories to calibrate the numerical model and evaluate
uncertainties in the computed deformations, the project focused on an extensive
parametric study using the calibrated model for anchored sheet pile bulkheads. The
parametric study included several analyses on the effects of varying; 1) zone of soil
improvement, 2) density of the backfill, 3) length of the tie rod, 4) stiffness of the sheet
pile wall, 5) the depth of sheet pile embedment, and 6) characteristics of the earthquake
motions. The study focused on walls designed using standard pseudo-static procedures
for anchored sheet pile walls without pore pressure generation (Ebeling and Morrison,
1993).
The walls were designed with heights ranging from 7.5 to 15 m, and horizontal
seismic coefficients ranging from 0.1 to 0.16. In most cases, seismic coefficients in
excess of 0.2 were not possible for the simplified walls with single anchor systems due to
the impracticable section properties required of the sheet piles. The vertical seismic
coefficient was assumed to be zero. The design followed the steps as outlined in Ebeling
and Morrison (1993), and included a factor of safety of 1.2 applied to the passive
resistance of the dredge soils in front of the sheet pile wall. Table 6.1 presents the results
of the design for the bulkheads utilized in this study.
The tie rod length was designed such that the active failure plane behind the wall
does not intersect the passive failure plane in front of the anchor. The active and passive
failure planes were assumed to originate at the base of the wall and anchor, respectively.
The anchor was assumed to be a continuous sheet pile wall, with the same stiffness as the
bulkhead wall, and the tie rod spacing was assumed to be two meters. The distance from
the ground surface to the ground water table and the tie rod was assumed to be 27 percent
of the wall height (I/), which is on average the elevation of the tie rod for many field
bulkheads.
94
Table 6.1:
Wall Height,
H (m)
7.5
7.5
15.0
15.0
Bulkhead Properties for Parametric Study
Anchor
Embedment
Coefficient, kh
Depth of
Embedment,
D (m)
0.10
0.16
0.10
0.16
2.9
3.7
5.4
6.9
4.5
6.0
7.6
9.9
Horizontal
Seismic
(m)
Tie Rod
Length, A (m)
Moment of
Inertia, I (m)
4
16
3.45x10-5
7.43 x10-5
22
30
40
3.82x10
8.11x104
The foundation and backfill soils are both cohesionless materials, each layer modeled
with uniform density (except for soil improvement zones within the backfill). The range
of densities used can be correlated with stress corrected penetration resistances ((N1)60) of
a) 25 blows/30 cm in the foundation soil, b) 10 to 25 blows/30 cm in unimproved
backfill, and c) 30 blows/30 cm in improved backfill. The soil improvement was modeled
as providing a uniform increase in soil density throughout the zone of treatment. This is a
simplification of the actual pattern of densification and soil stiffness variations that would
be expected from vibro-compaction, stone columns or other densification methods,
however it adequately modeled the general soil density increase due to improvement.
Five earthquake motions covering the magnitude range of engineering interest
(M 6 to 8) were selected for the parametric study (Table 6.2). The selected acceleration
time histories are slightly conservative in the sense that each one is characterized as
having greater than average duration for that magnitude, thereby yielding slightly
conservative displacement results. Each record was scaled to different maximum
acceleration values ranging from 0.1 to 0.4 g. Plots of the unsealed records are provided
in Appendix A.
In order to account for the duration of the earthquake motions a normalized ground
motion intensity parameter has been developed. This parameter is defined as the
maximum horizontal acceleration within the backfill at the elevation of the dredge line
(A max@dredge) divided by the appropriate magnitude scaling factor (MSF) given in Table
2.2. It is recommended that if a site specific seismic study is not performed to determine
A max@dredge 5
that the peak ground surface acceleration be reduced using the reduction
95
factor (rd) developed for estimating the variation of cyclic shear stress (or acceleration)
with depth (e.g., Seed and de Alba, 1983). The values of rd for 15 and 7.5 m walls are
approximately 0.78 and 0.95, respectively. It should be noted that the reduction factor
was developed using a linear one-dimensional dynamic soil response method and will
only yield approximate acceleration values for the two-dimensional soil-structure
interaction applications discussed herein.
Table 6.2:
Parametric Study Earthquake Motions
Earthquake
1984 Morgan Hills EQ - Gilroy #4
1989 Loma Prieta EQ Capitola Fire Station
1989 Loma Prieta EQ Salinas
1992 Landers EQ - Joshua Tree Fire Station
1985 Michoacan Mexico EQ
Moment
Magnitude
6.0
6.9
6.9
7.4
8.1
Recorded
(g)
0.22
0.40
0.11
0.27
0.39
Am
Notable conditions in the parametric numerical models include:
1) no pore pressure generation in the foundation or improved backfill soils
((NI)60-30),
2) the soil improvement extends vertically to the base of the backfill (i.e. the dredge
line elevation), and
3) the tie rod and water level are located at an elevation that is 27 percent of H from
the top of the wall.
Parametric studies were then formulated to examine the effects of varying several
parameters when subjected to seismic motions. The parametric studies included the
following (with results and discussions presented in the following sections and tabulated
results in Appendix B):
96
1) depth of sheet pile embedment (Section 6.2),
2) tie rod length (Section 6.3),
3) sheet pile stiffness (Section 6.4),
4) density of the liquefiable backfill (Section 6.5), and
5) zone of soil improvement (Section 6.6).
6.2
DEPTH OF SHEET PILE EMBEDMENT
A study was performed to determine the effect on the lateral displacements by
varying the depth of embedment from the value determined using the 7.5 m bulkheads
designed with a kh of 0.1. The depths of embedment were halved and doubled from the
design value of 2.9 m, and the horizontal displacements at the top of the wall were
monitored. The 1984 Morgan Hills Earthquake and the 1985 Mexico Earthquake were
used. The entire backfill was modeled as improved, with (A/d60 values of 30 blows/30 cm.
The monitored horizontal displacements for varying the depths of embedment are
presented in Figure 6.1. The depth of embedment (D) is normalized by the depth of
embedment calculated using the Mononobe-Okabe method (Dm_ 0) and the earthquake
intensity at the dredge line has also been normalized by the magnitude scaling factor
(MSF). Even though there are a limited number of data points, it is noted that for these
specific cases, the depth of embedment has very little effect on the horizontal
displacements at the top of the wall.
6.3
TIE ROD LENGTH
A study was performed to determine the effect on the lateral displacements by varying the
tie rod length from the value determined using the 7.5 m bulkheads designed with a kh of
0.1. The tie rod lengths were halved and doubled from the design value of 16 m, and the
horizontal displacements at the top of the wall were monitored. The 1984 Morgan Hills
Earthquake, 1992 Landers (Joshua Tree) Earthquake and the 1985 Mexico Earthquake
were used. The entire backfill was modeled with (N 1)60 values of 30 blows/30 cm.
97
1.50
Earthquake (Amax@dredge/MSF)
Morgan Hills (0.26)
1.25
Mexico (0.94)
1.00
0.75
0.50
0.25
0.00
0.0
0.5
1.0
1.5
2.0
2.5
Normalized Depth of Embedment ( DID wo
Figure 6.1: Top of Wall Displacements for the Depth of
Embedment Study
The monitored horizontal displacements for varying the tie rod length are presented
in Figure 6.2. The tie rod length (A) is normalized by tie rod length calculated using the
Mononobe-Okabe method (Am_0), as defined in Section 6.1. The earthquake intensity at
the dredge line has been normalized by the magnitude scaling factor (MSF). Even though
there are a limited number of data points, it is noted that for these specific cases, the tie
rod length has a large effect on the horizontal displacements at the top of the wall. This is
especially observed when the normalized tie rod length is less than 1.0.
6.4
SHEET PILE STIFFNESS
A study was performed to determine the effect on the lateral displacements by
varying the sheet pile stiffness (El) from the value determined using the 7.5 m bulkheads
98
1.50
Earthquake (Amax@dredge/MSF)
Morgan Hills (0.29)
1.25
Joshua Tree (0.50)
Mexico (0.94)
1.00
0.75
0.50
0.25
0.00
0.0
0.5
1.0
1.5
Normalized Tie Rod Length ( A/A
2.0
2.5
)
Figure 6.2: Top of Wall Displacements for the Tie Rod Length
Study
designed with a kh of 0.1. The stiffness was halved and doubled from the design, and the
horizontal displacements at the top of the wall were monitored. The 1984 Morgan Hills
Earthquake and the 1985 Mexico Earthquake were used. The entire backfill was modeled
as improved, with (Nj)60 values of 30 blows/30 cm for two sets of runs. For the other run
(hollow diamonds in the figure) the backfill was modeled as partially improved to a
normalized soil improvement value ( n = SI/(H+D) ) of 2.0, while the remaining
backland had a (Nj)60 values of 10 blows/30 cm. The normalized soil improvement
parameter is discussed in more detail in Section 6.6.
The monitored horizontal displacements for varying the stiffness are presented in
Figure 6.3. The stiffness (El) is normalized by the stiffness calculated using the
Mononobe-Okabe method (E/m_o) and the earthquake intensity at the dredge line has also
99
been normalized by the magnitude scaling factor (MSF). Even though there are a limited
number of data points, it is noted that for these specific cases, the sheet pile stiffness has a
negligible effect on the horizontal displacements at the top of the wall. It should be noted
that if the plastic moment of the sheet pile wall is exceeded, the deformations at the top of
the wall will increase dramatically.
1.50
Earthquake, Wall Height, (Amax@dredge/MSF)
Morgan Hills, H=7.5m (0.29)
1.25
o Morgan Hills, H=7.5m, n=2 (0.29)
Mexico, H=7.5m (0.94)
1.00
0.75
0.50
6
8
0.5
1.0
8
0.25
0.00
0.0
1.5
2.0
2.5
Normalized Stiffness ( El/EI m.o )
Figure 6.3: Top of Wall Displacements for the Sheet Pile Stiffness
Study
6.5
PENETRATION RESISTANCE OF THE BACKFILL
A study was performed to determine the effect on the lateral displacements by
varying the normalized penetration resistance ((N/)60) of the backfill. The study utilized
the 7.5 m bulkheads designed with a kh value of 0.1. The (N/)60 values were varied
between 10 and 25 blows/30 cm, and the horizontal displacements at the top of the wall
100
were monitored. The 1984 Morgan Hills Earthquake, 1992 Landers (Joshua Tree)
Earthquake and the 1985 Mexico Earthquake were used. The entire backfill was modeled
as uniformly dense.
The monitored horizontal displacements for varying the penetration resistance are
presented in Figure 6.4. Even though there are a limited number of data points, it is noted
that for these specific cases, the penetration resistance of the backfill has a large effect on
the horizontal displacements at the top of the wall. An especially large increase is noted
for the Morgan Hills record when the (Nj)60 values were less than 15 blows/30 cm. The
analyses for the Joshua Tree and Mexico earthquakes with (Nj)60 values less than 15 and
20 blows/30 cm, respectively, experienced complete flow failures of the bulkheads and
caused instability in the numerical models.
4.5
Earthquake (Amax@dredge/MSF)
Morgan Hills (0.29)
Joshua Tree (0.50)
Mexico (0.94)
4.0
........
E
35
g 3.0
E
a)
C 2.5
6
2.0
To
o 1.5
N
T-*
0
1.
0
0.5
0.0
5
10
15
20
Penetration Resistance (blows/30 cm)
Figure 6.4: Top of Wall Displacements for the Penetration
Resistance of the Backfill Study
25
30
101
6.6
SOIL IMPROVEMENT
The largest parametric study was to determine the effect on the lateral displacements
by varying the zone of soil improvement. The study utilized the 7.5 and 15 m bulkheads
designed with kh values of 0.1 and 0.16. The standard penetration of the improved and
backfill soils was 30 blows/30 cm, and 10 blows/30 cm, respectively. All five earthquake
motions were used, with each one scaled to three different peak accelerations.
A typical diagram of the soil improvement parameters is presented in Figure 6.5. The
distance from the wall to the most landward portion of the improved soil region (SI) has
been normalized by the total wall height (H+D). The resulting factor, n, is given as the
normalized width of soil improvement (SI/(H+D)). The monitored horizontal
displacements for varying n are presented in Figure 6.6. The data is plotted as ranges of
normalized earthquake intensity because of the numerous data points with varying
intensities. There is a definite trend noted in Figure 6.6 that as the amount of soil
improvement decreases, the displacements increase. The scatter within each intensity
range is due to the variations in the earthquake motions that were not accounted for in the
normalized factor (i.e. frequency content). There is no distinction made in the plot
between the different wall heights and different kh values due to the number of data
Figure 6.5: Definition of Soil Improvement Variables
102
points. It should be noted that even without a distinction between different wall heights
and designs, the scatter in the data for each range is quite small, given the wide range of
earthquake motions used in the analyses.
1.6
Amax@dredge/MSF
o 0.0 to 0.2
0.2 to 0.4
o 0.4 to 0.6
0.6 to 0.8
1.4
1.2
1.0
o0.8to1.1
0.8
0.6
0.4
0
0.2
9
0.0
0
1
2
3
4
5
6
7
8
Normalized Soil Improvement, n ( S1/(H+D) )
Figure 6.6: Horizontal Displacments at the Top of the Wall for
Variations in the Zone of Soil Improvement
6.7
COMPARISON OF THE PARAMETRIC STUDY WITH AN EXISTING DEFORMATIONBASED DESIGN METHOD
The results of the parametric studies which included non-liquefiable soils (the entire
backland was improved) can be compared to the design chart that has been proposed by
Dennehy (1985) and Gazetas et al. (1990). The location of four points was determined
using the method outlined in Gazetas et al. (1990). Figure 6.7 presents the comparison
between the parametric study (for non-liquefiable soils) plotted as solid stars (with the
103
calculated displacements in parenthesis) and the design chart presented by Dennehy and
Gazetas. One point (center-right) is the standard 7.5 m wall presented in Table 6.1. Two
points (top-right and bottom-right) are from the parametric study varying the length of the
tie rod anchor, and the last point (left-center) comes from the parametric study varying the
depth of embedment.
It is evident from Figure 6.7 that the computed deformations vary significantly at
each data point, and that some of the plotted points have calculated displacements that
both fit, and do not fit the proposed design chart (especially the point on the center-right).
The variations in the displacement values for each of the plotted parametric study points
can be attributed to variations in the earthquake motions. Larger earthquake motions
produced larger displacements, whereas the method proposed by Dennehy and Gazetas
does not directly include any earthquake motion parameters (intensity, frequency or
duration). It is significant to note that many of the computed deformations that fall in
Zone I (deformations approximately less than 10 cm) would be considered unacceptable
by many port engineers for an operating or contingency level earthquake motion (Ferritto,
1997).
A comparison can also be made between the parametric study on the depth of sheet
pile embedment (Section 6.2) and the proposed chart by Dennehy and Gazetas. It was
noted from the parametric study that the depth of embedment had very little effect on the
performance of the bulkhead over the range of the modeled values, but the contour lines
constructed by Dennehy and Gazetas show a clear variation in performance over the
range of interest (EPI = 0.25 to 0.75).
6.8
DEVELOPMENT OF A NEW SEISMIC DESIGN METHOD
The main objective of this research was the development of a new seismic design
chart and recommendations for deformation-based design. In the development of a
seismic design chart, the results of the parametric study have been synthesized into
normalized parameters, where possible, to incorporate key variables into straightforward
design parameters. For example, aspects of the wall geometry and sheet pile stiffness
104
2
1.5
1
0.5
0
-0.5
0.5
0
1
Embedment participation index (EPI)
Figure 6.7: The Developed Seismic Design Chart (Gazetas et al.,
1990), Including Data from the Parametric Study for
Models with Improved Soils (solid stars)
have been combined as into a flexibility factor, (EI /(H +D)5), where El is the stiffness of
the wall, H is the height of the wall above the dredge line, and D is the depth of sheet pile
embedment. The computed displacements at the top of the wall (La) have been combined
with the flexibility factor and the buoyant unit weight of the soil adjacent to the wall (fl),
to yield a deformation factor as follows;
105
AX EI
(H +D)5
6.1
Yb
The results of the parametric study are presented in Figure 6.8. The contour lines
indicate various levels of earthquake intensity for backfill soils with blowcounts of 10
and 20 blows/30 cm. The effectiveness of soil improvement for minimizing bulkhead
deformations is clearly demonstrated by the design chart. It is also noted that an
incremental benefit of soil improvement beyond n values of approximately 2.0 decreases
considerably. In comparison, the n values as determined from the PHRI (1997)
recommendations for the parametric study sheet pile bulkhead geometries are
approximately 1.9 to 2.5.
There are fourteen case histories plotted on the chart, which are arranged according
to the blowcounts of the backfill soils. The case history data is summarized in Table 6.3.
It should be noted that seven case histories are closely predicted by the chart (within
10%), five case histories are significantly over-predicted
10%) and only two of the
case histories are significantly under-predicted (5_ 10%). These results indicate that the
design chart can be used as a conservative preliminary design chart or screening tool.
The results of the study indicate that it would be very difficult to limit the
deformations to 10 cm utilizing only densification methods of soil improvement for
moderate to high earthquake motions (Amax ?_ 0.3g). In cases such as these, it may be
necessary to consider other soil improvement techniques (e.g. grouting, soil mixing, etc.)
or structural improvements to strengthen the wall and/or anchor systems.
The chart has a relative error of approximately 30% for any of the calculated values,
this is due to the variations in earthquake motion that are not accounted for (e.g.
frequency) and other unknown wall behaviors. The value of 30% was determined from
the results of the validation case histories and from the variation of values on the design
chart that should be equal (e.g. normalized displacements for different wall heights).
106
0.014
I
1
1
0.012
1
Penetration Resistance of the
Unimproved Backfill Soil:
0 510
(N1)60 (blows/30 cm)
0
0.010
0.008
1
Case History
(N1)60 (blows/30 cm)
_
- 10
20
15
20
--
(
1
t
0.006
1
L
%
I
0.004
Ground Motion Intensity Factors
( Anwor.dg./MSF )
SIM
1
0.002
0.000
0
Ill
1
05
0I 3
01
4
6
5
Normalized Soil Improvement, n ( SW(H+D) )
2
3
7
8
Figure 6.8: Design Chart for Lateral Displacements
Table 6.3:
Earthquake
1968 Tokachi-Oki
1993 Kushiro-Oki
1964 Niigita
1983 Nihonkai-Chubu
1968 Tokachi-Oki
1968 Tokachi-Oki
1968 Tokachi-Oki
1973 Nemuro-Hanto
1978 Miyagi-Ken-Oki
1968 Tokachi-Oki
1983 Nihonkai -Chubu
1993 Kushiro-Oki
1993 Guam
1993 Kushiro-Oki
Plotted Case History Validations
(N d 60
6
6
10
10
15
15
15
15
15
20
20
20
20
30
19
Normalized
Displacement
AXE1/((H+D)511)
0.0120
0.0020
100
0.0031
110 to 160
16 to 57
12 to 19
30
30
87 to 116
0.0066
0.0090
0.0029
0.0077
A,,*,fredge/MSF
Displacement
(g)
(cm)
0.26
0.20
0.14
0.10
0.23
0.23
0.12
0.16
0.10
0.26
0.09
0.16
0.18
12 to 23
0.21
-60
-5
50 to 70
61
no damage (-5)
0.0031
0.0129
0.0049
0.0006
0.0044
0.0019
0.0004
107
6.9
RECOMMENDATION DESIGN PROCEDURES
The design chart is useful in estimating lateral deformations, but it is not
recommended for use in specifying bulkhead properties (stiffness, depth of embedment,
tie rod length, etc.). These should be specified by utilizing the current "standard of
practice" design methods, as outlined in Chapter 4. The recommended procedures for
utilizing the results of the parametric study to estimate the permanent displacement at the
top of sheet pile walls include;
1) determine the wall specifications (H, D, EI, and anchor length), either through a
new design or from an in-situ wall,
2) determine A,,,,@dredge with a site-specific seismic study or approximate with
empirical soil amplification factors to yield the peak ground surface acceleration
and the reduction factor (rd),
3) determine the magnitude scaling factor for the specified earthquake magnitude,
and divide into Amax@dredge to yield the intensity factor, and
4) based on the standard penetration resistance of the backfill soils, the width of the
ground treatment behind the sheet pile wall, and the ground motion intensity
factor, enter Figure 6.8 and obtain the deformation factor. From this, the
deformation at the top of the wall (dX) can be estimated.
Using these recommended procedures it is possible for an engineer to determine the
deformations of a new or in-situ wall. It is also possible to estimate the zone of soil
improvement necessary to limit deformation to an acceptable limit. An example design
problem is provided in the following section.
108
6.10 EXAMPLE DESIGN PROBLEM
This section provides the calculations involved in determining the extent of soil
improvement necessary to limit wall deformations for a sheet pile bulkhead, using the
developed design method from Section 6.8. The example bulkhead is shown in Figure
6.9, and was designed using the Mononobe-Okabe Method. The example problem was
taken from Appendix C in The Seismic Design of Waterfront Retaining Structures by
Ebeling and Morrison (1993). The example wall was designed using the standard of
practice Mononobe-Okabe design methods with a factor of safety of 1.5 applied to the
passive earth pressure coefficient. This was a minimum design, no consideration was
given for scour, dredging, corrosion or other design considerations.
This example illustrates the use of Operating and Contingency Level earthquake
motions that are likely for design in the San Francisco-Oakland Bay area. The Operating
Level earthquake motion (OLE) is estimated as a magnitude 6.5 earthquake with peak
ground accelerations of 0.25 g. The Contingency Level earthquake motion (CLE) is
estimated as a magnitude 7.5 earthquake with peak ground accelerations of 0.45 g. The
following design example uses both levels of earthquake motion to estimate the expected
deformations for various extents of densification improved soil.
improved soil
= 19 kN/m3
(Ndso 15 blows/30 cm
9 kN/m3
backfill soil
7v1440'114k
foundation soil
sheet pile wall (PZ27)
Figure 6.9: Anchored Sheet Pile Bulkhead Design Problem
109
The sheet pile wall values for a PZ27 section are:
E = 2x1011 Pa
I= 2.51x104 m4/m
The wall height (H) is 9 m, and the wall depth (D) is 6 m.
The soil buoyant unit weight (A) directly adjacent to the wall is 9000 N/m3.
Substituting into the normalized lateral displacement factor (Equation 6.1) yields:
AX EI
AX (2 x 10" Pa). (2.51x10-4 m4/m) AX
(H + D)5 7b
(9 m + 6 m)5 (9000 N/m3)
136
The ground motion intensity factor is determined for the earthquake motions using
Table 2.2 to determine the MSF and Figure 2.1 to determine rd, using the following
equation:
Amax@ dredge
MSF
Amax@ surface
rd
MSF
The magnitude scaling factor values for both earthquake motions are tabulated in
Table 6.4.
Table 6.4:
Magnitude Scaling Factors for Design Example
A max@surface
rd (at 9 m)
A mar@d redge
MSF
A max*Iredge
OLE
0.25
0.92
0.23
1.63
0.14
CLE
0.45
0.92
0.41
1.00
0.41
MSF
110
The normalized lateral displacement factor is then determined as a function of soil
improvement (n) using Figure 6.8 for blowcounts of 15 and a maximum acceleration of
0.2 g, as provided in Table 6.5.
Table 6.5:
OLE
CLE
Lateral Displacements for Design Example
Normalized Soil
Improvement (n)
Normalized Lateral
Displacement (b)
Lateral Displacement
(8x 136) (m)
Displacement (cm)
0
0.0050
0.68
68
1
0.0018
0.24
24
2
0.0007
0.10
10
3
0.0005
0.07
7
8
0.0004
0.05
5
0
0.0130
1.77
177
1
0.0041
0.56
56
2
0.0021
0.29
29
3
0.0018
0.24
24
8
0.0017
0.23
23
Lateral
By plotting the values tabulated above, it is possible to develop a figure relating the
extent of soil improvement to lateral displacements (Figure 6.10). It is interesting to note
that for this example, it is impossible to limit the deformations to 5 cm for an OLE
motion and 23 cm for a CLE motion, using only soil improvement techniques. Other
forms of earthquake mitigation are necessary to limit earthquake-induced deformations to
smaller values. The design tie rod length was 28 m, which is at the break in the curve in
Figure 6.10. Improvement beyond this point has little effect on the earthquake-induced
lateral displacements.
It should also be noted that the example problem varied from the developed design
method in that the tie rod and water elevations were not equal.
111
200
180
OLE motion
CLE motion
160
E
140
E 120
(.3
as
100
c.
80
11'
co
60
ea
40
20
0
0
20
40
60
80
Soil Improvement (m)
Figure 6.10: Lateral Displacements for the Design Problem
100
120
112
7
SUMMARY AND CONCLUSIONS
There have been numerous recent earthquakes, which have resulted in severe damage
to many international ports. The damage has been due largely to the liquefaction of
saturated, sandy soils. Waterfront structures, particularly anchored sheet pile bulkheads
are very susceptible to liquefaction-induced damages. Due to the recent damage caused
by liquefiable soils, many port authorities are instigating soil improvement programs to
limit the generation of excess pore pressures during design level earthquakes. The case
histories of liquefaction-damaged sheet pile bulkheads have highlighted the need for a
simplified, performance-based design method.
The "standard of practice" seismic design procedures for sheet pile wall involves
using pseudo-static, limit-equilibrium methods, developed for rigid retaining walls. There
are several recent modifications to the "standard of practice" design methods, but are very
limited in their applicability to design. The method proposed by Dennehy (1985) and
Gazetas et al. (1990) is performance-based, but is limited to non-liquefiable soils. The
other recent additions to sheet pile design are limit-equilibrium based.
A study that incorporates field performance and numerical modeling has been
conducted using a validated numerical model, utilizing several empirical relationships
developed for geotechnical earthquake parameters. The study involved determining the
effects of varying several design parameters, including the depth of sheet pile
embedment, tie rod length, sheet pile stiffness, penetration resistance of the backfill,
extent of soil improvement, and ground motion characteristics. Charts were developed
depicting the effect of varying these various parameters. A final design chart was
developed utilizing the results from the study, and includes data from several case
histories. There is some inherent scatter in the data due to the variability of the earthquake
motions, but the design chart provides a reasonable method for estimating lateral
deformations of sheet pile bulkheads with or without soil improvement. This design chart
is applicable for the preliminary design of new bulkheads and as a screening tool for
existing bulkheads.
113
This investigation has highlighted the following pertinent aspects of the seismic
design and performance of anchored sheet pile walls;
from a practical perspective, the "failure" of an anchored sheet pile bulkhead can
be considered as corresponding to lateral deformations in the range of 10 cm to 20
cm,
the factors of safety computed with "standard of practice" design methods are not
adequately correlated with wall deformations to facilitate estimates of seismicallyinduced lateral displacements,
in order to estimate the deformations of sheet pile walls, the intensity and duration
of the earthquake, and potential pore pressure generation in the backfill and
foundation soils must be evaluated,
in light of the flexible nature of anchored sheet pile walls, lateral deformations
should be anticipated even in competent, non-liquefiable soils subjected to
moderate- to high-intensity ground motions. In relationship to the Navy guidelines
proposed by Ferritto (1997) for sheet pile walls, it may be impossible to design a
sheet pile bulkhead to stay within the recommended deformations utilizing only
soil densification techniques, and
a simplified method of estimating seismically induced lateral deformations for
anchored sheet pile bulkheads in unimproved or improved soils has been
proposed.
Several recommendations for future work are also noted, including;
studies to examine the effects of different tie rod elevations and other structural
properties such as the anchor and tie rod stiffness, and different anchor
configurations (such as single pile, deadman, or battered piles),
studies that examine the effect of pore pressured dissipation, as well as generation,
114
studies that examine other bulkhead variables, such as liquefiable foundation
soils, cohesive foundation soils,
studies on the specific effects of various soil improvement strategies (e.g. vibrocompaction, soil mixing, deep dynamic compaction), and
utilize scale-model testing (i.e. centrifuge) for validation with case history and
numerical model studies.
115
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122
APPENDICES
123
APPENDIX A
(EARTHQUAKE MOTIONS USED IN THE PARAMETRIC STUDY)
124
1984 Morgan Hill EQ- Gilroy #4 (Mw.0, Am ax.222g)
4
E
2
0
O
2
4
1989 Loma Prieta EQ - Captola Fire Station (Mw.9, Amax=-0.399)
E
2
o
0
0
O
111111111111
iTIMIP,111111111,011111.1111
2
11E11
4
1989 Loma Prieta EQ - Salinas (Mw.9, Amax=-0.112)
0
-2
1992 Landers EQ - Joshua Tree Fire Station (Mw=7.4, Am ax).274)
1985 Michoacan Mexico EQ (Mw=8.1, Scaled to Amax.350g)
I
I
rrirl T rip prrr
RI
g
0
'S
2
u
a
INN
I
0
10
20
30
Time (seconds)
40
50
60
125
APPENDIX B
(TABLE OF PARAMETRIC STUDY RESULTS)
126
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7.5
0.10
Morgan
all
0.30
0.52
0.26
72.8
5.8
15.9
3.45
30
10.4
6.0
0.29
0.37
1.4
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2.4
2.6
0.0005
7.5
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Morgan
all
0.30
0.52
0.26
72.8
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15.9
3.45
30
10.4
6.0
0.32
0.40
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2.4
2.6
0.0017
7.5
0.10
Morgan
all
0.30
0.52
0.26
72.8
1.5
15.9
3.45
30
10.4
6.0
0.31
0.38
1.3
2.2
2.4
2.6
0.0036
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
1.5
15.9
3.45
30
10.4
8.1
0.64
0.82
1.3
2.2
2.3
2.6
0.0074
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
2.9
15.9
3.45
30
10.4
8.1
0.61
0.80
-2.3
2.2
2.5
2.6
0.0033
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
5.8
15.9
3.45
30
10.4
8.1
0.59
0.78
-2.1
2.2
2.5
2.6
0.0009
Variations is Stiffness
7.5
0.10
Morgan
all
0.30
0.59
0.29
72.8
2.9
15.9
1.73
30
10.4
6.0
0.34
0.42
1.2
2.2
2.3
2.6
0.0009
7.5
0.10
Morgan
all
0.30
0.59
0.29
72.8
2.9
15.9
3.45
30
10.4
6.0
0.31
0.28
1.2
2.2
2.3
2.6
0.0017
7.5
0.10
Morgan
all
0.30
0.59
0.29
72.8
2.9
15.9
6.90
30
10.4
6.0
0.30
0.37
-2.0
2.2
2.4
2.6
0.0033
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
2.9
15.9
1,73
30
10.4
8.1
0.63
0.85
-1.5
2.2
2.4
2.6
0.0017
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
2.9
15.9
3.45
30
10.4
8.1
0.62
0.67
-2.2
2.2
2.3
2.6
0.0034
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
2.9
15.9
6.90
30
10.4
8.1
0.62
0.75
-3.0
2.2
2.6
2.6
0.0067
15.0 0.10
Morgan
all
0.30
0.44
0.22
142.0
5.4
30.0
19.12
30
10.4
6.0
0.43
0.54
8.4
13.0
7.9
8.7
0.0004
15.0 0.10
Morgan
all
0.30
0.44
0.22
142.0
5.4
30.0
76.48
30
10.4
6.0
0.36
0.46
-9.4
13.0
8.2
8.7
0.0015
15.0 0.10
Mexico
all
0.30
0.48
0.68
142.0
5.4
30.0
19.12
30
10.4
8.1
0.61
0.81
9.1
13.0
8.1
8.7
0.0006
7.5
0.10
Morgan
1.9
0.30
0.59
0.29
20.2
2.9
15.9
1.73
10
10.4
6.0
0.35
0.40
1.0
2.2
1.9
2.6
0.0010
7.5
0.10
Morgan
1.9
0.30
0.59
0.29
20.2
2.9
15.9
3.45
10
10.4
6.0
0.34
0.37
1.1
2.2
1.9
2.6
0.0018
7.5
0.10
Morgan
19
0.30
0.59
0.29
20.2
2.9
15.9
6.90
10
10,4
6.0
0.33
0.35
-1.7
2.2
2.0
2.6
0.0036
Variations is Tie Rod Length
7.5 0.10
Morgan
all
0.10
0.20
0.10
72.8
2.9
8.0
3.45
30
10.4
6.0
0.01
0.01
0.8
2.2
1.7
2.6
0.0001
7.5
0.10
Morgan
all
0.30
0.59
0.29
72.8
2.9
8.0
3.45
30
10.4
6.0
0.48
0.44
-1.1
2.2
1.1
2.6
0.0026
7.5
0.10
Morgan
all
0.30
0.59
0.29
72.8
2.9
15.9
3.45
30
10.4
6.0
0.22
0.28
1.2
2.2
2.3
2.6
0.0012
7.5
0.10
Morgan
all
0.30
0_59
029
72.8
2.9
31.8
3.45
30
10.4
6.0
0.07
0.22
1.4
2.2
2.6
2.6
0.0004
7.5
0.10
Joshua
all
027
0.52
0.50
72.8
2.9
8.0
3.45
30
10.4
7.4
0.69
0.60
1.4
2.2
2.7
2.6
0.0038
7.5
0.10
Joshua
all
0.27
0.52
0.50
72.8
2.9
15.9
3.45
30
10.4
7.4
0.31
0.37
1.3
2.2
2.4
2.6
0.0017
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
2.9
8.0
3.45
30
10.4
8.1
1.19
1.00
-1.5
2.2
2.5
2.6
0.0065
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
2.9
15.9
3.45
30
10.4
8.1
0.50
0.67
-2.2
2.2
2.3
2.6
0.0028
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
2.9
31.8
3.45
30
10.4
8.1
0.13
0.51
-2.5
2.2
2.4
2.6
0.0007
Variations in the Penet ation Resistance of the Backfill
7.5
0.10
Morgan
0.0
0.30
0.59
0.29
0.0
2.9
15.9
3.45
10
7.2
6.0
3.50
1.79
3.2
2.2
1.6
2.6
0.0276
7.5
0.10
Morgan
0.0
0.30
0.59
0.29
0.0
2.9
15.9
3.45
15
8.4
6.0
1.51
0.94
-1.5
2.2
1.8
2.6
0.0102
7.5
0.10
Morgan
0.0
0.30
0.59
0.29
0.0
2.9
15.9
3.45
20
9.2
6.0
0.82
0.69
-2.1
2.2
1.8
2.6
0.0051
7.5
0.10
Morgan
0.0
0.30
0.59
0.29
20.2
2.9
15.9
3.45
20
10.4
6.0
0.31
0.35
1.1
2.2
2.0
2.6
0.0017
7.5
0.10
Morgan
0.0
0.30
0.59
0.29
0.0
2.9
15.9
3.45
25
9.8
6.0
0.50
0.51
-1.9
2.2
1.9
2.6
0.0029
7.5
0.10
Joshua
0.0
0.27
0.52
0.50
0.0
2.9
15.9
3.45
15
8.4
7.4
2.86
1.55
1.9
2.2
1.7
2.6
0.0194
7.5
0.10
Joshua
0.0
0.27
0.52
0.50
0.0
2.9
15.9
3.45
25
9.8
7.4
0.79
0.72
-2.1
2.2
1.9
2.6
0.0045
7.5
0.10
Mexico
0.0
0.30
0.66
0.94
0.0
2.9
las
3.45
20
9.2
8.1
2.57
1.60
-1.5
2.2
1.9
2.6
0.0159
7.5
0.10
Mexico
0.0
0.30
0.66
0.94
0.0
2.9
15.9
3.45
25
9.8
8.1
1.56
1.19
-2.2
2.2
1.8
2.6
0.0090
Variations in the Zone of Soil Improvement
15.0 0.10
Morgan
1.9
0.10
0.15
0.07
38.7
5.4
30.0
38.24
10
10.4
6.0
0.05
0.06
4.7
13.0
5.4
8.7
0.0001
0.10
Morgan
1.9
0.10
0.20
0.10
20.2
2.9
15.9
3.45
10
10.4
6.0
0.04
0.05
0.7
2.2
1.6
2.6
0.0002
15.0 0.10
Capitola
1.9
0.10
0.14
0.11
38.7
5.4
30.0
38.24
10
10.4
6.9
0.05
0.06
4.5
13.0
5.3
8.7
0.0001
15.0 0.10
Mexico
1.9
0.05
0.08
0.11
38.7
5.4
30.0
38.24
10
10.4
8.1
0.02
0.02
3.9
13.0
4.8
8.7
0.0000
7.5
0.10
Salinas
0.8
0.11
0.16
0.12
8.0
2.9
15.9
3.45
10
10.4
6.9
0.36
0.28
-1.1
2.2
1.4
2.6
0.0020
7.5
0.10
Salinas
1.5
0.11
0.16
0.12
15.9
2.9
15.9
3.45
10
10.4
6.9
0.13
0.13
0.8
2.2
1.5
2.6
0.0007
7.5
0.10
Salinas
1.9
0.11
0.16
0.12
20.2
2.9
15.9
3.45
10
10.4
6.9
0.08
0.09
0.8
2.2
1.6
2.6
0.0004
7.5
0.16
Salinas
2.4
0.11
0.16
0.12
26.3
3.7
22.0
7.43
10
10.4
6.9
0.04
0.06
1.0
2.9
2.0
3.6
0.0004
7.5
0.10
Salinas
2.9
0.11
0.16
0.12
30.0
2.9
15.9
3.45
10
10.4
6.9
0.05
0.06
0.9
2.2
1.9
2.6
0.0003
7.5
127
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= 1- S m
0.10
Salinas
all
0.11
0.16
0.12
72.8
2.9
15.9
3.45
30
10.4
6.9
0.05
0.06
1.0
2.2
2.0
2.6
0.0002
15.0 0.10
Salinas
1.9
0.11
0.18
0.14
38.7
5.4
30.0
38.24
10
10.4
6.9
0.07
0.09
5.3
13.0
5.7
8.7
0.0002
15.0 0.16
Salinas
2.2
0.11
0.18
0.14
48.2
6.9
39.5
92.45
10
10.4
6.9
0.04
0.06
6.3
17.9
6.4 11.9
0.0001
15.0 0.10
Morgan
1.9
0.22
0.33
0.16
38.7
5.4
30.0
38.24
10
10.4
6.0
0.25
0.28
6.8
13.0
6.4
8.7
0.0005
15.0 0.16
Morgan
2.2
0.22
0.33
0.16
48.2
6.9
39.5
92.45
10
10.4
6.0
0.17
0.22
8.1
17.9
7.5 11.9
0.0006
15.0 0.10
Capitols
1.9
0.20
0.28
0.21
38.7
5.4
30.0
38.24
10
10.4
6.9
0.23
0.23
6.2
13.0
6.3
8.7
0.0005
7.5 0.10
Morgan
0.8
0.22
0.43
0.22
8.0
2.9
15.9
3.45
10
10.4
6.0
0.74
0.58
-1.6
2.2
1.2
2.6
0.0041
7.5 0.10
Morgan
1.5
0.22
0.43
0.22
15.9
2.9
15.9
3.45
10
10.4
6.0
0.30
0.31
-1.1
2.2
1.6
2.6
0.0016
7.5 0.10
Morgan
1.9
0.22
0.43
0.22
20.2
2.9
15.9
3.45
10
10.4
6.0
0.20
0.23
1.0
2.2
1.7
2.6
0.0011
7.5 0.16
Morgan
2.4
0.22
0.43
0.22
26.3
3.7
22.0
7.43
10
10.4
6.0
0.15
0.19
-1.3
2.9
2.0
3.6
0.0012
7.5 0.10
Morgan
2.9
0.22
0.43
0.22
30.0
2.9
15.9
3.45
10
10.4
6.0
0.15
0.18
1.0
2.2
2.0
2.6
0.0008
0.10
Morgan
all
0.22
0.43
0.22
72.8
2.9
15.9
3.45
30
10.4
6.0
0.14
0.18
1.1
2.2
2.1
2.6
0.0008
15.0 0.10
Morgan
1.9
0.30
0.44
0.22
38.7
5.4
30.0
38.24
10
10.4
6.0
0.42
0.48
7.5
13.0
6.6
8.7
0.0009
15.0 0.10
Mexico
1.9
0.10
0.16
0.23
38.7
5.4
30.0
38.24
10
10.4
8.1
0.10
0.10
5.6
13.0
6.1
8.7
0.0002
7.5
0.10
Morgan
1.9
0.30
0.59
0.29
20.2
2.9
15.9
3.45
10
10.4
6.0
0.34
0.37
1.1
2.2
1.9
2.6
0.0018
7.5
0.10
Morgan
all
0.30
0.59
0.29
72.8
2.9
15.9
3.45
30
10.4
6.0
0.22
0.28
1.2
2.2
2.3
2.6
0.0012
15.0 0.10
Morgan
1.9
0.40
0.59
0.29
38.7
5.4
30.0
38.24
10
10.4
6.0
0.66
0.75
8.0
13.0
7.1
8.7
0.0014
7.5
0.10
Mexico
1.9
0.10
0.22
0.31
20.2
2.9
15.9
3.45
10
10.4
8.1
0.12
0.13
0.9
2.2
1.7
2.6
0.0007
7.5
0.10
Mexico
2.9
0.10
0.22
0.31
30.0
2.9
15.9
3.45
10
10.4
8.1
0.07
0.09
1.0
2.2
2.0
2.6
0.0004
15.0 0.10
Capitola
1.9
0.30
0.43
0.32
38.7
5.4
30.0
38.24
10
10.4
6.9
0.42
0.44
7.2
13.0
6.9
8.7
0.0009
15.0 0.10
Joshua
1.9
0.27
0.35
0.33
38.7
5.4
30.0
38.24
10
10.4
7.4
0.67
0.68
8.3
13.0
7.5
8.7
0.0014
15.0 0.16
Joshua
2.2
0.27
0.35
0.33
48.2
6.9
39.5
92.45
10
10.4
7.4
0.44
0.51
-11.0 17.9
7.8
11.9
0.0016
7.5 0.10
Morgan
1.9
0.40
0.78
0.39
20.2
2.9
15.9
3.45
10
10.4
6.0
0.51
0.56
1.4
2.2
2.0
2.6
0.0028
15.0 0.10
Capitols
1.9
0.40
0.57
0.43
38.7
5.4
30.0
38.24
10
10.4
6.9
0.62
0.66
7.7
13.0 7.1
8.7
0.0013
15.0 0.16
Capitola
2.2
0.40
0.57
0.43
48.2
6.9
39.5
92.45
10
10.4
6.9
0.48
0.56
-10.1
17.9
7.9 11.9
15.0 0.10
Mexico
1.9
0.20
0.32
0.45
38.7
5.4
30.0
38.24
10
10.4
8.1
0.41
0.43
7.9
13.0
7.3
8.7
0.0009
7.5
0.10
Joshua
1.5
0.27
0.52
0.50
15.9
2.9
15.9
3.45
10
10.4
7.4
0.73
0.69
-1.7
2.2
1.5
2.6
0.0040
7.5
0.10
Joshua
1.9
0.27
0.52
0.50
20.2
2.9
15.9
3.45
10
10.4
7.4
0.52
0.52
-1.5
2.2
2.3
2.6
0.0028
7.5
0.16
Joshua
2.4
0.27
0.52
0.50
26.3
3.7
22.0
7.43
10
10.4
7.4
0.38
0.43
-2.2
2.9
1.9
3.6
0.0031
7.5
0.10
Joshua
2.9
0.27
0.52
0.50
30.0
2.9
15.9
3.45
10
10.4
7.4
0.36
0.39
-1.3
2.2
2.2
2.6
0.0019
7.5
0.10
Joshua
all
0.27
0.52
0.50
72.8
2.9
15.9
3.45
30
10.4
7.4
0.31
0.37
1.3
2.2
2.4
2.6
0.0017
7.5
0.10
Capitols
0.8
0.40
0.76
0.58
8.0
2.9
15.9
3.45
10
10.4
6.9
1.39
1.10
-1.9
2.2
1.2
2.6
0.0076
7.5
0.10
Capitols
1.5
0.40
0.76
0.58
15.9
2.9
15.9
3.45
10
10.4
6.9
0.72
0.72
-1.6
2.2
1.5
2.6
0.0039
7.5
0.10
Capitola
1.9
0.40
0.76
0.58
20.2
2.9
15.9
3.45
10
10.4
6.9
0.57
0.60
-1.3
2.2
2.1
2.6
0.0031
7.5
0.16
Capitola
2.4
0.40
0.76
0.58
26.3
3.7
22.0
7.43
10
10.4
6.9
0.46
0.53
-2.2
2.9
1.7
3.6
0.0038
7.5
0.10
Capitola
2.9
0.40
0.76
0.58
30.0
2.9
15.9
3.45
10
10.4
6.9
0.47
0.52
-1.3
2.2
2.1
2.6
0.0026
7.5
0.10
Capitols
all
0.40
0.76
0.58
72.8
2.9
15.9
3.45
30
10.4
6.9
0.45
0.52
1.3
2.2
2.4
2.6
0.0024
7.5
0.10
Mexico
1.9
0.20
0.44
0.62
20.2
2.9
15.9
3.45
10
10.4
8.1
0.49
0.51
-1.5
2.2
2.1
2.6
0.0027
15.0 0.10
Mexico
1.9
0.30
0.48
0.68
38.7
5.4
30.0
38.24
10
10.4
8.1
0.79
0.87
-9.0
13.0
7.4
8.7
0.0017
0.10
Mexico
1.9
0.35
0.56
0.79
38.7
5.4
30.0
38.24
10
10.4
8.1
0.96
1.08
-10.4
13.0
7.4
8.7
0.0020
15.0 0.16
Mexico
2.2
0.35
0.56
0.79
48.2
6.9
39.5
92.45
10
10.4
8.1
0.70
0.84
-15.5
17.9
8.2 11.9
15.0 0.10
Mexico
1.9
0.40
0.64
0.91
38.7
5.4
30.0
38.24
10
10.4
8.1
1.14
1.30
-11.6
13.0
7.4
81
0.0024
7.5
0.10
Mexico
all
0.30
0.66
0.94
72.8
2.9
15.9
3.45
30
10.4
8.1
0.50
0.67
-2.2
2.2
2.3
2.6
0.0028
7.5
0.10
Mexico
0.8
0.35
0.77
1.09
8.0
2.9
15.9
3.45
10
10.4
8.1
3.60
2.09
-1.4
2.2
1.5
2.6
0.0197
7.5
0.10
Mexico
1.5
0.35
0.77
1.09
15.9
2.9
15.9
3.45
10
10.4
8.1
1.49
1.37
-2.7
2.2
1.5
2.6
0.0081
7.5
0.10
Mexico
1.9
0.35
0.77
1.09
20.2
2.9
15.9
3.45
10
10.4
8.1
1.07
1.12
-2.5
2.2
2.1
2.6
0.0059
7.5
0.16
Mexico
2.4
0.35
0.77
1.09
26.3
3.7
22.0
7.43
10
10.4
8.1
0.87
0.96
-3.3
2.9
1.8
3.6
0.0071
7.5
0.10
Mexico
2.9
0.35
0.77
1.09
30.0
2.9
15.9
3.45
10
10.4
8.1
0.83
0.95
-2.7
2.2
2.4
2.6
0.0045
7.5
0.10
Mexico
all
0.35
0.77
1.09
72.8
2.9
15.9
3.45
30
10.4
8.1
0.61
0.80
-2.5
2.2
2.4
2.6
0.0033
7.5
7.5
15.0
0.0017
0.0025