Guide to Durable
Concrete
ACI 201.2R-16
Reported by ACI Committee 201
First Printing
November 2016
ISBN: 978-1-945487-39-2
Guide to Durable Concrete
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ACI 201.2R-16
Guide to Durable Concrete
Reported by ACI Committee 201
Thomas J. Van Dam, Chair
Reza Ahrabli
James M. Aldred
Jon B. Ardahl
Mohamed Bassuoni
Bruce Blair
Andrew J. Boyd
Paul W. Brown
Ramon L. Carrasquillo
Rachel J. Detwiler
Jonathan E. Dongell
R. Douglas Hooton, Secretary
Thano Drimalas
Kevin J. Folliard
Harvey H. Haynes
Jason H. Ideker
Francis Innis
Donald J. Janssen
Roy H. Keck
Mohammad S. Khan
Kimberly E. Kurtis
Michael L. Leming
Tyler Ley
Darmawan Ludirdja
Mohamad Nagi
Robert E. Neal
Charles K. Nmai
Karthik H. Obla
Robert C. O’Neill
Kyle Austin Riding
David A. Rothstein
Hannah C. Schell
Lawrence L. Sutter
David G. Tepke
Michael D. A. Thomas
Paul J. Tikalsky
David Trejo
Orville R. Werner II
Terry J. Willems
Michelle L. Wilson
Consulting Members
W. Barry Butler
Bernard Erlin
Odd E. Gjorv*
William G. Hime
Charles J. Hookham
Alexander M. Leshchinsky
Stella Lucie Marusin
Howard H. Newlon Jr.
Mauro J. Scali
George V. Teodoru
Niels Thaulow
J. Derle Thorpe
Claude B. Trusty Jr.
*
Deceased.
This guide describes specific types of concrete deterioration. Each
chapter contains a discussion of the mechanisms involved and the
recommended requirements for individual components of concrete,
quality considerations for concrete mixtures, construction procedures, and influences of the exposure environment, which are all
important considerations to ensure concrete durability.
This guide was developed for conventional concrete but is generally applicable to specialty concretes; however, specialty concretes,
such as roller-compacted or pervious concrete, may have unique
durability-related issues that deserve further attention that are not
addressed herein.
CONTENTS
CHAPTER 1—INTRODUCTION AND SCOPE, p. 2
1.1—Introduction, p. 2
1.2—Scope, p. 3
CHAPTER 2—DEFINITIONS, p. 3
2.1—Definitions, p. 3
CHAPTER 3—MASS TRANSPORT, p. 3
3.1—Introduction, p. 3
3.2—Transport processes in nonreactive porous media, p. 4
3.3—Factors affecting mass transport in concrete, p. 5
3.4—Measurement of transport properties, p. 8
3.5—Obtaining durable concrete, p. 10
Keywords: abrasion resistance; alkali-aggregate reaction; chemical attack;
curing; deterioration; durability; freezing and thawing; physical salt attack,
sulfate attack.
ACI Committee Reports, Guides, and Commentaries are
intended for guidance in planning, designing, executing, and
inspecting construction. This document is intended for the use
of individuals who are competent to evaluate the significance
and limitations of its content and recommendations and who
will accept responsibility for the application of the material it
contains. The American Concrete Institute disclaims any and
all responsibility for the stated principles. The Institute shall
not be liable for any loss or damage arising therefrom.
Reference to this document shall not be made in contract
documents. If items found in this document are desired by
the Architect/Engineer to be a part of the contract documents,
they shall be restated in mandatory language for incorporation
by the Architect/Engineer.
CHAPTER 4—FREEZING AND THAWING OF
CONCRETE, p. 10
4.1—Introduction, p. 10
ACI 201.2R-16 supersedes ACI 201.2R-08 and was adopted and published
November 2016
Copyright © 2016, American Concrete Institute.
All rights reserved including rights of reproduction and use in any form or by
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reproduction or for use in any knowledge or retrieval system or device, unless
permission in writing is obtained from the copyright proprietors.
1
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
4.2—Frost attack of concrete made with durable aggregates, p. 11
4.3—Frost attack of concrete made with nondurable
aggregates, p. 17
CHAPTER 5—ALKALI-AGGREGATE REACTION,
p. 19
5.1—Introduction, p. 19
5.2—Types of reactions, p. 19
5.3—Evaluating aggregates for potential alkali-aggregate
reactivity, p. 22
5.4—Preventive measures, p. 25
5.5—Tests for evaluating preventive measures, p. 28
5.6—Protocols for minimizing the risk of alkali-aggregate
reactivity, p. 29
CHAPTER 6—SULFATE ATTACK, p. 30
6.1—External sulfate attack, p. 30
6.2—Internal sulfate attack, p. 36
6.3—Seawater and brine exposure, p. 37
CHAPTER 7—CHEMICAL ATTACK, p. 39
7.1—General, p. 39
7.2—Seawater, p. 39
7.3—Acid attack, p. 41
7.4—Fresh water, p. 42
7.5—Carbonation, p. 42
7.6—Industrial chemicals, p. 43
7.7—Deicing and anti-icing chemicals, p. 44
7.8—Environmental structures, p. 45
CHAPTER 8—PHYSICAL SALT ATTACK, p. 45
8.1—Introduction, p. 45
8.2—Occurrence, p. 46
8.3—Background, p. 47
8.4—Mechanism, p. 47
8.5—Recommendations, p. 48
CHAPTER 9—CORROSION OF METALS
AND DEGRADATION OF OTHER MATERIALS
EMBEDDED IN CONCRETE, p. 48
9.1—Introduction, p. 48
9.2—General principles of corrosion initiation in concrete,
p. 48
9.3—Propagation of corrosion, p. 49
9.4—Corrosion-related properties of concreting materials,
p. 49
9.5—Mitigating corrosion, p. 50
9.6—Corrosion of prestressed steel reinforcement, p. 53
9.7—Degradation of materials other than steel, p. 53
9.8—Summary, p. 54
CHAPTER 10—ABRASION, p. 54
10.1—Introduction, p. 54
10.2—Testing concrete for resistance to abrasion, p. 55
10.3—Factors affecting abrasion resistance of concrete,
p. 55
10.4—Recommendations for obtaining abrasion-resistant
concrete surfaces, p. 57
10.5—Studded tire and tire chain wear on concrete, p. 58
10.6—Skid resistance of pavements, p. 58
10.7—Erosion, p. 59
CHAPTER 11—SUMMARY, p. 60
CHAPTER 12—REFERENCES, p. 60
Authored documents, p. 62
CHAPTER 1—INTRODUCTION AND SCOPE
1.1—Introduction
Concrete is the most widely used construction material in
the world. The design, detailing, and execution of concrete
to resist weathering action, chemical attack, abrasion, and
other processes of deterioration over its intended service life
will determine its durability. Durable concrete will retain its
original form, quality, and serviceability when exposed to its
environment. Properly designed, proportioned, transported,
placed, finished, and cured concrete is capable of providing
decades of service with little or no maintenance. Yet certain
conditions or environments exist that can lead to concrete
deterioration. Deterioration mechanisms are either chemical or physical in nature and may originate from within the
concrete, or may be the result of the external environmental
exposure. Chemical and physical attacking mechanisms often
work synergistically. Depending on the nature of the attack,
distress may be concentrated in the paste, aggregate, or reinforcing components of the concrete, or a combination thereof.
The various factors influencing durability and a particular
mechanism of deterioration should be considered in the
context of the environmental exposure of the concrete. In
addition, consideration should be given to the microclimate
to which the specific structural element is to be exposed.
The type and severity of deterioration of a given structure
may be affected by its proximity to sources of deleterious
agents or agents that facilitate distress, exposure to wind,
precipitation, or temperature. For instance, exterior girders
in a bridge structure may be exposed to a more aggressive
environment than interior girders.
The concept of service life is increasingly used for the
design of new structures. To produce concrete suitable for a
particular application, required service life, design requirements, and expected exposure environments, both macro
and micro, should be determined before defining the necessary materials and mixture proportions.
The use of good materials and proper mixture proportioning
will not, by itself, ensure durable concrete. Appropriate placement practices and workmanship are essential to the production of durable concrete. Fresh concrete can be consolidated
and molded to the shape desired to serve its intended purpose.
During this stage, a number of properties significantly influencing the durability of the hardened concrete are established.
Pore structure development, air-void system formation, material uniformity, and potential for cracking are established
at early ages and are important to the ultimate durability of
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
concrete. As such, durable concrete requires the application
of good quality control during construction. Inspection and
testing by trained and certified personnel can help ensure the
use of durable mixtures and proper practices.
1.2—Scope
This guide discusses the important mechanisms of concrete
deterioration and gives recommendations on how to mitigate
or minimize such damage. This guide also addresses durability by first discussing the importance of mass transport
and then addressing specific modes of attack in separate
chapters. These include freezing and thawing, alkali-aggregate reaction (AAR), sulfate attack, aggressive chemical
attack, physical salt attack, corrosion of metals and other
embedded materials, abrasion, or a combination of these.
Fire resistance of concrete and cracking are not addressed
directly. Fire resistance is covered in ACI 216.1 and cracking
is covered in ACI 224R and ACI 224.1R. While cracking
does impact the durability of concrete in severe exposures,
the different causes of cracking and their specific impacts are
not discussed. Cracking is only mentioned in general terms
regarding its impact on fluid ingress.
CHAPTER 2—DEFINITIONS
2.1—Definitions
ACI provides a comprehensive list of definitions through
an online resource, “ACI Concrete Terminology,” https://
www.concrete.org/store/productdetail.aspx?ItemID=CT16.
Definitions provided herein complement that source.
advective transport––transfer of heat or matter via the
bulk motion of a fluid.
alkali loading (or content)––total amount of equivalent
alkalis (Na2Oe) in a concrete mixture expressed as mass per
volume.
calcium sulfoaluminate cement––product obtained
by pulverizing clinker containing mainly ye′elimite
[Ca4(AlO2)6SO4] that is often used in expansive cements and
ultra-high-early-strength cements.
diffusion––movement of species, such as ions, gas, or
vapor, from an area of higher concentration to an area of lower
concentration, independent of the bulk motion of a fluid.
electrical migration––transport of electrons or ions due
to an electric potential gradient.
ice lens––layer of ice, generally parallel to the exposed
surface of the concrete, that can produce internal damage
and also lead to scaling or delamination.
leaching––dissolution and removal of soluble components such as calcium hydroxide from concrete.
permeability—the ability of a given concrete to permit
liquids or gases to pass through.
permeation––flow of a liquid, gas, or vapor within a solid
under the action of a pressure gradient.
physical salt attack––mechanism in which concrete or
mortar is damaged as a result of salt crystallization pressure.
reactive silica––form of silica, often amorphous or
crypto-crystalline, that dissolves when in contact with
3
concrete pore solution having a sufficiently high concentration of hydroxyl ions.
salt weathering––form of deterioration most commonly
observed in arid climates where exposure to soluble salts
and cyclic variations in temperature and relative humidity
can lead to salt crystallization.
thaumasite––silicate mineral, colorless to white prismatic hexagonal crystals typically as acicular radiating
groups, with the chemical formula {[Ca3Si(OH)6·12(H2O)]
(SO4)(CO3)}.
CHAPTER 3—MASS TRANSPORT
3.1—Introduction
Concrete is a multiphase porous medium consisting of a
multiscale porous cement paste matrix with aggregate inclusions. Liquid and gas may be present in any pores and microcracks. As such, it is susceptible to the ingress and movement of substances (fluids or ions) from its environment
within and through its pore system. This chapter discusses
the transport of gases, liquids, and ions in solution through
concrete (Lichtner et al. 1996; Baer 1988; Hearn et al. 2006;
Hall and Hoff 2012). Methods for improving the durability
of concrete and some of the common test methods used to
measure the transport properties, along with their advantages
and limitations with regard to assessing concrete durability,
are also discussed. It is recognized that the rate of ingress
of fluids and ions will increase by the presence of cracks.
However, the specific influences of different types of cracks
and crack widths are not discussed herein.
The ingress of gases, liquids, or ions in solution through
concrete may initiate chemical processes, physical processes,
or both, that affect the durability of the concrete under a given
set of service conditions. Water itself may be harmful because
of its ability to leach calcium hydroxide (CH) from the hardened cement paste and because of osmotic pressures generated as water flows to sites of higher alkalinity (Powers et al.
1954; Powers 1975; Helmuth 1960b,c). In addition, water
may also be acidic or carry harmful dissolved chemicals, such
as chlorides or sulfates, into the concrete. The ingress of gases
such as oxygen and carbon dioxide through the concrete pores
can contribute to the corrosion of steel reinforcement.
Different substances may interact with components of the
concrete in different ways; therefore, transport of a substance
through concrete is unique to that substance. For example,
water can hydrate previously unhydrated cement particles or
leach calcium. Chloride ions may be bound by the hydration
products of cement or supplementary cementitious materials
(SCMs). The size of the molecules or ions that are transported through the concrete, viscosity of the fluid, valence of
the ions, and other ionic species present also affect the transport properties. Thus, permeability and diffusivity must be
expressed in terms of the substance that is migrating through
the concrete. In general, concrete with transport properties
that limit the rate of ingress of external agents is not immune
to chemical deterioration, but the effects are mainly near the
exposed surfaces, so the concrete tends to be more durable.
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
3.2—Transport processes in nonreactive porous
media
This section provides a brief overview of the transport
processes of fluids (gases and liquids) and ions in solution
within a nonreactive porous medium. This is a simplifying
assumption because concrete changes chemically and physically with time in response to its environment. Physical
changes, chemical changes, or both, in the internal structure
of the porous medium resulting from interactions with the
migrating fluids or ions are not discussed herein. The transport pathways described here include:
a) Transport by permeation
b) Advective transport
c) Hydrodynamic dispersion
d) Diffusion within the pores
e) Transport due to electrostatic interactions or electrical
migration
Martín-Pérez et al. (2001) modeled transport related to
corrosion of reinforcement in concrete based on chloride
transport, moisture diffusion, heat transfer, and oxygen
transport using a two-dimensional finite element model.
They used a modified version of Fick’s second law.
Johannesson (2003) developed a theoretical model for
diffusion of different types of ions in concrete pore solution.
The model incorporates diffusion caused by concentration
gradients of ions (for example, due to drying), internal electrical potential, convection, effects of changes in moisture
content, and mass exchange of ions between solution and
solid hydration phases.
Chung and Consolazio (2005) developed a finite difference
model to simulate heat and mass transport in rapid heating
conditions, such as fires in reinforced concrete structures. The
model accounts for the interference between liquid and gas
phases, slip-flow effects in steam flow, and the interference of
steel reinforcement in moisture movement in concrete.
3.2.1 Transport by permeation—Permeation is the flow of
a fluid under the action of a pressure gradient. Permeability
is the property that characterizes the ease with which fluid
passes through a porous material under a pressure gradient.
For a steady laminar flow through a saturated porous
medium, the fluid flow is related to the hydraulic pressure
gradient according to Darcy’s law.
dq/dt = K1A∆h/l
(3.2.1)
where dq/dt is the flow (expressed as a rate); K1 is the permeability coefficient; A is the cross-sectional area; ∆h is the
hydraulic head; and l is the thickness of the specimen. The
permeability coefficient K1 is the rate of discharge of water
under laminar flow conditions through a unit cross-sectional
area of a porous medium under a unit hydraulic gradient
and standard temperature conditions. Darcy’s law indicates
that for a given cross-sectional area and permeability coefficient, the flow is proportional to the hydraulic gradient ∆h/l.
Under service conditions, flow is three-dimensional and the
concrete may not be saturated. In concrete, the permeability
coefficient may change with increased hydration, cracking,
Fig. 3.2.1‒Graphical representation of a simple percolation
theory model used to model the permeability of concrete.
or changes in the pore structure due to various physical and
chemical processes.
Permeability coefficients of plastic portland cement pastes
of 0.5 water-cement ratio (w/c), calculated from measurements
of bleeding, ranged from 5 to 8 × 10–7 m/s for four cements
with different chemical composition but the same specific
surface (180 m2/kg by the Wagner turbidimeter). The permeability coefficient of mature paste (for example, at greater than
28 days of age) is between 1 millionth and one 10 millionth of
that of fresh paste. It ranges from 1 × 10–15 to 1.2 × 10–12 m/s
for w/c ranging from 0.3 to 0.7 (Powers et al. 1954).
To understand the effects of microstructure on the permeability of concrete, Bentz et al. (1999) at the National Institute of Standards and Technology (NIST) used percolation
theory. One useful application of percolation theory is the
examination of the time needed for a material to progress
through a complex maze (Stauffer and Aharony 1992). This
maze consists of areas that can allow free movement, as
well as areas that impede the transport of a fluid to different
degrees. These models can be made in two and three dimensions and can include the effects of cracks. Lu et al. (2012)
were able to use a three-dimensional version of the NIST
model to predict chloride ingress into cracked concrete.
Work has also been done by NIST that models changes
in the properties of these systems with time. This allows the
change in the microstructure of the concrete to be examined
with time and observe the effects on the ability of a fluid to
move through the system.
A graphical representation of a simple percolation
theory model has been used by Bentz (2000) to model
the transport of a fluid through concrete in two dimensions, as shown in Fig. 3.2.1. Very dense areas are used
for aggregates. Moderately dense material is used for the
cement paste and low-density material is used to model
the interfacial transition zone (3.3.4).
3.2.2 Advective transport—Advective transport refers to
the movement of molecules and ions with the bulk solution
flow. This transport process is related directly to the velocity
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
5
Fig. 3.3.1a–‒Relative sizes of different types of pores and other microstructural features
(adapted from Mehta [1986]).
of fluid flow and concentration of ions in solution. Depending
on exposure condition and degree of saturation, the velocity
of fluid flow depends on the sorptivity or permeability of
concrete, viscosity of the fluid phase, and pressure head of
the permeating fluid or rate of evaporation from the exposed
surface in the case of wick action (Buenfeld et al. 1995).
3.2.3 Hydrodynamic dispersion—Dispersion is the spreading
of the ion concentration during advective transport due to
variations in the pore fluid velocity. These variations can be
a result of the tortuosity of the pore structure, the connectivity of the pore network, or variation in the fluid properties.
3.2.4 Diffusion within pores—Diffusion refers to the transport mechanism whereby ions or gases migrate from areas of
higher concentration to areas of lower concentration. For the
idealized one-dimensional case, Fick’s second law describes
the non-steady-state diffusion of ions within the pores
dc/dt = De · d2c/dx2
(3.2.4)
where c is the concentration of the ion at distance x from the
surface after time t, and De is the effective diffusion coefficient or effective diffusivity. The effective diffusivity is a
function of the porosity and tortuosity of the porous medium
and the molecular diffusivity of the ion of concern. Many
factors affect diffusion of ions in concrete. Based on measurements obtained under controlled conditions in the laboratory,
diffusion coefficients increase with temperature and waterto-cementitious materials ratio (w/cm) and decrease with
increasing degree of hydration. Because concrete pore solutions have high ionic strength, electrical charge effects can
be significant. Diffusion coefficients can also vary with the
species of other ions present in solution. For these and other
reasons discussed in 3.4.1.2, the values obtained experimentally using Fick’s second law are generally termed “apparent
diffusion coefficients”. Diffusion coefficients for Na+ in
concrete are on the order of 10–11 to 10–13 m2/s, and for Cl–, on
the order of 10–11 to 10–12 m2/s (Taylor 1997).
3.2.5 Transport due to electrostatic interactions or electrical migration—Migration refers to the transport mechanism due to the charged nature of ions and is the result of
the potential difference across the specimen. The electrical
coupling between ions in concentrated solutions was demon-
strated by Snyder (2001) and Snyder and Marchand (2001).
Electrical migration occurs when an external electric field
such as in ASTM C1202 (or AASHTO T277) is applied to
the medium (Buenfeld et al. 1998). The migration flux Ji of
ion i is given by
J i = − Di Ci
zi F ∂ ϕ
RT ∂x
(3.2.5)
where ∂φ/∂x is the potential difference; Zi is the charge of the
ion; F is the Faraday constant; T is the temperature; Di is the
ion diffusivity; R is the gas constant; and Ci is the concentration of the ion in solution.
Further information on ionic transport in concrete can
be found in McGrath and Hooton (1996) and for the more
complex case of multi-species transport in Truc et al. (2000)
and Samson et al. (1999).
3.3—Factors affecting mass transport in concrete
3.3.1 Porosity and pore size distribution—Porosity is
defined as the volume of voids as a fraction that is usually
expressed as a percent of the total volume
porosity (%) = (volume of voids/total volume) × 100% (3.3.1)
Figure 3.3.1a shows the size ranges for the various types
of pores in concrete. Pores in concrete range in size from
nanometers to millimeters. The capillary pores, also ranging
in size from tens of nanometers to millimeters, have the
most significant effect on the transport properties. Transport properties, however, depend more on the connectivity
of the pores than on either the porosity or size of the pores.
Figure 3.3.1b shows two hypothetical porous materials with
approximately the same porosity. In one material, the pores
are discontinuous, as would be the case with entrained air
bubbles, whereas in the other the pores are continuous. The
latter material would allow for more rapid rates of transport
than the former.
3.3.2 Water-cement ratio (w/c)—The initial porosity of a
cement paste is determined by the w/c. As cement hydrates,
hydration products fill some of the void space formerly occupied by water. With time, this process results in a continued
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
Fig. 3.3.1b–‒Porosity and permeability are related but
distinct. The two hypothetical materials shown have approximately the same porosity (total volume of pores), but different
permeabilities. Discrete pores, such as those resulting from
air entrainment, have almost no effect on permeability, but
interconnected pores increase permeability.
decrease in the porosity of the cement paste. Figure 3.3.2a
(Mehta 1986) illustrates the relationship among w/c, degree
of hydration, and capillary porosity. For a w/c of 0.45, the
degree of hydration must reach approximately 70 percent to
bring the porosity down to 30 percent. For a w/c of 0.60, the
degree of hydration must reach approximately 100 percent
to reach the same porosity. The degree of hydration that
could be expected for good curing conditions—for example,
moist curing for 5 to 7 days—would range between 70 and
80 percent, depending on cement chemistry and fineness,
and on hydration temperatures; a 100 percent degree of
hydration is not a practical possibility. Figure 3.3.2b shows
the relationship between porosity and permeability. Above
a porosity of approximately 30 percent, the coefficient of
permeability increases sharply.
Powers (1962b) calculated that for cement paste with a
w/c of 0.38, all of the capillary pore space was just filled by
maximum density gel when all of the cement was hydrated.
Sealed, fully hydrated cement pastes made at w/c above
0.38 have remaining capillary pore space equal to the excess
above 0.38. Partially hydrated mixtures have proportionately less gel and more capillary space. Powers et al. (1959)
calculated the time required for capillary pores to become
discontinuous with increasing hydration of the cement, as
shown in Table 3.3.2. It is notable that mixtures with a w/c
greater than 0.7 will always have continuous pores. Even for
w/c of 0.40 to 0.45, extended moist curing or other favorable
curing conditions are necessary to achieve the desired low
permeability.
3.3.3 Curing temperature
3.3.3.1 At normal temperatures—Soon after mixing
cement with water, a gel layer forms on the surfaces of the
cement grains (Taylor 1997). Between 3 and 24 hours after
mixing cement with water, approximately 30 percent of the
cement reacts. Rapid formation of calcium silicate hydrate
(C-S-H) and CH is accompanied by significant evolution of
heat. The CH forms massive crystals in the originally waterfilled space. The C-S-H forms a thickening layer around
Fig. 3.3.2a‒–Water-cement ratio versus capillary porosity
for cement paste at different degrees of hydration (Mehta
1986) based on equations developed by Powers and Brownyard (1948).
Fig. 3.3.2b‒–Both compressive strength and permeability
are related to the capillary porosity of the cement paste
(adapted from Powers [1958]).
Table 3.3.2–Approximate age required to
produce maturity at which capillaries become
discontinuous for concrete continuously moistcured (Powers et al. 1959)
w/c by mass
Time required
0.40
3 days
0.45
7 days
0.50
14 days
0.60
6 months
0.70
1 year
Over 0.70
Impossible
the cement grains. As the shells grow outward, they begin
to coalesce at about 12 hours, a time coinciding with the
maximum rate of heat evolution (Fig. 3.3.3.1) and corresponding approximately to completion of setting (Taylor
1997). In Fig. 3.3.3.1, the first heat peak is associated
with the initial hydrolysis of the C3S and the hydration of
the C3A. The acceleration period begins with the renewed
evolution of heat at the beginning of the second peak as the
initial hydration products of the C3S begin to form. Initial
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
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Fig. 3.3.3.1––Heat evolution of Type I/II portland cement
paste as measured by conduction calorimetry (Image courtesy of E. Shkolnik.).
set coincides with the beginning of the acceleration period.
Final set takes place just before the maximum point of the
second peak. The shoulder of the second peak is associated with the renewed formation of ettringite (Taylor 1997).
Further hydration of the cement continues at a much slower
rate, asymptotically approaching 100 percent (Mindess and
Young 1981). After the spaces between the hydration shells
and the cement grains fill with hydration products, further
hydration is slow (Taylor 1997). Capillary pores remaining
in mature cement paste increase in size with w/c and have
diameters of 10 nm and higher (Mindess and Young 1981).
3.3.3.2 At high or low temperatures—Like most chemical reactions, cement hydration is faster with increasing
temperature. Verbeck and Helmuth (1969) postulated that at
elevated temperatures, cement hydration products would not
have time to diffuse any significant distance from the cement
grain, thus forming relatively dense hydration shells around
the cement grains. A consequence of the uneven distribution
of the solid phases is a coarser pore structure. Goto and Roy
(1981) found that the total porosities of pastes hydrated at
140°F (60°C) were greater than those of pastes hydrated at
81°F (27°C). For cement pastes hydrated at low temperatures, on the order of 50°F (10°C), the hydration products
are more evenly distributed and the pores fine and discontinuous. For cement pastes hydrated at elevated temperatures, pores are coarser and more interconnected (Kjellsen
et al. 1991). Cement pastes containing fly ash, slag cement,
or both, are less sensitive to the effects of elevated temperatures, as discussed in 3.3.5.
3.3.4 Aggregates—Aggregates generally have fundamentally different transport properties from those of cement
paste. For example, the permeability of granite is typically
two to three orders of magnitude lower than that of cement
paste. The presence of the aggregate in a cement-paste
matrix creates an inhomogeneity in the structure of hardened
concrete known as the interfacial transition zone between
the cement paste and aggregate. Mehta (1986) reported that,
compared to the bulk cement paste, the interfacial transition
zone has a higher void content, higher contents of CH and
ettringite, reduced content of C-S-H, and larger crystals of
Fig. 3.3.4‒–Representation of transition zone at paste/
aggregate interface in concrete, showing more coarsely
crystalline and porous microstructure than in interzonal
mass (Mehta 1986).
CH strongly oriented parallel to the aggregate surface (Fig.
3.3.4). Factors contributing to the anomalous nature of the
interfacial transition zone include bleeding, which creates
pockets of water-filled space beneath aggregate particles;
less efficient packing of particles of cementitious materials
in the vicinity of a surface, which is called the wall effect;
and the one-sided growth effect of dissolved cementitious
materials and hydration products diffusing in from the bulk
cement paste, but not from the aggregate (Bentz et al. 1995).
As the cementitious materials hydrate, the interfacial transition zone fills preferentially with CH and ettringite. Because
of the relatively open space, the crystals can grow large.
Thus, in most concrete, the interfacial transition zone is the
weakest link in terms of mechanical behavior and transport
properties. For the latter, the interfacial transition zone can
serve as a relatively open channel for fluids and ions, and
the CH is vulnerable to leaching and acid attack. For a given
w/c and degree of hydration, water permeability of concrete
made with low-permeability aggregates is approximately
one to two orders of magnitude lower than that of cement
paste due to the interfacial transition zone between aggregate and cement paste (Mehta 1986). It was found that the
diffusivity of chloride in the interfacial transition zone is 10
times greater than that in bulk cement paste (Delagrave et al.
1997). The connectivity of pores in the interfacial transition
zone may be high, leading to significantly greater rates of
transport for some concretes than might be predicted from
their mixture proportions. For mixtures with high coarse
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
aggregate contents, such as paving concretes, flow through
the interfacial transition zone can completely dominate
moisture movement in the concrete over a wide range of w/c
(Janssen and Snyder 1994).
3.3.5 Supplementary cementitious materials—The use of
SCMs can significantly reduce the permeability and diffusivity of concrete. These materials may not reduce the total
porosity to any great extent, but instead act to refine and
subdivide the pores so that they become less continuous.
Some benefits are obtained by the improvements in workability, particularly the reduction in bleeding, afforded by
these materials. The use of fly ash can also reduce the water
demand of some concrete mixtures, allowing reductions in
water-to-cementitious materials ratio (w/cm) while maintaining equivalent workability. The lower specific gravity
of SCMs relative to cement results in an increased volume
of solids for a given w/cm with increasing dosage. The
greatest benefits from the standpoint of durability, however,
derive from the pozzolanic reaction associated with many
SCMs. In this reaction, CH from the hydration of the cement
reacts with noncrystalline silica in SCMs and water to form
C-S-H. Because C-S-H has a greater volume than the CH
and pozzolan from which it forms, the pozzolanic reaction
results in a finer system of capillary pores.
Slag cement and alumina-bearing pozzolans have an additional benefit as the hydration products are highly effective
in binding chloride ions, preventing further penetration into
the concrete (Thomas et al. 2012). This effect is particularly
important in applications such as parking garages, bridge
decks, and marine construction, where the reinforcing steel
is vulnerable to chloride-induced corrosion.
The indiscriminate use of SCMs is not necessarily beneficial. Their pozzolanic and hydraulic reactions take time,
partly because the cement must hydrate sufficiently to
produce CH to participate in the reaction, and partly because
SCMs may vary significantly in terms of kinetics or rate
of reactivity. The engineer must consider the properties of
the concrete at early ages. Because the pozzolanic reactions typically proceed more slowly than the hydration of
cement, extended moist curing is necessary to achieve the
best results.
Supplementary cementitious materials can often mitigate
the deleterious effects of elevated-temperature curing. Cao
and Detwiler (1995) found that both silica fume and slag
were effective in refining the pore structure. Campbell and
Detwiler (1993) tested a series of steam-cured concretes in
which the proportions of the various cementitious materials
were varied. They found that the total charge passed in 6
hours using AASHTO T277 varied by two orders of magnitude, with CSA Type 10 (now designated Type GU) portland
cement alone performing the worst, and optimized blends of
cement, slag, and silica fume performing the best. They did
not use fly ash in their study.
Bentz et al. (1995) showed that silica fume particles both
reduce the initial thickness of the interfacial transition zone
and react to convert CH to C-S-H. Fly ash has a similar,
but less-pronounced, effect due to its larger particle size and
lower pozzolanic activity.
3.4—Measurement of transport properties
Measurement of the transport properties of concrete is
complicated by the interactions between the concrete and
the substance that is moving through it; the changing properties of concrete with time; and the sensitivity of transport
properties to variations in moisture, temperature, and other
conditions. Because many tests accelerate the transport
mechanism of the fluid or ion in question, they may induce
different or additional mechanisms of transport than what
would occur in service. They often make it impossible to
achieve the steady-state conditions that form the basis of
the various equations used to describe mass transport, and
may invalidate the assumption of laminar flow used in many
calculations (Eq. (3.2.1)). Furthermore, laboratory tests are
often conducted under highly controlled conditions that may
not accurately reflect actual service conditions. When used
judiciously, however, the tests described in the following
may be helpful in comparing the suitability of different
concrete mixtures for a given exposure condition, or for
quality assurance/quality control purposes during construction (Puerto Rico DOT SP934). The following sections
discuss available ASTM tests used to characterize transport
properties, along with commonly used variations.
3.4.1 Ions
3.4.1.1 Coulomb test (ASTM C1202/AASHTO T277)—
Standard tests of ion transport focus on the penetration of
chloride ions into concrete. The most common test used for
this purpose is ASTM C1202 (or AASHTO T277), in which
a cylindrical specimen with a nominal diameter of 4 in.
(100 mm) and a length of 2 in. (50 mm) is vacuum saturated with water before being placed in a test cell. The cell
contains a 3 percent solution of sodium chloride on one side
and a 0.3 N solution of sodium hydroxide on the other side.
An electrical potential of 60 V dc is applied for 6 hours. The
total charge passed during the test period is an indirect indication of the chloride ion penetrability of the concrete.
Essentially, ASTM C1202 (or AASHTO T277) uses the
electrical conductivity of the concrete as a rapid index test
or surrogate for diffusivity. The main objections to the use of
this test method stem from the indirect nature of the measurement. While ion diffusion depends primarily on the microstructure and chemical binding capacity of the matrix, electrical conductivity depends on both microstructure and pore
solution chemistry (Buenfeld and Newman 1987). Different
proportions of SCMs can profoundly affect the pore solution chemistry. For example, Page and Vennesland (1983)
found that 10 percent silica fume, by substitution, reduced
the concentrations of Na+, K+, and Ca2+ by approximately
50 percent and that of OH– by approximately 75 percent in
the pore solution. Detwiler and Fapohunda (1993) compared
the results of AASHTO T277 to those of a direct measure of
chloride ion migration for portland-cement concretes with
and without slag cement, and found that AASHTO T277
unduly favored the concretes containing SCMs. They attributed the differences between the two sets of test results to
differences in electrical conductivity due to differences in
pore solution chemistry.
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Corrosion inhibitors such as calcium nitrite add ions to
the pore solution that increase the pore fluid conductivity
and the charge passed. There is some evidence that calcium
nitrite increases the penetrability of the concrete matrix.
That is, increased coulomb values may be due to microstructural effects as well as changes in the pore solution
chemistry (Ann et al. 2006; Reou and Ann 2008). If ASTM
C1202 (or AASHTO T277) is used in qualifying concrete
mixtures for use in construction, typically the concrete will
be tested without the corrosion inhibitor to qualify it, and
then with the corrosion inhibitor to provide a baseline value
for quality-control purposes. The latter measurement will be
considerably higher.
The relationship between electrical conductivity and
diffusion may also vary with the mechanism and type of
diffusion. Diffusion through voids and cracks differs from
bulk diffusion; thus, the presence of flaws in the concrete
can significantly affect the results of the test. In addition,
the measurement is taken before steady-state conditions are
reached (Zhang and Gjørv 1991).
Low-quality concrete is difficult to evaluate properly
because the temperature of the specimens rises when current
is applied, increasing the rate of diffusion. McGrath and
Hooton (1999) proposed a modification to reduce the test
period to 30 minutes to eliminate this problem. High-quality
concretes may be difficult to distinguish from one another
because the total charge passed is so low, and because the
test results are variable (Hooton 1989).
ASTM C1202 (or AASHTO T277) has been criticized for
many of the previously-cited interferences (Andrade 1993;
Feldman et al. 1994; Streicher and Alexander 1994; Shi 2004).
Despite its limitations, ASTM C1202 (or AASHTO T277)
is rapid, convenient, and according to Hooton (1989), whatever property it is measuring probably is coincident with
permeability. If properly interpreted, it can be used effectively for quality control in construction (Bognacki et al
2010), although it should not be used to make fine distinctions among concretes and cannot be used to compare
concretes made with different materials or mixture proportions, or both (Shi 2004).
3.4.1.2 Ponding (ASTM C1543; AASHTO T259) and
bulk diffusion (ASTM C1556)—Another standard test that
is sometimes used is ASTM C1543 (or AASHTO T259),
which involves ponding three concrete slabs at least 3 in. (75
mm) thick and a surface area of 46 in.2 (0.030 m2) with a 3
percent sodium chloride solution for 90 days. The sides of the
slab are sealed and the bottom exposed to a drying environment at 50 percent relative humidity. If desired, the exposure
period can be extended to 6 months or 1 year. At the end of
the exposure period, the excess solution and salt buildup are
removed. Half-inch (12 mm) thick samples of the concrete
can be taken at two or three depths and analyzed for chloride
ion content, which is compared to a baseline value determined
on a companion concrete specimen not exposed to external
chlorides. Alternatively, the concrete slab can be sampled and
tested according to ASTM C1556, in which a core from the
slab is milled or sliced to obtain samples at eight depths for the
purpose of determining the apparent chloride diffusion coeffi-
9
cient using Fick’s second law. This test can also be conducted
using a different salt in the ponding solution. Note that the
type of cation(s) present affects the rate of ingress of chloride ions because charge balance must be maintained, and the
associated cation(s) diffusing at a slower velocity will impede
the movement of chloride ions.
One of the most common objections to the use of ASTM
C1543 is its duration. As the specimens must be cured for
14 days and then dried for 28 days before the beginning of
the ponding, the 90-day version of the test takes 118 days,
or longer in the case of extended curing, to conduct, after
which the samples must be analyzed. Most test programs for
high-performance concretes would use a ponding period of
at least 180 days.
Although the ponding test does provide a crude onedimensional profile of chloride ion ingress, the profile is not
a reflection of chloride diffusion alone. The initial mode of
ingress of the ions is by sorption into the dried concrete. The
exposure of the bottom face to a 50 percent relative humidity
environment during the test induces vapor transmission from
the wet front on the top surface to the dry bottom surface,
and chloride ions penetrate by wick action (Buenfeld et al.
1995). Diffusion of the chlorides also takes place. McGrath
and Hooton (1999) observed that, while all three of these
mechanisms do occur in bridge decks, the test exaggerates
the importance of the sorption component.
The 28-day drying period before the ponding begins
significantly increases the apparent diffusion coefficient,
especially for concretes containing SCMs (Ngala and
Page 1997). For high-quality concretes, it may be difficult
to develop a chloride profile based on a 90-day ponding
period because so little chloride penetrates into the concrete.
Extending the ponding period to 180 days and increasing the
number of samples taken help to resolve this problem (Berke
and Hicks 1992; Andrade and Whiting 1996; Sherman et al.
1996; McGrath and Hooton 1999).
Precision of the sampling can make a significant difference in the conclusions drawn from the results. In analyzing
their data from AASHTO T259 (a predecessor of ASTM
C1543), McGrath and Hooton (1999) showed that imprecise
sampling makes it difficult to distinguish between a highquality concrete in which there is a high concentration of
chlorides near the surface, and a low-quality concrete in
which the chlorides penetrate much farther. Precision of the
sampling specified in ASTM C1556 avoids this problem.
ASTM C1556 avoids some of the problems associated
with ASTM C1543, as the concrete specimen is sealed on
all sides except the finished surface, and is saturated with
a CH solution before exposure to the sodium chloride solution. Thus, the chlorides penetrate into the specimen only
by diffusion, not by sorption. The specimen is then placed
in a concentrated (165 g/L) sodium chloride solution for at
least 35 days. Longer exposure times are recommended for
mature concretes, concretes with low w/cm, or high-performance concretes containing SCMs. In practice, the exposure
time could be extended to 90 days or longer.
The sampling is more precise than for ASTM C1543,
as indicated previously. A nominal 4 in. (100 mm) diam-
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
eter core from the slab is mounted in a mill or lathe and a
series of thin layers ground off. The dust from each layer is
collected separately and analyzed for acid-soluble chloride
content. The apparent diffusion coefficient is determined
using nonlinear regression analysis to fit the data to Fick’s
second law.
The use of Fick’s second law to reduce the data is a convenient but questionable practice. Pettersson (1994) noted that
the applicability of Fick’s second law, which is a simplification of a more general equation describing ion transport,
depends on the validity of three assumptions:
a) The material in which diffusion takes place is permeable and homogeneous.
b) The diffusion properties of the material do not change
with time or with the concentration of the diffusant.
c) No chemical reaction or physical binding of the diffusant occurs.
Pettersson (1994) further noted that all three of these
assumptions are violated in the diffusion of chloride ions
through concrete. That is, concrete is heterogeneous, its diffusion properties change with time and with the concentration
of the diffusant, and both chemical reactions and physical
binding can occur. ASTM C1556 uses the term “apparent
chloride diffusion coefficient” to make clear that the result
obtained is not a true diffusion coefficient. Although ASTM
C1556 in some cases may be conducted in less time than
ASTM C1543, it is still a lengthy test.
3.4.2 Fluids— One test for the absorption of water by hardened concrete is ASTM C642, in which a piece of concrete
at least 350 mL in volume is oven dried to constant mass and
then immersed in water until it again reaches constant mass.
The specimen is then boiled for 5 hours, allowed to cool, and
the mass determined again. The absorption after immersion
and the absorption after immersion and boiling are determined. This test is a measure of the absorption of the bulk
concrete. Note that oven drying may induce cracking in the
specimen, thus increasing the measured absorption.
ASTM C1585 measures the water sorptivity, which is the
rate of absorption, of a concrete surface, which is often of
greater interest than the bulk concrete. A cylinder or core
4 in. (100 mm) in diameter and 2 in. (50 mm) in length
is conditioned to an internal relative humidity of 50 to 70
percent and then sealed on all but one surface. The mass of
the specimen is determined initially and after being placed in
contact with water. The mass is determined at close intervals
initially and at longer intervals up to an exposure time of 7
days, after which one additional measurement is taken. The
initial slope of the absorption-versus-time curve is taken as
the rate of absorption.
Abbas et al. (1999) measured the permeability of concrete
to oxygen, which is easier than measuring its permeability to
water. Their calculations were based on Darcy’s Law, which
assumes laminar flow. This simplifying assumption is not
strictly true because in very small pores the flow is partly
molecular in nature. Gas permeability varies with the degree
of saturation of the concrete; the coefficient of permeability
varied over two orders of magnitude as a function of the
degree of saturation. Acceptable limits on the permeability
for durability vary depending on the exposure conditions
and performance requirements.
3.5—Obtaining durable concrete
Obtaining durable concrete for given conditions of exposure requires suitable mass transport properties for the fluids
and ions of interest. Proper attention to all aspects of good
concrete practice is important. The most significant factors,
however, are an appropriately low w/cm; judicious use of
SCMs; and good workmanship, including mixing, placement, compaction, and curing. Elevated curing temperatures
can be deleterious to the transport properties of concrete,
although the use of an appropriate combination of cementitious materials can largely mitigate this effect. Optimization
of the concrete mixture proportions should be done using
the curing regime anticipated on the job and a test method
that bears some relation to the anticipated exposure conditions. In particular, compressive strength is not a surrogate
for durability.
Proper attention to control of cracks is also important;
there is little to be gained from concrete of low permeability
between the cracks. Good aggregate grading to minimize the
paste content, control of temperature and moisture conditions, and appropriate structural design and detailing can
minimize the width of cracks. Further guidance is available from the Transportation Research Board (2006) and
Detwiler and Taylor (2003).
CHAPTER 4—FREEZING AND THAWING OF
CONCRETE
4.1—Introduction
Deterioration of concrete exposed to freezing can occur
when there is sufficient internal moisture that can freeze at
the given exposure conditions. The source of moisture can
be either internal or external. Internal is water that is already
in the pores of concrete that is redistributed by thermodynamic conditions to provide a sufficient degree of saturation
at the point of freezing to cause damage. External is when
the water enters the concrete from an external source, such
as rainfall). Dry concrete (generally below approximately 75
to 80 percent internal relative humidity) is normally immune
to damage from freezing.
Young concrete can be damaged by a single freeze (4.2.2).
Mature concrete may be able to withstand repeated cycles of
freezing and thawing without damage. Thus, concrete that is
properly cured and reaches sufficient maturity before being
exposed to freezing, such as concrete for columns or floor
slabs, may tolerate freezing from exposure to a single winter
season before it is protected from the elements. Similar
concrete that is not properly cured and is exposed to freezing
conditions at an early age, such as sidewalks and exposed
foundation walls, may show deterioration after a few years
of exposure to repeated cycles of freezing and thawing.
The severity of exposure should be quantified by a
combination of freezing, which is the number of annual
cycles of freezing plus average low temperature reached
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
during each cycle, and moisture condition before each
cycle of freezing (4.2.6.4).
4.1.1 Concrete made with durable aggregate—Concrete
made with aggregate that is resistant to freezing and thawing
is primarily protected from damage through the use of
entrained air in the concrete mixture, with secondary protection from steps taken to limit the fractional volume of freezable water in the concrete. A description of damage due to
freezing and thawing of concrete made with durable aggregates, along with protection methods, is given in 4.2.
4.1.2 Concrete made with frost-susceptible aggregate—
A properly proportioned concrete mixture that has received
adequate curing can suffer damage from freezing and thawing
if it contains an aggregate (generally the coarse aggregate)
that is susceptible to damage from freezing and thawing.
A description of damage in concrete made with aggregates
susceptible to freezing and thawing, along with a description
of aggregate identification procedures, is given in 4.3.
4.2—Frost attack of concrete made with durable
aggregates
4.2.1 Description of frost damage
4.2.1.1 Damage at early ages—Concrete in the early
stages of hydration ordinarily contains a considerable
amount of freezable water and has little or no tensile strength
to resist pressures due to freezing. Concrete that is allowed
to freeze under these conditions will develop ice lenses
approximately parallel to the surface exposed to freezing.
Additional ice lenses can develop under coarse aggregates.
When the concrete thaws and hydration resumes, the space
previously occupied by the ice lenses will form weak planes
that are susceptible to delamination or surface scaling.
4.2.1.2 Damage in cured concrete
4.2.1.2.1 Surface scaling—The most common form of
damage from freezing and thawing in hardened concrete is
surface scaling, that is, the loss of paste and mortar from the
surface of the concrete. Generally, layers less than 0.04 in.
(1 mm) thick are lost, but repeated cycles of freezing and
thawing can lead to removal of additional material. Scaling
is considerably accelerated by deicing salts. Vehicle traffic
or other surface contact can also accelerate scaling by
aiding in the removal of loosened material. Consequences
of scaling include change in appearance; change in surface
smoothness; and, in severe cases, loss of concrete cover over
reinforcing steel.
4.2.1.2.2 Internal deterioration—Though less common,
internal deterioration can have more severe consequences
than surface scaling due to the loss of structural integrity
of the concrete. Internal deterioration manifests itself as a
loss of strength in the paste of the concrete. Modern concrete
practice has practically eliminated this form of damage from
freezing and thawing by requiring a proper air-void system
and adequate curing before the first exposure to freezing
temperatures.
4.2.2 Preventing frost damage in new concrete
4.2.2.1 Protection from early freezing—Young concrete
should be protected from freezing by following the procedures and maintaining the minimum temperatures recom-
11
mended in ACI 306R. After consolidation and finishing, the
concrete should be protected from cooling too rapidly by the
use of insulated forms, curing blankets, and other procedures
described in ACI 306R. Allowing the concrete to cool too
rapidly could result not only in early freezing, but also in
thermal cracking of the concrete (ACI 306R; ACI 308R).
4.2.2.2 Minimum curing before freezing—Adequate
curing, including preventing excessive drying and maintaining adequate temperature, will ensure that the concrete
has hydrated sufficiently to substantially reduce the amount
of freezable water. A recommended minimum strength
that should be attained before the concrete temperature is
allowed to drop below freezing is 500 psi (3.5 MPa) (Powers
1962a). Once this strength has been achieved, a single freeze
will generally not permanently damage the concrete (ACI
308R). If repeated cycles of freezing and thawing are anticipated, the concrete should be kept warm long enough to
allow it to develop a compressive strength of at least 3500
psi (25 MPa) if it will not be exposed to deicing salts, and
4500 psi (32 MPa) if it will be exposed to deicing salts. The
strengths of 3500 and 4500 psi are average in-place strengths
needed before the concrete is exposed to repeated cycles of
freezing and thawing. Table 19.3.2.1 in ACI 318-14 refers
to the specified design strength. The two strengths are not
the same, as concrete could be exposed to repeated cycles of
freezing and thawing at an age significantly earlier than the
age associated with the specified design strength.
4.2.3 Preventing frost damage by proper design—Much
concrete now in service has withstood repeated cycles of
freezing and thawing for many years. While some of this
concrete has remained undamaged because it was never
allowed to contain enough freezable water to cause damage,
most of it has remained durable because proper precautions
were taken to avoid such damage (Mather 1990). The three
most important precautions to provide resistance to freezing
and thawing are discussed in 4.2.3.1 through 4.2.3.3.
4.2.3.1 Reducing freezable water—The likelihood of
damage from freezing and thawing is reduced by decreasing
the amount of freezable water in concrete. For conventional
mixtures, this has generally been accomplished by lowering
the w/cm, which is a maximum of 0.50 for moderate exposure and 0.45 for severe and very severe exposure (Table
4.2.3.1a), combined with adequate curing to ensure a
minimum compressive strength of approximately 3500 psi
(25 MPa) before exposure to repeated cycles of freezing and
thawing (4500 psi [32 MPa] if deicing salts are present).
Note that for instances in which corrosion is a concern, a
lower w/cm may be required if deicing salts are present (ACI
318-14).
Limiting the w/cm to a specified maximum has the effect
of reducing the amount of freezable water in the cured
concrete initially. Requiring a minimum strength before
freezing helps ensure that the tensile strength of the paste is
sufficient and the fractional volume that could be occupied
by freezable water in saturated concrete has been adequately
reduced by the formation of hydration products.
Modern concrete mixtures may contain admixtures, additives, and supplementary cementitious materials (SCMs)
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
Table 4.2.3.1a—Freezing-and-thawing exposure
classes
Exposure
Class
F0
F1
F2
F3
Maximum percent of total
cementitious materials by mass*
Severity
Condition
Cementitious materials
Not
applicable
Concrete not exposed to freezing conditions
Fly ash or other pozzolans
conforming to ASTM C618
25
Slag conforming to ASTM C989/
C989M
50
Moderate
Concrete exposed to freezing and thawing
conditions, but very low probability of
concrete being near saturation at time of
exposure*
Silica fume conforming to ASTM
C1240
10
Total of fly ash or other pozzolans,
slag, and silica fume
50†
Total of fly ash or other pozzolans
and silica fume
35†
Severe
Very severe
Concrete exposed to freezing and thawing
conditions, with a high probability of
concrete being near saturation at time of
exposure, but no deicing chemical exposure†
Concrete exposed to freezing and thawing
conditions as well as deicing chemicals‡
*Examples are vertical surfaces above the level of snow accumulation or horizontal
elevated floors in areas protected from direct exposure to moisture.
†
Examples are: vertical surfaces below the level of snow accumulation; vertical
surfaces with sufficient moisture exposure to allow the concrete to be near saturation
prior to freezing; retaining walls or other vertical elements with one side exposed to
moisture; and slab-on-ground that is not protected from freezing.
‡
Examples are: vertical surfaces that may have deicing-chemical-contaminated snow
piled against them; sidewalks or pavements that receive deicing chemicals; and
concrete that received frequent exposure to seawater as well as freezing-and-thawing
conditions.
Table 4.2.3.1b—Requirements by freezing-andthawing exposure class
Minimum
f c ,* MPa
(psi)
Maximum
w/cm†
Air content
F0
No
restriction
No
restriction
No
restriction
No restriction
F1
25 (3500)
0.50
Table
4.2.3.2.4
No restriction‡
F2
25 (3500)
0.45
Table
4.2.3.2.4
No restriction‡
F3a§
32 (4500)
0.45**
Table
4.2.3.2.4
Table 4.2.3.1c‡
F3b#
32 (4500)
0.45**
Table
4.2.3.2.4
No restriction‡
Exposure
Class
Table 4.2.3.1c—Cementitious materials limitations
for Exposure Class F3b
Limits on
cementitious
materials
*
The minimum average compressive strength that should be achieved before initial
exposure to freezing and thawing.
The maximum w/cm for the in-place concrete to provide adequate restriction of freezable water in the properly-cured concrete.
†
‡
High cementitious material replacement for portland cement frequently results in
lower rates of strength gain. Care should be taken to ensure that adequate curing (moisture, temperature, and time) is provided so that the minimum f c is achieved before
initial exposure to freezing and thawing.
§
Hand-finished surfaces.
#
Formed and machine-finished surfaces.
A lower w/cm may be needed when corrosion is of concern (ACI 318-14).
**
that permit reduction of the amount of freezable water in
concrete. Slag cement or pozzolans can refine the pore structure at a given w/cm, resulting in a smaller fraction of the
porosity containing freezable water. The use of slag cement
or pozzolans has the added beneficial effect of reducing
the rate that water can penetrate through the concrete. This
means that water removed by evaporation or hydration will
be replaced more slowly when the concrete is exposed to
*
The total cementitious materials also include ASTM C150/C150M, ASTM C595/
C595M, ASTM C845/C485M, and ASTM C1157/C1157M cements.
The maximum percentage should include:
(a) Fly ash or other pozzolans in Type IP blended cement, ASTM C595/C595M or
ASTM C1157/C1157M
(b) Slag used in the manufacture of an IS blended cement, ASTM C595/C595M or
ASTM C1157/C1157M
(c) Silica fume, ASTM C1240, present in a blended cement
†
Fly ash or other pozzolans and silica fume shall constitute no more than 25 and 10
percent, respectively, of total mass of the cementitious materials.
water. Numerous field observations have indicated when
surfaces are finished by hand, as opposed to machinefinished or cast against formwork, and exposed to deicing
salts (Exposure Class F3a in Table 4.2.3.1b), limits on the
SCMs may be required due to a variety of factors that appear
to include modification of the entrained air-void system
(4.2.3.2) as well as superficial changes to the w/cm. Table
4.2.3.1c gives recommended cementitious materials limitations for Exposure Class F3a (Thomas 1997).
Some researchers have hypothesized that it may be
possible to produce concrete with so little freezable water
that the concrete would not be damaged on freezing. Hooton
(1993) and Pigeon (1994) found that the cost and difficulty
in placing and finishing such low w/cm concrete would make
it impractical.
4.2.3.2 Entrained air-void system—Resistance to freezing
and thawing of a concrete mixture is substantially improved
by incorporating entrained air voids into the concrete. To
achieve maximum effectiveness, these air voids should
be evenly distributed throughout the paste portion of the
concrete. Their spacing should be close enough to prevent
the development of sufficient pressures from freezing to
fracture the concrete.
Because voids reduce the strength and stiffness of most
concrete mixtures, there is a natural tendency to limit
entrained air in concrete. An adequate air-void system,
however, is necessary for resistance to freezing and thawing.
The specific parameters normally used to evaluate an airvoid system along with the generally accepted minimum
(or maximum) values are described (4.2.3.2.1 through
4.2.3.2.4).
4.2.3.2.1 Spacing factor L —Spacing factor L is an
approximation of the average distance from anywhere in the
cement paste to an air void. The following assumptions are
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made for this parameter: the voids are spherical, of equal size,
and evenly distributed in a simple cubic lattice throughout
the paste (Powers 1949). The method for determining this
parameter is described in ASTM C457/C457M. The range of
spacing factors is generally from 0.004 in. (0.1 mm) or less to
approaching 0.04 in. (1 mm) for mixtures that do not contain
entrained air. The generally accepted maximum spacing
factor value for concrete with good resistance to freezing and
thawing is approximately 0.008 in. (0.20 mm) (Powers 1954;
Backstrom et al. 1954, 1958b). Other researchers (Dubovoy
et al. 2002; Attiogbe et al. 1992) have shown that concrete
with spacing factors greater than 0.008 in. (0.20 mm) may
exhibit satisfactory freezing-and-thawing resistance under
specific conditions in laboratory tests. CSA A23.1-14/CSA
A23.2 requires an average spacing factor of less than 0.009 in.
(0.23 mm) with no single value to exceed 0.010 in. (0.26 mm).
This is relaxed to an average value of 0.0098 in. (0.25 mm) for
concrete with specified strength greater than 10,000 psi
(70 MPa).
4.2.3.2.2 Specific surface α—Specific surface α is a
measure of the surface area per unit volume of voids. The
method for determining this parameter is described in ASTM
C457/C457M. A basic assumption of specific surface is that
all voids are spherical, which makes it a function of average
chord length alone (Powers 1949). Specific surface is, therefore, a good indicator of average void size. As average void
size goes up, specific surface goes down. Smaller voids
provide compliance with the L requirement at lower air
contents. The generally accepted value of specific surface
for resistance to freezing and thawing is a minimum of
600 in.2/in.3 (25 mm2/mm3) (Powers 1949), though other
researchers have shown that concrete with a specific surface
area less than 600 in.2/in.3 (25 mm2/mm3) may exhibit satisfactory freezing-and-thawing resistance in laboratory tests
(Dubovoy et al. 2002; Attiogbe et al. 1992).
4.2.3.2.3 Philleo factor F′—Philleo (1955) developed an
air-void parameter as a means to eliminate the assumptions
made for the spacing factor—namely, that all voids are of
equal size and spacing. His equation is based on the assumption that voids are randomly sized and distributed and establishes a relationship between the air-void distribution and
the percentage of paste that is within a given distance of an
air void. Philleo used the work of Willis and Lord (1951) to
establish a relationship between air-void chord lengths and
voids per unit volume. The equation is used to either determine the percentage of paste that is protected because it is
within a specified distance of an air-void, or alternately, to
determine the distance from an air-void resulting in a specified percentage of the paste. This distance, called the Philleo
factor, is often compared with the spacing factor. In actuality, however, the Philleo factor is more sensitive to the true
air-void distribution than the air content, paste content, and
number of voids that influence the spacing factor.
This parameter has not been widely accepted as a measure
of resistance of concrete to freezing and thawing, primarily
due to the difficulty in acquiring the data necessary for
calculation. These data consist of a record of all chord
13
lengths measured in a linear traverse (ASTM C457/C457M).
Most petrographers, however, use the alternate modified
point count method (ASTM C457/C457M), which does not
collect the necessary data. While no specific criterion for the
maximum F′, which is the distance for a given percentage
of paste protected, has been determined, an examination
of a considerable amount of linear traverse (ASTM C457/
C457M) data for a number of concrete specimens having
spacing factors of approximately 0.008 in. (0.20 mm) and
specific surface values of approximately 600 in.2/in.3 (25
mm2/mm3) suggests that a maximum acceptable distance
between an air void and 90 percent of the paste, P90′, should
be approximately 0.002 in. (0.04 mm) (Janssen and Snyder
1993, 1994).
4.2.3.2.4 Air content—The aforementioned air-void
parameters, while excellent indicators of the protection from
freezing and thawing provided by the air-void system, are
difficult to measure in the field. Total air content is therefore
generally specified and measured. Total air content includes
both the entrained air voids and the larger air voids that are
not removed by consolidation.
The use of an air-entraining admixture complying with
ASTM C260/C260M can provide a proper system of
entrained air-voids when a specified total air content of
the concrete mixture is achieved. The actual air content
necessary to ensure the production of the necessary airvoid system is affected by mixing action, workability of
the mixture, cement composition, types and amounts of
other admixtures, and others (Whiting and Stark 1983). In
addition, concrete handling during transport, placing, and
finishing can affect the entrained air-void system (4.2.4).
Recommended air contents for fresh concrete are given
in Table 4.2.3.2.4. These recommendations consider the
higher air requirements of concrete mixtures with higher
paste contents, which would be smaller nominal maximum
aggregate sizes, as determined by Klieger (1952, 1956) and
the severity of exposure; higher exposure severity increases
the probability of damage from freezing and thawing and,
therefore, demands greater protection. The values shown are
general recommendations; local conditions and experience
with specific mixtures, admixtures, and construction procedures could warrant other values.
Achieving the total air content specified in Table 4.2.3.2.4
does not always ensure frost protection of the paste. Rather,
a mixture should achieve the minimum and maximum
values for the air-void parameters discussed previously. In
most cases, the minimum air contents from Table 4.2.3.2.4
will achieve the necessary air-void parameters.
4.2.3.3 Design details—Physical details that allow for the
repeated wetting or restricted drying of concrete surfaces
should be avoided. Examples of these include roof driplines on sidewalks, and structural members directly below
unsealed joints of bridge slabs. Be careful to always provide
positive drainage of runoff from flat areas by methods
such as sloping surfaces, which is typically 2 percent.
Special precautions should be taken when runoff is likely to
contain deicing salts or other aggressive chemicals to avoid
increasing the risk of damage to concrete surfaces.
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
Table 4.2.3.2.4—Recommended air contents
Air content, percent*
Nominal maximum
aggregate size, in.
(mm)
Exposure Class F1
Exposure Class F2
and F3
3/8 (9.5)
7
7.5
1/2 (12.5)
7
7
3/4 (19)
6.5
7
1 (25)
6.5
6.5
1-1/2 (37.5)
6
6.5
2 (50)
6
6
3 (75)
5
5.5
*
Field tolerance on air content is recommended as ±1-1/2 percent. Air content recommendations are based on 18 percent air in the paste portion of the concrete with a
Vinsol resin air-entraining agent (from an analysis of work by Klieger [1952]). Mixture
proportions based on guidance in ACI 211.1 for angular coarse aggregates along with
the maximum w/cm values from Table 4.2.3.1b were used to determine the air content
recommendations. Mixtures using rounded aggregates will require approximately 1
percent less air due to the lower paste contents associated with rounded aggregates.
Consideration should also be given to structures that allow
water contact with the side away from freezing, such as water
tanks. In many cases, the evaporation rate on the outside of
the tank will exceed the rate that water can migrate through
the tank wall to replenish the evaporation. In these instances,
the amount of freezable water in the portions of the concrete
that freeze will be below critical degree of saturation. In
some cases, however, water migration to the freezing surface
can exceed the evaporation rate, leading to potential damage
from freezing and thawing. In very cold climates, freezing
can extend through the concrete wall, resulting in severe
damage. Low rates of moisture movement and consideration
of the depth of freezing are necessary in the design of these
types of structures (ACI 350).
4.2.4 Preventing frost damage by proper practice—A properly proportioned concrete mixture can still suffer damage
from freezing and thawing if either the air-void system or
the amounts of freezable water are adversely affected during
construction. The effects of construction practices on maintaining the quality of the concrete mixture are described in
4.2.4.1 through 4.2.4.4.
4.2.4.1 Transporting and placing—The air voids that
provide protection against freezing-and-thawing damage
are produced during mixing (Whiting and Stark 1983). Airentraining admixtures stabilize the bubbles that are produced
during mixing, but do not generate them. Therefore, mixing
is critical in forming and distributing the bubbles throughout
the mixture. Upon delivery and discharge of the concrete, the
number, volume, and size distribution of the bubbles depend
on the mixing action and the degree to which the particular
combination of ingredients has worked to generate, stabilize, and retain the bubbles.
The system of air bubbles in the fresh concrete at placement will be sensitive to many factors, including the type
and effectiveness of the mixer, duration of mixing, mixing
speed (rpm), haul time, batch size, and general condition of
the mixing equipment. These factors affect the time required
for the air-entraining admixture to stabilize the bubbles.
Fast-acting air-entraining admixtures may perform well
even with short mixing periods or brief haul times, whereas
more slowly-acting admixtures can perform better with a
longer mixing or haul time. The degree that air is incorporated in the fresh concrete during mixing depends on the
shearing, tumbling, or wave-breaking action of the concrete
in the mixer itself. This action will depend on the slump
of concrete and cleanliness and efficiency of the mixing
blades. Air incorporated in concrete is subdivided into
smaller bubbles by continued mixing and stabilized by the
air-entraining admixture, thus minimizing air loss. Immediately before being discharged, the freshly mixed concrete
contains air bubbles of a wide distribution of sizes and with
varying effectiveness in terms of providing frost resistance,
which all have their origin in the engulfment or entrapment
of air during mixing (Mielenz et al. 1958a,b; Backstrom et
al. 1958a,b).
Concrete is often placed by pumping from the delivery
truck to the formwork. Advances in pumping technology
have resulted in increased distances, lift heights, and delivery
rates. These advances have been accompanied by increases
in the pressure capacity of the pumps, with pressures in the
range of 300 to 500 psi (2 to 3.5 MPa) being fairly common
(Cooke 1990). Unfortunately, concerns have developed
when the air content measured after pumping was not the
same as the air content before pumping. In most cases,
the air content had decreased after pumping, but in some
instances, it increased (Cooke 1990; Dyer 1991; Whiting
and Nagi 1998). The possible loss of entrained air can be a
major concern for concrete exposed to severe freezing-andthawing environments.
Pumping concrete includes a number of activities that
could contribute to a change in the air content of the concrete.
During pumping, the concrete falls through a grating in the
pump hopper, is forced under pressure through a relatively
small-diameter pipe, moves through a series of bends, experiences changes in both elevation and pipe material, which is
generally steel to rubber, is then released from pressure as it
exits the pipe, and often falls some distance into the formwork. Pumping may also be continuous or interrupted at
various times. This sequence of events makes it difficult to
isolate the mechanism(s) that induce(s) the loss or gain of air.
A number of mechanisms have been proposed to explain
the change in air content that sometimes occurs when
concrete is pumped. During pumping, a vacuum could form
in the line, especially at low pumping rates with a long
portion of the pumping being downhill, which would enlarge
and remove air voids. Air voids could also be lost when the
concrete falls through the grating in the pump hopper or
when it falls into the formwork after exiting the pump line
(Yingling et al. 1992; Janssen et al. 1995). While changes
in the air-void system have been documented in multiple
cases (Whiting and Nagi 1998; Janssen et al. 1995; Pleau
et al. 1995), evidence of reduced resistance to freezing and
thawing has not been identified (Elkey et al. 1993; Whiting
and Nagi 1998).
4.2.4.2 Consolidating—On discharge of fresh concrete,
air voids could be entrapped in the concrete. While filling
forms, it is virtually impossible to avoid the inclusion of air
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pockets. When such large air pockets form during mixing,
continued mixing breaks them down to smaller sizes and
distributes them into the mixture. No such opportunity exists
once the concrete is cast. For this reason, it is necessary to
reduce the number and size of these pockets trapped in the
mixture by vibration.
Vibration accomplishes two purposes with regard to the
removal of air pockets in fresh concrete. First, vibration
can liquefy the concrete in the same manner as earthquakes
liquefy certain types of soils. Many of the large, buoyant air
pockets that were trapped in the semifluid mixture can rise
to the surface through the temporarily fluid mixture. Second,
vibration imposes a cyclic compression in the concrete
that locally increases, then decreases, the water pressure
surrounding the air bubbles. This causes the bubbles themselves to compress and decompress at the frequency of
the vibrator (Young 1989). Air bubbles break when forced
to compress and decompress at a critical frequency that
varies with their size—the larger the bubble, the lower the
frequency. Conversely, the higher the vibration frequency,
the smaller the bubble that can be broken. Simon et al.
(1992) showed that vibration at approximately 6000 vibrations per minute begins to break bubbles in the size range
of the so-called entrained air. Evidence also indicates that
some portion of the air from broken bubbles escapes, while
the rest is incorporated into other air bubbles in the mixture.
It takes time for the energy imparted by the vibrator to
liquefy the surrounding concrete and for the water pressure
to build up so as to compress the air bubbles. It also takes
time for the bubbles to rise to the surface, break, or both. For
these reasons, short intermittent vibration may have little
effect on the concrete in general or on the air bubble system.
Refer to ACI 309R for guidelines for appropriate vibration
procedures.
In summary, vibration of fresh concrete refines the airvoid system by encouraging the loss of larger air voids while
retaining the smaller ones. The exact distinction between
the size of bubbles removed and those left in place depends
on the concrete mixture and on the frequency, duration, and
intensity of vibration.
4.2.4.3 Finishing—After concrete is placed and consolidated, it is still possible to modify the air bubbles at the
surface of the fresh concrete, and therefore the air voids in
the hardened concrete, during finishing. Repeated passes of a
finishing tool can force bubbles together, resulting in fewer,
larger bubbles. The air content of the concrete at the surface
can be reduced by over-finishing, that is, over-manipulation
of the surface can reduce resistance to freezing and thawing.
This is particularly the case if the surface is finished while it
is still covered with bleed water or if water has been applied
to make it easier to finish. Finishing with free water on the
surface not only weakens that surface by increasing the w/
cm locally, but also increases porosity. When coupled with a
localized reduction in air content, resistance to freezing and
thawing is likely to be significantly reduced. Thomas (1997)
found that this was especially true for hand-finished concrete
having a high dosage of SCMs (50 percent Class F fly ash).
15
Machine-placed and finished areas of the same concrete did
not show scaling damage.
For concrete slabs, it has been found that hard-troweling of
air-entrained concrete has the potential to blister or delaminate at the surface (ACI 301; ACI 302.1R; Tarr and Farny
2008). Hence, as per ACI 301, concrete for slabs to receive a
hard trowel finish should not contain an air-entraining agent
or have a total air content greater than 3 percent.
4.2.4.4 Curing—Curing is defined as the maintenance of
a satisfactory moisture content and temperature in concrete
during its early stages so that the desired properties may
develop (ACI 308R). Overall, resistance to freezing and
thawing increases as continued curing develops microstructure and reduces the porosity and hydraulic conductivity
of the concrete. The result is a concrete that is less likely
to become critically saturated. Curing further increases the
compressive and tensile strength of the concrete, which
increases the resistance to pressure from freezing. These
attributes combine to produce a concrete that is less susceptible to freezing-and-thawing damage. The issue is made
more complicated, however, when concrete is cast in weather
where there is risk of freezing. In this situation, attempts
to cure the concrete to improve its overall durability can
conflict with the fact that curing can increase the risk of saturating concrete during exposure to early freezing. It is generally accepted that properly air-entrained concrete can sustain
one freeze cycle when it has attained a compressive strength
of 500 psi (3.5 MPa), and repeated freezing-and-thawing
cycles when saturated at a compressive strength (in-place) of
3500 psi (25 MPa). This means that curing procedures need
to be carried out to maintain moisture content to improve the
quality of the concrete, as well as measures to prevent early
freezing (ACI 306R; ACI 308R).
4.2.5 Preventing frost damage in existing concrete that
lacks adequate air-void system—Concrete that lacks an
adequate air-void system to protect it from the anticipated
exposure conditions is sometimes encountered. Protecting
the concrete from damage caused by freezing and thawing
then requires keeping it dry. Powers and Brownyard (1947)
presented the thermodynamic calculations to show that
concrete dried to 85 percent relative humidity at room
temperature would contain no freezable water at –0.4°F
(–18°C). Therefore, concrete dried to an internal relative
humidity below approximately 75 to 80 percent would
rarely, if ever, contain freezable water.
4.2.5.1 Sealers—Concrete that can be adequately dried
can sometimes be kept dry by sealing the surface of the
concrete with some sort of protective barrier system to
prevent the reintroduction of moisture. Slab-on-ground and
similar construction would also require a vapor retarder
beneath the concrete to reduce the movement of water in
either liquid or vapor form from entering the concrete from
underneath. Details of various protective barrier systems can
be found in ACI 515.2R. Vapor retarders for use under slabs
are discussed in ACI 302.1R.
In many cases, not all sides of a concrete member are
accessible for sealer treatments. Care should be taken so
that the sealed surface does not prevent the evaporation of
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
moisture that may have entered from unsealed surfaces. An
example would be the sealing of the top surface of a slabon-ground. The sealer could restrict evaporation of moisture
entering the slab from the bottom. The concrete could end
up wetter than if it had never been sealed. Vapor-permeable
sealers, which limit the intrusion of liquid water and permit
evaporation of moisture from the sealed surface, should be
evaluated for application-specific conditions.
4.2.5.2 Drainage and other methods—The moisture in
concrete can often be sufficiently limited to reduce the possibility of damage due to freezing and thawing if attention is
given to the removal of water from the area of the concrete.
While methods of moisture reduction are generally application-specific, the examples in 4.2.5.2.1 through 4.2.5.2.3 are
presented.
4.2.5.2.1 Drainage—Water allowed to pond on a concrete
surface can contribute significantly to the moisture in the
concrete. By providing adequate drainage to prevent ponding,
the water absorbed in the concrete can be minimized.
4.2.5.2.2 Maintenance—Designs that provide for rapid
drainage from concrete surfaces can be defeated by poor
maintenance. An example is the ponding of water on a
concrete slab being caused by leaves blocking a drain. Proper
design should be supported by adequate maintenance. Other
times, improper maintenance or other activities could result
in the unnecessary accumulation of water. Snow pushed off
a sidewalk and against a concrete wall could result in the
accumulation of moisture in the wall. As the snow against
the wall melts, the remaining snow could serve as a dam,
holding the water against the wall. Normally, a surface with
little moisture absorption, this vertical wall section could
absorb enough melted snow to lead to damage from subsequent freezing.
4.2.5.2.3 Redirection of water flow—Downspouts
emptying water across a concrete slab can be redirected
so that runoff does not flow across the concrete. This can
reduce the moisture exposure of the surface enough to
reduce damage from freezing and thawing.
4.2.6 Theories for frost damage—Damage in concrete
from freezing and thawing occurs as internal damage or
surface deterioration (Cordon 1966). Historically, the capillary-void system of the cement paste has been the focus of
most investigations. There is no consensus regarding the
mechanisms responsible for damage in cement paste. The
damage has been attributed to hydraulic pressure buildup
as water is forced away from the freezing front, to osmotic
pressure gradients driving water toward the freezing centers,
to vapor pressure potentials, and to combinations of these
processes (Powers 1945, 1954, 1955, 1975; Powers and
Helmuth 1956; Helmuth 1960a; Litvan 1972; Penttala 1998;
Setzer 1999, 2002; Scherer and Velenza 2005; Coussy and
Monteiro 2008, 2009). These are described in more detail in
the following sections.
4.2.6.1 Moisture expulsion—Powers proposed that ice
nucleated and grew in capillary pores, forcing them to dilate
or expel excess water from freezing sites. Elevated hydraulic
stresses would arise as water was expelled due to the relatively low permeability of cement paste. Distance from
the void boundary, the degree of saturation, and the rate of
freezing would influence the magnitude of hydraulic-stress
buildup (Cordon 1966; Powers 1945, 1954, 1955).
4.2.6.2 Osmotic pressure—The hydraulic-pressure
concept was later modified when experiments showed
significant evidence that moisture was moving toward, rather
than away from, freezing sites (Powers and Helmuth 1956;
Helmuth 1960a; Powers 1975). Marchand et al. (1995) and
Penttala (1998) proposed that not all of the water in the capillary pores is freezable due its surface tension and the small
diameter of the pores containing it. Water in the largest voids
would freeze before the water in the smaller voids. When
the water in the larger voids freezes, the concentration of
the dissolved salts increases locally, causing a concentration
gradient in the pore solution. Water is thought to move from
the smaller voids to the larger ones to reduce this gradient.
The resulting flow is thought to cause damage. Litvan (1972)
proposed a similar theory, also founded on thermodynamic
arguments, but cast in terms of vapor-pressure gradients
between supercooled water and ice instead of salt concentration gradients.
Penttala (1998), Scherer and Valenza (2005), Setzer
(1999, 2002), and Coussy and Monteiro (2008, 2009) have
combined the moisture expulsion and osmotic pressure theories to account for the rate of freezing, degree of saturation,
dispersion of air voids, and the paste microstructure. These
theories agree that forces are exerted on the paste from
the movement of water from the small to the large voids.
Scherer and Valenza (2005) add that when the larger voids
fill with ice, local pressure from ice crystallization would be
expected. These pressures will increase with the shape and
curvature of the pores. A combination of these factors lead
to initial damage.
4.2.6.3 Ice lens growth—Whereas the previous mechanisms
may describe the deterioration in small saturated samples
frozen rapidly and cyclically in the laboratory, progressive
ice accretion in cracks during periods of sustained temperatures slightly below 32°F (0°C) may generally dominate the
further degradation of previously damaged or microcracked
concrete (Litvan 1978). A theoretical model was developed
and proven successful for predicting freezing-and-thawing
damage in rock (Walder and Hallet 1985; Hallet et al. 1991).
In this model, freezing attack in homogeneous porous solids
is viewed as occurring in an open system where microcracks
are internally pressurized by ice accretion fueled by migration moisture from either liquid or vapor; the latter is in turn
induced by thermally driven free-energy gradients. This ice
accretion model, also known as the segregation ice model, is
consistent with Gilpin’s (1980) study of freezing effects in
porous media as applied to soils.
4.2.6.4 Implications of freezing-and-thawing damage
mechanisms—Uncertainty and opposing views of fundamental processes governing freezing in concrete undermine efforts to develop tests of the resistance to freezing
and thawing of concrete. Several testing strategies have
been used that involved relatively rapid cycling of freezing
and thawing in saturated and dry environments. In North
America, a freezing-and-thawing cycling test, ASTM C666/
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
C666M, is used to determine the resistance of concrete
mixtures to internal damage. ASTM C672/C672M evaluates
scaling resistance. The latter relies on visual classifications,
sometimes supplemented with mass loss measurements.
These tests, like some of their predecessors, are criticized
for not adequately representing typical environmental conditions. The tests cycle between extreme temperatures too
rapidly. The CDF test (RILEM TC 117-FDC 1996) provides
greater reproducibility and is more quantitative with respect
to scaling measurements. Hallet et al. (1991) proposed that
conventional tests give only limited guidance for understanding the processes governing freezing at field sites
where the thermal or hydraulic regimes are very different
from those in the laboratory. Extrapolation from diurnal or
more frequent freezing-and-thawing experiments to field
conditions should be viewed with particular caution because
distinct physical processes may govern each. This situation
calls for consideration of diagnostic freezing-and-thawing
tests in which both processes can be distinguished and
probed systematically. This also suggests that considerably
more caution is needed when attempting to relate standard
laboratory test results to spacing factors and to other design
criteria for effective resistance to freezing and thawing in
service. Some concrete mixtures pass accepted laboratory
tests and do not perform well in the field, while others fail
the tests and perform quite satisfactorily in the field. This is
probably because the tests for internal damage and scaling
due to freezing and thawing do not address all of the significant variables. Stark (1989a) indicated that the potential
role of several key factors in freezing-and-thawing damage
has not been appreciated; these include the magnitude and
duration of exposure to sustained temperature and moisture
gradients, and the cumulative time of exposure to specified
temperature ranges.
4.3—Frost attack of concrete made with
nondurable aggregates
Deterioration due to freezing and thawing of properly
proportioned, air-entrained concrete made with aggregate susceptible to freezing-and-thawing damage is often
referred to as D-cracking. Many types of coarse aggregate
have been identified as susceptible to D-cracking, while
other sources of the same kind of rock have not been found
susceptible. The pore structure of the coarse aggregate is
thought to be the primary contributing factor to susceptibility to D-cracking. Descriptions of the appearance and
development of D-cracking are presented in 4.3.1, while the
prevention of D-cracking in new construction is covered in
4.3.2. Mitigation of D-cracking in existing construction is
given in 4.3.3. Theories and mechanisms of D-cracking are
covered in detail in 4.3.4.
4.3.1 Description of D-cracking
4.3.1.1 General description—D-cracking is characterized by cracks through the coarse aggregate and mortar of
the concrete. Away from the cracks, both mortar and coarse
aggregate are strong and show no signs of deterioration.
The development of D-cracking requires considerable
moisture and repeated cycles of freezing and thawing. As
17
a result, D-cracking usually appears close to joints, cracks,
edges, and corners where moisture enters from more than one
surface. D-cracking generally appears as a series of cracks
approximately parallel to the primary moisture source.
4.3.1.2 Flatwork—The most common appearance of
D-cracking is in at-grade flatwork such as highway pavements, parking lots, and sidewalks. These are areas that
frequently have readily available moisture from precipitation runoff, from multiple directions at joints and cracks,
and on the bottom of the slab unless there is an effective
vapor retarder beneath it. Often, the earliest appearance
of D-cracking will be at the intersection of transverse and
longitudinal joints in a pavement. At these locations, moisture is often available at the tops and the bottoms of the
vertical joint faces.
In climates cold enough to freeze through the thickness
of the concrete slab, D-cracking usually starts at the bottom
and progresses to the surface (Schwartz 1987). This is probably due to the greater availability of moisture at the bottoms
of slabs. By the time telltale cracks appear on the surface,
deterioration can extend 1.5 ft (0.5 m) or more away from
the joint.
The appearance of D-cracking in milder climates is somewhat different because the concrete never freezes all the way
through. D-cracking in these cases often appears as shallow
spalling at joints; closer examination reveals the characteristic cracks parallel to the joint.
4.3.1.3 Vertical construction—Though much less
common, D-cracking can also appear on vertical construction. Construction details, maintenance practices, or both,
that allow the accumulation of moisture against the corners
of walls or columns can contribute to D-cracking if susceptible aggregates were used in the construction. An example
of such maintenance practices at a building would be the
shoveling of snow off a sidewalk and depositing it against a
concrete foundation wall.
4.3.2 Prevention of D-cracking
4.3.2.1 Role of mixture proportioning—The primary factor
in a concrete mixture that contributes to the development
of D-cracking is the susceptibility of the coarse aggregate,
whereas the air-void system and the w/cm have little or no
effect (Schwartz 1987; Missouri Highway and Transportation Department 1990). Most coarse aggregates identified as
susceptible to D-cracking are sedimentary rocks, although
many sedimentary rocks have not been found to be susceptible to D-cracking. Igneous rocks are generally not considered to be susceptible to D-cracking unless the rocks are
weathered. Weathered rocks would probably be undesirable
for concrete production anyway due to their low strength and
likelihood to break down from handling. Most metamorphic
rocks have not shown D-cracking susceptibility; however,
some partially metamorphosed sedimentary rocks have been
identified as susceptible (Stark 1976).
The maximum aggregate size is also important in the
development of D-cracking. Numerous studies (Stark and
Klieger 1973; Klieger et al. 1974; Stark 1976; Missouri
Highway and Transportation Department 1990; Almond
and Janssen 1991) have shown that reducing the nominal
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maximum size of the aggregate reduces its susceptibility.
Unfortunately, reducing the nominal maximum aggregate
size can have less-desirable side effects, including increased
paste demand to maintain workability at a given strength
level, increased drying shrinkage potential, and reduced
joint load transfer in pavements.
4.3.2.2 Importance of aggregate identification—Dcracking can require a number of years to fully develop,
during which time much susceptible concrete could be
placed before a problem is identified. This, combined with
natural variability of aggregate sources, leads to the need for
identification of D-cracking-susceptible aggregates before
they are used in concrete exposed to moisture and cycles of
freezing and thawing.
D-cracking has been known since the 1930s (Stark and
Klieger 1973); wide ranges of tests have been developed to try
to identify susceptible aggregates. The most common procedure for identifying susceptible aggregates is ASTM C666/
C666M. Concrete specimens made with the aggregate in question are subjected to repeated cycles of freezing and thawing
in the laboratory and are evaluated in terms of either increase
in length or decrease in dynamic modulus of elasticity.
4.3.2.3 Aggregate beneficiation—A variety of techniques
have been proposed to improve the performance of susceptible aggregates. Limiting the nominal maximum size was
discussed in 4.3.2.1. Schwartz (1987) summarized other
methods, including coating the aggregates to prevent their
absorption of water, heavy media separation, and blending
durable aggregate with a nondurable aggregate to reduce
its D-cracking susceptibility. He reported that aggregate
size reduction was the most effective method of reducing
D-cracking susceptibility.
4.3.3 Mitigation of existing D-cracking—A considerable
amount of concrete containing D-cracking-susceptible aggregates has been placed where it is exposed to both moisture
and cycles of freezing and thawing. This is especially true
of concrete pavements. Joint deterioration associated with
D-cracking can significantly reduce the service life of such
pavements. The concrete as little as 1.5 ft (0.5 m) away from
the joints often shows no deterioration or loss of strength.
A typical concrete pavement can have transverse joints
12 ft (4 m) or more apart. With less than 1.5 ft (0.5 m)
of D-cracked concrete at each end of a slab, most of the
concrete is in good condition. While replacement of the
deteriorated concrete near the joints with a full-depth patch
would seem to be a cost-effective method of extending the
life, D-cracking often appears at the newly created joints
adjacent to the patches in as little as 5 years. Thus, the
D-cracking continues as before, but at twice as many joints
(Janssen and Snyder 1994).
4.3.3.1 General—Three conditions are necessary for the
development of D-cracking: concrete made with susceptible
aggregate, moisture, and cycles of freezing and thawing. As
the aggregates already in the concrete would be difficult or
impossible to render nonsusceptible, D-cracking mitigation
should attempt to either prevent freezing and thawing or
remove the source of moisture.
4.3.3.2 Preventing freezing—Portland-cement concrete
pavements often receive asphalt concrete overlays to improve
the condition of the pavement and extend its life. In climates
that do not get too cold in winter, freezing in a concrete pavement that contains D-cracking-susceptible aggregates could
be prevented by covering it with a sufficient thickness of
asphaltic concrete. Janssen et al. (1986) found that the freezing
should be almost completely prevented in the concrete to stop
the progression of D-cracking; merely decreasing the number
of cycles of freezing and thawing with an overlay could actually accelerate the rate of D-cracking due to the increased
potential for moisture migration during slower freezing rates.
More than 6 in. (150 mm) of asphalt concrete overlay would
be required to prevent freezing at the surface of the concrete
pavement for central latitudes of the United States. Using
asphalt concrete overlay to prevent freezing in concrete made
with D-cracking-susceptible aggregates is probably not an
effective D-cracking mitigation method for these conditions
(Janssen et al. 1986; Janssen and Snyder 1994).
4.3.3.3 Reducing moisture—The use of sealers on the
cut ends of the existing concrete pavement sections (4.3.3)
before placing the patches, could reduce the lateral movement of moisture into the concrete. This could increase the
time before D-cracking appeared in the patched concrete.
This method was attempted in a D-cracked concrete pavement section in Ohio in 1992 (Janssen and Snyder 1994).
Though initial laboratory testing indicated that sealer treatment delayed the resumption of D-cracking, field monitoring showed that the D-cracking reappeared after six years
(Janssen 2001).
4.3.4 Theories and mechanisms of D-cracking—Theories of damage to concrete from freezing and thawing have
already been discussed in 4.2.5. With the exception of the
role of air voids in protecting concrete from damage, these
same theories generally apply to D-cracking. This section,
therefore, will concentrate on the characteristics of specific
aggregates that make them susceptible, while other aggregates of the same type are not susceptible.
4.3.4.1 Pore size and size distribution—Kaneuji et al.
(1980) observed qualitative correlations between concrete
durability and pore size distributions of aggregates. At a
constant total pore volume, aggregates with smaller pores
result in a lower resistance to freezing and thawing. For
aggregates with similar predominating pore sizes, a greater
pore volume results in less resistance to freezing and thawing
aggregate. By correlating aggregate service records with
mercury porosimetry studies, Marks and Dubberke (1982)
found that with one exception, the D-cracking-susceptible
aggregates analyzed exhibited a predominance of pore sizes
of 1.5 × 10–6 to 8 × 10–6 in. (0.04 to 0.2 μm) whereas aggregates with good to excellent service records had a majority
of pores that were larger than this.
4.3.4.2 Deicing salt effect—Dubberke and Marks (1985)
noted a reduced resistance to D-cracking for some aggregates
when pavements containing susceptible aggregates were
exposed to deicing salt. Other aggregates that they examined showed no effect. A possible explanation is a change
in pore structure due to etching of the pore walls by the salt.
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Such action has been identified for aggregates containing
calcite (Gillott 1978). This possible deicing salt effect should
be noted when using field performance records to evaluate
the D-cracking potential of an aggregate. Satisfactory performance in a pavement that never received deicing salt may not
ensure the same performances where deicing salts are used.
CHAPTER 5—ALKALI-AGGREGATE REACTION
5.1—Introduction
This chapter presents guidelines for minimizing the risk
of damaging expansion caused by alkali-aggregate reaction (AAR) in concrete construction. This risk may include
potential damage from alkali-carbonate reactive or alkalisilica reactive aggregates. Procedures are discussed for evaluating aggregates and selecting appropriate measures for
controlling expansion are discussed. ACI 221.1R provides
more detailed information on types of reaction, reaction
mechanisms, reactive rock types, methods of testing aggregates, and preventive measures.
Alkali-silica reaction (ASR) was first detected in the late
1930s and first reported in 1940 (Stanton 1940a,b). Since its
initial detection, ASR has been identified as a major cause of
premature concrete deterioration throughout the world. Since
Stanton’s first investigations, a wide range of testing procedures for assessing aggregate reactivity has been developed.
Most recently, effort has focused on developing accurate
testing methods to determine the effectiveness of mitigation
measures. Much focus has been placed on the incorporation
of supplementary cementitious materials (SCMs) as well as
chemical admixtures, namely, lithium salts such as LiNO3.
The reliability of these techniques varies and depends on, to
some extent, the nature of the aggregate being tested and the
testing environment.
The first case of alkali-carbonate reaction (ACR) was
observed in Ontario, Canada, in the late 1950s (Swenson
1957). ACR occurs between alkali hydroxides and certain
argillaceous dolomitic limestones. This reaction is characterized by rapid expansion and extensive cracking of the
affected concrete. Structures suffering from ACR generally exhibit deleterious effects within 5 years or less from
initial construction. The only way to avoid ACR is through
selective quarrying to avoid construction with potential
alkali-carbonate reactive aggregates. This type of reaction is
limited to select geographical regions.
Criteria for interpreting test results vary among different
national standards. They also differ among and within states
or provinces, with different limits being adopted by various
local agencies and state or provincial authorities. Methods
for controlling expansion due to AAR also vary regionally. Many specifications do not permit the use of reactive
aggregates. When reactive aggregates are used in concrete,
recommended preventive measures include limiting the
alkali content of the concrete; using SCMs such as fly ash,
slag cement, silica fume, or natural pozzolans; using lithium
nitrate; or combining these methods.
A history of satisfactory field performance may be the
most effective method for evaluating the potential for an
19
aggregate to cause AAR (5.3.1). Where such satisfactory
field performance can be demonstrated, aggregates may be
accepted for use in concrete without AAR testing, provided
that similar materials are incorporated in the batching
process. An example is concrete with an alkali content less
than or equal to that of the satisfactory concrete in service.
In the absence of such field performance data, however,
aggregates should be subjected to suitable laboratory testing
procedures to establish their degree of reactivity. If the
results of such laboratory testing do not indicate a potential
for AAR, aggregates may be used without any precautionary
measures. Even aggregates that demonstrate the potential
for ASR may be used in concrete, provided that suitable
measures are implemented to control the risk of expansion.
Alkali-carbonate reactive aggregates are normally avoided,
however, as it has been proven difficult, or economically
unfeasible, to control expansion with such materials.
5.2—Types of reactions
Two types of AAR have been recognized: ACR and ASR.
Alkali-carbonate reaction is associated with the use of certain
argillaceous dolomitic limestones. Confirmed cases of ACR
have been restricted to a few locations in North America:
mainly in Virginia, Kentucky, Indiana, Iowa, Illinois, and in
Ontario, Canada. Alkali-carbonate reaction involves a reaction between an alkali source and certain calcium-magnesium carbonate rocks (dolomites). Alkali-silica reaction
is distinctly different, and results from a reaction between
alkali hydroxides in the pore solution and certain forms of
reactive silica present in some types of siliceous or carbonate
aggregates. Alkali-silica reaction can occur in limestone
aggregates that contain siliceous components as well. Table
5.2 presents a list of several common reactive rock types and
mineral forms that are susceptible to ASR.
Alkali-silica reaction is far more widespread than ACR,
and is further subdivided into two categories: 1) reactions
involving poorly crystalline or metastable silica materials;
and 2) reactions involving certain varieties of quartz. From
an engineering perspective, the main distinctions between
these two categories of ASR involve the time to the onset of
expansion and cracking, and the perceived duration of the
reaction in the field. Reactions involving such silica materials, which are sometimes referred to as classical ASR,
are characterized by a relatively short time to the onset of
cracking where cracking usually occurs within 5 to 10 years,
whereas the manifestation of reactions involving quartz
minerals usually takes much longer, although the reaction
may continue for many decades.
5.2.1 ACR background—The ACR occurs between alkali
hydroxides and certain argillaceous dolomitic limestones;
these dolomites are characterized by a matrix of fine calcite
and clay minerals with scattered dolomite rhombohedra. The
reaction is manifested in the rapid expansion and extensive
cracking of concrete; structures affected by ACR usually
show cracking within 5 years. Although there is a lack of
consensus regarding the precise mechanisms involved, it
is generally agreed that the reaction is accompanied by the
dedolomitization process, as follows
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20
Table 5.2—Some examples of rock types and
minerals susceptible to ASR
Reactive rocks
Reactive minerals
Shale
Opal
Sandstone
Tridymite
Silicified carbonate rock
Cristobalite
Chert
Volcanic glass
Flint
Cryptocrystalline (or microcystalline)
quartz
Quartzite
Strained quartz
Quartz-arenite
Gneiss
Argillite
Granite
Greywacke
Siltstone
Arenite
Arkose
Hornfels
Ca·Mg(CO3)2 + 2ROH → CaCO3 + Mg(OH)2 + R2CO3
(5.2.1a)
(dolomite + alkali hydroxide → calcite + brucite + alkali carbonate)
where R represents K or Na. Because this reaction results in
a reduction in solid volume, however, the expansion must
be attributed to an alternative mechanism. Several theories
have been proposed to explain the expansion mechanism
(Swenson and Gillott 1964; Tang et al. 1987; Fournier and
Bérubé 2000), which include:
a) Hydraulic pressures caused by the migration of water
molecules and alkali ions into the restricted spaces of the
calcite/clay matrix around the dolomite rhombs
b) Adsorption of alkali ions and water molecules on the
surfaces of the active clay minerals scattered around the
dolomite grains
c) Growth and rearrangement of the products of dedolomitization (brucite and calcite)
The alkali carbonate produced in the dedolomitization
reaction may react with calcium hydroxide (CH) in the
cement paste as follows
R2CO3 + Ca(OH)2 → CaCO3 + 2ROH
(5.2.1b)
thereby regenerating alkalis for further reaction. Thus,
provided there is sufficient alkali available to initiate the reaction, the process may continue independently of the amount
of alkalis available in the concrete. This could explain why
low-alkali cements are not effective in controlling damaging
reaction in some instances (Thomas and Folliard 2007).
5.2.2 ASR background—Alkali-silica reaction has been
a known cause of concrete deterioration for more than 70
years. Since its discovery in the late 1930s (Stanton 1940a,
b), ASR has been observed as a cause of premature concrete
deterioration throughout the world. Although the factors that
lead to deleterious ASR are commonly agreed on, the mechanism by which the alkali-silica gel causes expansion and
subsequent cracking in concrete is not yet entirely understood by researchers in the field.
Alkali-silica reaction is a chemical reaction that is the
result of hydroxyl ions attacking siliceous species in certain
aggregates. The attack liberates silica, which then combines
with alkalis (Na+ and K+) and with lesser amounts of calcium
(Ca++) that are present in the concrete pore solution to maintain charge balance. The resulting alkali-silica gel then
absorbs water and expands, which may result in cracking
of the aggregates, the cement paste, and ultimately the
concrete matrix. For the ASR to cause damage in concrete,
it is widely accepted that three components are necessary:
sufficient alkali, reactive silica, and adequate moisture.
5.2.2.1 Alkalis—The alkalis (Na+ and K+) are typically
supplied by portland cement. However, SCMs; chemical
admixtures; and external sources such as seawater, deicing
salts, and anti-icing chemicals can also contribute to the
alkalinity of the pore solution. Certain aggregate species,
particularly those containing feldspars, may also release
alkalis to the pore solution (Bérubé et al. 2002). The amount
of alkali in cement is usually expressed as the sodium oxide
equivalency, written Na2Oeq.
Equation (5.2.2.1) is used to determine the sodium oxide
equivalency in the portland cement
Na2Oeq = Na2O + 0.658K2O
(5.2.2.1)
where Na2Oeq is the total sodium oxide equivalent, in percent
by mass; Na2O is sodium oxide content, in percent by mass;
and K2O is potassium oxide content, in percent by mass.
The concentration of alkalis in portland cement generally
ranges from 0.2 to 1.3 percent Na2Oeq, which is relatively
low in comparison to other compounds and oxides. Initial
research on ASR proposed that expansion due to ASR was
unlikely to occur when the percentages of alkalis in the
cement fell below 0.6 percent Na2Oeq (Stanton 1940b). This
approach has since been used as a mitigation option to limit
ASR in new concrete. Reducing the percent contribution
of alkalis from portland cement, however, does not effectively mitigate ASR for all reactive aggregate types because
it does not limit the total alkali content of the concrete or
alkali loading; within the pore solution of concrete, the
alkalis dissociate in solution, leaving K+ and Na+, which
must then be balanced by an equivalent concentration of
hydroxyl ions (OH–) to maintain charge equilibrium. The
increased concentration of dissociated alkalis in the concrete
pore solution effectively increases the concentration of
hydroxyl ions, which in turn increases the pH in the pore
solution. As referenced in 5.2.2.4, this OH–-induced increase
in pH, if high enough, leads to the initial breakdown of the
reactive silica in the aggregate, resulting in the formation
of alkali-silica gel. Diamond (1983a) reports that the OH–
ion threshold concentration for ASR is unlikely to be less
than 0.25M and Kollek et al. (1986) suggest a threshold of
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0.20M; this equates to a pH threshold of 13.2 to 13.3 for
ASR to occur.
5.2.2.2 Reactive silica—The degree of reactivity of aggregates depends on a number of factors, including the mineralogy of the aggregate, and the crystallinity and solubility
of the silica. Reactive silica is amorphous or disordered
silica found in certain aggregates. This poorly crystalline
silica dissolves more readily in the alkaline pore solution of
concrete than do well-crystallized or dense forms of silica.
Because of its increased solubility, amorphous silica is more
susceptible to ASR. Common reactive minerals susceptible to ASR include microcrystalline, cryptocrystalline,
and strained quartz; cristobalite; tridymite; opal; chert; and
volcanic glass such as obsidian.
5.2.2.3 Adequate moisture—The third and final component necessary for ASR to occur is adequate moisture, which
is one of the key components in the expansion of the gel.
Water is found within the pore solution of concrete and is
also introduced from external sources. A minimum relative
humidity of 80 percent is required to provide enough moisture to drive the expansion of the alkali-silica gel and sustain
the reaction (Pedneault 1996).
5.2.2.4 Mechanism of gel formation—The term “ASR” is
somewhat misleading, as the initial reaction occurs between
the reactive siliceous aggregate and the hydroxyl (OH–) ions
and not the alkalis of the pore solution. The high concentration of the hydroxyl ions in the pore solution is equal to that
of the alkali cations to maintain charge equilibrium. This
high OH– concentration is what results in a high pH, which,
in turn, leads to the initial breakdown of the reactive silica
in the aggregate. When poorly-crystalline hydrous silica is
exposed to a strong alkaline solution, there is an acid-base
reaction between the hydroxyl ions in solution and the acidic
silanol (Si-OH) groups (Dent Glasser and Kataoka 1981) as
follows
≡Si-OH + OH– → ≡Si-O- + H2O
(5.2.2.4a)
As further hydroxyl ions penetrate the structure, some of
the siloxane linkages (Si-O-Si) are also attacked as follows
(Dent Glasser and Kataoka 1981)
≡Si-O-Si≡ + OH– → 2 ≡Si-O- + H2O (5.2.2.4b)
The negative charges on the terminal oxygen atoms are
balanced by alkali cations (Na+ and K+) that simultaneously
diffuse into the structure. The disruption of siloxane bridges
weakens the structure and, provided sufficient reserves of
alkali hydroxide are available, the process continues to
produce an alkaline silicate solution. The extent or rate of
dissolution is controlled by the alkalinity of the solution and
the structure of the silica.
5.2.2.5 Role of calcium—Bleszynski and Thomas (1998)
and Thomas (2006a) concluded that significant expansion
only occurs when an adequate supply of calcium is available as CH. In systems with abundant alkali hydroxides
and reactive silica but no CH, silica dissolved and remained
in solution. Although the precise role calcium plays in gel
21
expansion remains unclear, a series of mechanisms has been
proposed:
a) Calcium can replace alkalis in the reaction product,
regenerating alkalis for further reaction (alkali recycling)
(Thomas 2006b; Hansen 1944)
b) Calcium hydroxide may act as a buffer maintaining a
high level of OH– in solution (Wang and Gillott 1991)
c) High calcium concentrations in the pore solution prevent
the diffusion of silica away from reacting aggregate particles
(Bleszynski and Thomas 1998; Chatterji 1979; Chatterji and
Clausson-Kaas 1984)
d) If calcium is not available, reactive silica may merely
dissolve in alkali hydroxide solution without causing damage
(Thomas 2006a; Diamond 1989)
e) The formation of calcium-rich gels is necessary to
cause expansion either directly or through the formation
of a semi-permeable membrane around reactive aggregate
particles (Thomas 2006a; Thomas et al. 1991; Bleszynski
and Thomas 1998)
Promoting the formation of C-S-H at the expense of CH
(for example, through the use of pozzolans) may result in
successful mitigation of expansion due to ASR.
5.2.3 Mechanism of gel expansion—Although the mechanisms behind the formation of the gel are well understood,
the actual mechanism for expansion of gel remains uncertain. Four main theories have emerged over the past 70 years
to explain the mechanism of gel expansion, all maintaining
that water is the main component driving the process. The
four theories of expansion include the double-layer, osmotic
pressure, CSH-shell, and the calcium/alkali exchange theory.
The osmotic pressure theory speculates that the cement
paste surrounding the reactive aggregates acts as a semipermeable membrane, preventing the presence of large silicate ions while allowing the water and alkali hydroxides to
diffuse through. Under these conditions, the alkali silicate
formed on the surface of the aggregate particle draws solution through the cement paste, resulting in continued formation of the alkali silica gel. As the gel continues to swell, an
osmotic pressure cell is formed and increasing hydrostatic
pressure is applied to the cement paste, eventually resulting
in cracking (Hansen 1944).
The C-S-H shell theory considers the effect of calcium
on the durability of concrete. This theory hypothesizes
that in the presence of CH, alkali ions from alkali salts and
hydroxyl ions from Ca(OH)2 enter the reactive silica aggregate grains, leaving calcium and anions in the pore solution.
The penetration of the solvated hydroxyl and alkali ions
causes the Si-O-Si bonds of the reactive aggregate to break
apart, opening the grains for further penetration of ions,
and permitting the release of some SiO2 into the pore solution. As the solvated hydroxyl and alkali ions infiltrate the
aggregate grains, calcium, hydroxyl, and water molecules
also migrate into the reactive siliceous material. When high
concentrations of Ca(OH)2 and alkali salts are present in the
pore solution, only a limited amount of SiO2 can diffuse out,
while additional materials penetrate into the aggregate structure. This imbalance results in an expansive force within
the aggregate grain. If, however, the Ca(OH)2 and alkali
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salt concentrations are low, the rate of penetration of the
hydroxyl and alkali ions is also low, while the migration of
SiO2 is increased, thus resulting in a lower expansive force
(Chatterji et al. 1987).
The calcium/alkali exchange theory also considers calcium
to be an essential component in the expansion of alkali-silica
gel. This theory hypothesizes that the gel absorbs Ca2+ ions
and, in turn, exchanges them for alkali ions, which then react
with the pore solution to create additional alkali-silica gel.
As additional gel forms, it fills more space and eventually
causes cracking. The ion and water uptake is governed by
the temperature and moisture conditions of the material and
thus is coupled to physical transport (Rotter 1995).
The double-layer theory suggests that the expansion of
the gel is caused by swelling due to electrical double-layer
repulsive forces. When a liquid and a solid come into contact,
the surface of the solid carries excess charge, which electrifies the interface. This excess charge alters the properties
of both the solid and liquid materials. Alkali-silica reaction
involves the interaction of the highly-charged silica aggregate surface with the alkaline pore solution of the concrete.
This is the reaction that leads to the breakdown of the silica
and the formation of alkali-silica gel. Within the gel, negatively charged solid silica particles attract positively charged
cations that thus bind to form a rigid layer around the solid
particle. Surrounding this rigid layer, a diffuse layer is formed
that comprises more cations and anions found within the gel.
The electrical double layer, therefore, is composed of sodium,
potassium, and calcium ions, which surround the negatively
charged silica surface. Once this double layer has been established, it imbibes water and swells. As water is introduced into
the layer, electrostatic forces predominate and particles are
pushed apart as the gel expands (Prezzi et al. 1997).
While it is commonly understood that water is the primary
driving force for expansion of alkali-silica gel, no single
theory about the mechanism of gel expansion is widely
agreed upon, and none appears to have completely and accurately explained the mechanism. For example, none of the
four main theories discussed previously considers the potentially crucial role that short-range forces might play in expansion. Neither the osmotic pressure theory nor double-layer
theory considers the potential effect of calcium on expansion. Although at this time it remains unknown exactly how
alkali-silica gel expands in concrete, portions of the theories
described previously may together explain the mechanism
or lead to a more complete explanation.
5.3—Evaluating aggregates for potential alkaliaggregate reactivity
5.3.1 Field performance—A history of satisfactory field
performance is possibly the best method for evaluating
the potential for an aggregate to cause AAR. A number of
factors have to be considered when analyzing field performance data. These include:
a) The cement content of the concrete and the alkali
content of the cement should be the same or higher in the
field concrete as proposed in the new structure.
b) The concrete examined should be at least 10 years old
and preferably more than 15 to 20 years old.
c) The exposure conditions of the field concrete should be
at least as severe as those in the proposed structure.
d) In the absence of documentation conclusively demonstrating that the aggregate to be used in the proposed structure is sufficiently similar to the field structure under investigation, a petrographic examination should be conducted to
make that determination.
e) The possibility that SCMs, lithium-based admixtures, or
both, were used in the field structure should be considered.
f) Provided that satisfactory field performance can be
demonstrated, the aggregate can be used in concrete,
following the prior-listed guidance, with no further testing
for AAR.
5.3.2 Petrographic examination (ASTM C295/C295M)—
A petrographic examination should be the first step in
assessing the suitability of a particular aggregate source
for use in concrete construction. Petrography is a powerful
tool that yields a wide range of information regarding the
physical, chemical, and mineralogical characteristics of an
aggregate, including the presence of rocks or mineral phases
that are known to cause deleterious reaction in concrete.
In some cases, a petrographic examination may produce
sufficient evidence to reject an aggregate on the basis of
potential alkali reactivity or require that suitable preventive measures be explored and implemented. Generally, the
examination cannot predict whether the type and distribution of reactive minerals present will cause damaging expansion in concrete, and further laboratory testing is usually
required. Results of petrographic examination could form
the basis for directing the laboratory test program in terms
of selecting the type and sequence of tests and any relevant
evaluation criteria.
The reliability of petrographic examination for screening
aggregates for potential reactivity is strongly dependent
on the skill and experience of the individual petrographer.
There have been cases where aggregates that were accepted
for use on the basis of results of petrographic examination
have been later implicated in AAR. This is not necessarily
the result of incorrect material classification, but more likely
is due to a failure to recognize certain minerals as potentially
reactive (Rogers and Hooton 1991). Furthermore, the reactive constituents of some rocks may not be readily identified
by optical microscopy.
5.3.3 Laboratory tests to identify alkali-silica reactive
aggregates—Many test methods have been developed for
identifying alkali-silica reactive aggregates. These methods
vary in terms of testing environment, duration of test, and
reliability of results. A comprehensive review is provided by
Thomas et al. (2006). Generally, the tests of longer duration such as the concrete prism test (5.3.3.4) produce more
reliable results than shorter-duration, highly aggressive tests
such as the accelerated mortar bar test (Thomas et al. 2006).
In this section, a brief description is provided of test methods
to detect ASR in aggregates.
5.3.3.1 Mortar bar test (ASTM C227)—This test method
was originally developed to assess the potential alkali-silica
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reactive of cement-aggregate combinations and is not recommended for evaluating ACR aggregates. Mortar bars (aggregate/cement = 2.25) are monitored for expansion under
storage conditions at 100°F (38°C) in storage containers
designed to maintain high humidity. For testing, in ASTM
C227 coarse aggregates are crushed to a fine aggregate size
fraction and both coarse and fine aggregates follow specific
gradation requirements set forth in ASTM C227. The alkali
content of the cement is not specified, but the cement should
be selected to have the highest alkali content representative
of the cement generally intended for use with the aggregate.
The standard specification for concrete aggregates
(ASTM C33/C33M) regards as reactive those cementaggregate combinations that have an expansion greater than
0.05 percent at 3 months or 0.10 percent at 6 months. More
stringent limits have been applied by certain agencies. For
example, the U.S. Bureau of Reclamation requires specific
combinations to expand by less than 0.05 percent at 6
months, or 0.10 percent at 12 months.
There are a number of limitations associated with this
test. For example, it is difficult to attain high humidity in the
containers without leaching the alkalis from the mortar bars.
Given the relatively small size of these mortar bars (1 x 1 in.
[25 x 25 mm] cross section), leaching may be severe, especially during early testing period. Alkali leaching can lead to
an underestimation of the expansion of certain combinations
of cement and aggregate, especially if the reactive component of the aggregate reacts relatively slowly. This has been
observed in argillites and greywackes in which the reactive component is microcrystalline quartz (Grattan-Bellew
1989; Rogers and Hooton 1991). The consequences of alkali
leaching may be reduced by storing bars over water without
wicks and by raising the equivalent alkali content or Na2Oeq
(note: Na2Oeq = Na2O + 0.658 K2O) of the cement to 1.0
percent, but expansions will still be less than 0.10 percent at
6 months with some reactive aggregates. Due to these limitations, ASTM C227 is not recommended for identifying
alkali-silica reactivity of aggregates or for cement-aggregate
combinations.
5.3.3.2 Quick chemical method (ASTM C289)—In this
test, a sample of the aggregate, crushed to pass a No. 50
(300 μm) sieve and retained on a No. 100 (150 μm) sieve,
is immersed in 1 molar NaOH solution for 24 hours, and
the resulting solution is analyzed. The amount of silica
dissolved and the reduction in alkalinity of the host solution
are plotted on a graph with zones that classify the aggregate
as innocuous, potentially reactive, or deleterious.
The reliability of this test in detecting aggregate reactivity is poor for a number of reasons. Other mineral phases
present in the aggregate may reduce the dissolved silica by
precipitation (Bérubé and Fournier 1992a). Furthermore,
reactive phases may be lost during crushing and sieving.
Thus, reactive aggregates may appear innocuous based on
the test results. In contrast, the high surface area and temperature used in this test dissolve some siliceous mineral phases
that are stable under the conditions that prevail in concrete.
This results in aggregates with good field performance being
23
classified as deleterious (Bérubé and Fournier 1993). Note
that ASTM C289 was withdrawn by ASTM in 2016.
5.3.3.3 Accelerated mortar bar test (ASTM C1260)—This
test is essentially the same as that developed by Oberholster and Davies (1986). In this method, the length change of
mortar bars (measuring nominally 1 x 1 x 11.25 in. [25 x 25 x
285 mm]) stored in 1 molar NaOH solution at 176°F (80°C)
is monitored for 14 days. For testing, in ASTM C1260, coarse
aggregates are crushed to a fine aggregate size fraction and
both coarse and fine aggregates follow specific gradation
requirements set forth in ASTM C1260. Some agencies
specify that length change be measured over 28 days in 1
molar NaOH. The expansions obtained in this rapid test are
generally comparable to or higher than those obtained by
ASTM C227 (100°F [38°C] at 100 percent humidity) after
1 year (Oberholster and Davies 1986; Hooton and Rogers
1989). The test has been successfully used to identify alkalisilica reactive aggregates from across Canada (GrattanBellew 1989; Hooton and Rogers 1989, 1992; Bérubé and
Fournier 1992b; Durand et al. 1990; Hooton 1991) and the
United States (Stark et al. 1993), but is not considered suitable for evaluating reactive alkali-carbonate rocks.
Interpretation of results is not simple, and various expansion criteria have been suggested (Thomas et al. 2007).
Bérubé and Fournier (1992b) proposed a limit of 0.10
percent expansion after 14 days in 1M NaOH for quarried
silicate and siliceous carbonate aggregates, and a limit of
0.20 percent for natural sands and gravels. Many aggregates with satisfactory field performance, however, produce
expansions in excess of 0.25 percent in this test (Bérubé and
Fournier 1992b). Consequently, aggregates should not be
rejected on the basis of this test unless petrographic examination confirms that the material is similar to known deleteriously reactive aggregates.
Recent research has shown that certain aggregate types
may pass this test, having expansions less than 0.10 percent
at 14 or 28 days, yet cause deleterious expansion in ASTM
C1293 and in the field (Ideker et al. 2012). In summary, this
test should be used with caution, owing to the many potential discrepancies between the performance of aggregates in
this test and in the field (Thomas et al. 2007).
5.3.3.4 Concrete prism test (ASTM C1293)—This test is
considered the most reliable for correctly identifying alkalisilica reactive aggregates (Thomas et al. 2006). A concrete
mixture is proportioned with a cement content of 708 ± 17
lb/yd3 (420 ± 10 kg/m3) using a portland cement with an
equivalent alkali content (Na2Oe) of 0.90 ± 0.10 percent.
Sodium hydroxide is then added to the mixing water to
provide a total alkali loading of the concrete of 8.85 lb/yd3
(5.25 kg/m3). This high alkali loading is necessary to induce
expansion of slowly reactive rocks such as greywacke and
argillites (Magni et al. 1987). Prisms are stored at 100°F
(38°C) and the expansion is monitored for at least 1 year.
An expansion limit of 0.04 percent at 1 year is currently
specified in both Canada and France (Thomas et al. 1997)
and is recommended in ASTM C1293. This expansion correlates approximately to the point where cracking and signs of
distress are first observed on the prisms. It also relates well
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24
GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
to field performance. CSA A23.2-27A and CSA A23.2-28A
carry a caveat stating that lower expansion limits may be
appropriate for assessing aggregates for use in critical structures, such as nuclear containment, or dimensionally sensitive structures, such as hydraulic dams, where small expansions may result in relatively large movements.
This test method has distinct advantages over the mortar
bar tests (ASTM C227; ASTM C1260), as coarse aggregates
can be tested without crushing to sand sizes. Furthermore,
the larger test specimen reduces the effect of alkali leaching.
The 12-month duration is necessary unless the temperature
is raised or other changes are made to accelerate expansion;
however, further acceleration of the test may have undesirable side effects (Ideker et al. 2012).
5.3.3.5 Accelerated concrete prism test (RILEM AAR 4 –
proposed)—Due to the duration of the concrete prism test,
there has been a push since the early 1990s to accelerate the test
by increasing the temperature from the standard 100 to 140°F
(38 to 60°C) (Ranc and Debray 1992; Murdock and Blanchette
1994; DeGrosbois and Fontaine 2000; Touma et al. 2001).
Unfortunately, simply raising the temperature and identifying expansion limits was not enough to produce a consistent and reliable test method. Fournier et al. (2004) identified factors leading to the variability in the accelerated
test, including increased leaching of alkalis, reduced pore
solution pH, increased mass loss, and a potential concern
related to selection of the nonreactive fine aggregate; they
also proposed expansion limits. Further work by Ideker et
al. (2010) made clear the profound effect of the selection of
the nonreactive fine aggregate on expansions in the accelerated test: expansions were reduced by as much as 50 percent
compared to the 1-year expansions obtained with the Spratt
coarse aggregate.
The accelerated version of the concrete prism test results
in a reduction in expansion at 3 months as compared to
the 1-year expansion obtained in ASTM C1293 (Ideker et
al. 2010). Thus, the accelerated concrete prism test is not
recommended for assessing aggregate reactivity or for determining the efficacy of mitigation measures.
5.3.3.6 Chinese accelerated concrete microbar test (for
alkali-silica reactive aggregates)—A new test method,
now commonly referred to as the concrete microbar test
(formerly the Chinese accelerated mortar bar method), was
introduced by Xu et al. (1998, 2000) to capture reactivity of
alkali-carbonate reactive rocks. In this test method, mortar
bars measuring 1.58 x 1.58 x 6.30 in. (40 x 40 x 160 mm) are
cast, cured for 24 hours, and then soaked in water at 176°F
(80°C) for an additional 24 hours. The bars are then moved
into a 1 N NaOH at 176°F (80°C) and length change is monitored for 28 days. Aggregates are graded (or crushed where
appropriate) to produce to a size fraction in the range of 0.2
to 0.4 in. (5 to 10 mm) and cast into mortar bars at a fixed
cement-aggregate ratio and a w/cm of 0.33. Since that time,
promising results have also been obtained for the detection
of ASR across a wider range of rock types (Lu et al. 2008;
East 2007). In particular, the ability of this test to detect
deleterious ASR in coarse aggregates is advantageous where
reactive phases are removed due to crushing and processing
in other accelerated methods (ASTM C227; ASTM C1260).
This accelerated test method is gaining momentum, and
other researchers have shown its efficacy in testing potentially alkali-silica reactive coarse aggregates. This method
may serve as a complementary test method, especially when
ASTM C1260 produces a false negative. Lu et al. (2008)
have recommended it as a universal accelerated test for
alkali-aggregate reactivity.
5.3.4 Laboratory tests to identify reactive alkali-carbonate
rock aggregates
5.3.4.1 Rock cylinder method (ASTM C586)—In this
method, cylinders (or prisms) cut from the rock are immersed
in a solution of 1M NaOH at room temperature (after having
attained dimensional stability in distilled water) and the
expansion is monitored for at least 1 month. Expansions in
excess of 0.10 percent at 1 month are generally taken to indicate a potentially deleterious chemical reaction between the
alkalis and the rock. This test does not provide an indication
of the potential for expansion in concrete, and further testing
of the aggregate in concrete (ASTM C1105) is recommended
if the rock cylinder expansion exceeds 0.10 percent.
5.3.4.2 Chemical composition (CSA A23.2-26A)—The
determination of potential ACR by chemical composition involves analysis for CaO, MgO, and Al2O3 (CSA
A23.2-26A). Limestones or dolomites with a composition
outside of the range indicated as potentially alkali-carbonate
reactive in Fig. 6 of CSA A23.2-26A require further testing
for ASR. Potentially reactive dolomitic limestones plot in
the potentially expansive area of a CaO/MgO-versus-Al2O3
plot, and such aggregates should be tested by ASTM C1105.
This test has helped to remove some of the difficulty in
identifying reactive dolomitic limestones by petrographic
examination.
5.3.4.3 Concrete prism test (ASTM C1105)—This test is
similar to ASTM C1293 used to assess ASR (5.3.3.4), except
for differences in storage temperature and the alkali content of
the concrete. ASTM C1105 requires the testing to be carried
out using a specific concrete mixture, with specimens stored
at 73°F (23°C). Potentially deleterious reactivity is indicated
if the expansion exceeds 0.015 percent at 3 months, 0.025
percent at 6 months, or 0.030 percent at 1 year.
Users of this test should recognize that the test yields information about the specific cement-aggregate combination
tested, and that the absence of significant expansion in this
test does not necessarily indicate that the aggregate is nonreactive. For instance, deleterious expansion may occur if the
aggregate is used in concrete with a higher alkali content.
CSA A23.2-14A requires potentially alkali-carbonate reactive rocks to be tested in concrete prisms under the same
conditions as those used for ASR (that is, with 8.85 lb/yd3
[5.25 kg/m3] Na2Oe and stored at 100°F [38°C]). Furthermore, the same expansion criteria are applied; aggregates
are deemed to be reactive if the expansion exceeds 0.040
percent at 12 months, with no criteria for earlier test results.
CSA A23.2-14A is aimed at establishing the reactivity of
the rock and not assessing the performance of a particular
cement-aggregate combination.
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5.3.4.4 Chinese accelerated concrete microbar test (for
alkali-carbonate reactive aggregates)—The Chinese accelerated concrete microbar test was created to capture reactivity of alkali-carbonate rocks (Lu et al. 2008; Xu et al.
1998, 2000; Sommer et al. 2005). Further work by Lu et
al. (2008) has shown the ability of this test to detect alkalicarbonate reactive aggregates across a range of aggregate
gradations (0.2 to 0.4 in. [5.0 to 10.0 mm] particles and 0.1
to 0.2 in. [2.5 to 5.0 mm] particles). The advantage to using a
0.1 to 0.2 in. (2.5 to 5.0 mm) particle size is that the test can
be used to test a single aggregate sample for both ASR and
ACR. Further, petrographic examination would be needed to
distinguish the type of reaction present so that appropriate
mitigation measures could be followed. This test shows
promise for detecting alkali-carbonate reactive aggregates in
a relatively short time of 4 weeks, while providing more reliable results than the accelerated concrete microbar test. This
test is an attractive alternative to the longer-duration ASTM
C1105 test for detecting ACR.
5.4—Preventive measures
There are a number of preventive measures that can be used
to minimize the risk of damage due to ASR, which include:
a) Using nonreactive aggregates
b) Limiting the alkali content of the concrete
c) Incorporating SCMs
d) Using chemical admixtures, namely, lithium compounds
These approaches are discussed in the following sections.
For alkali-carbonate rock reactive aggregates, avoidance or
reduction in proportion to the reactive phases is the only
recommended practice. The other methods listed, though
proven effective with alkali-silica reactive aggregates, are
typically not a remedy for ACR (Rogers and Hooton 1992).
5.4.1 Use of nonreactive aggregate—This approach
is perhaps the most obvious and certain way of avoiding
damaging reaction in concrete structures. Nonreactive aggregates are not available in many locations, and
importing nonreactive material may not be economically
viable. Furthermore, AAR has occurred in a number of cases
where prior testing of the aggregates indicated they were
not deleteriously reactive. Methods of testing aggregates for
reactivity have increased in severity, and acceptance criteria
have become more stringent to reflect the increasing number
of aggregates implicated in field cases of AAR. Adoption
of existing testing practices, however, does not guarantee
that aggregates will give satisfactory performance in every
situation. Consequently, even if aggregates are found not to
be deleteriously reactive, further precautions are frequently
taken as circumstances demand. Such circumstances may
include prestigious (or critical) structures, aggressive environments such as external source of alkalis like seawater or
deicing salts, high cement contents, or extended service life.
5.4.2 Limiting alkali content of concrete—Stanton’s
(1940a,b) work on AAR indicated that expansive reaction
is unlikely to occur when the alkali content of the cement
is below 0.60 percent Na2Oe. This value has become the
accepted maximum limit for cement to be used with reactive aggregates in the United States, and appears in ASTM
25
C150/C150M as an optional limit. This criterion, however,
takes no account of the cement content of the concrete that,
together with the cement alkali content, governs the total
alkali content of concrete and is considered a more accurate
index of potential reactivity. Some national specifications
recognize this fact by specifying a maximum alkali content
in the concrete; this limit was reported (Nixon and Sims
1992) to range from 4.21 to 7.58 lb/yd3 (2.5 to 4.5 kg/m3)
Na2Oe. In some countries, the limit may vary depending on
the reactivity of the aggregate (Oberholster 1994). In CSA
A23.2-27A, the limit ranges between 3.03 to 5.05 lb/yd3 (1.2
and 3.0 kg/m3) Na2Oe. A similar range has been adopted in
Thomas et al. (2008a) and in AASHTO PP065.
The use of low-alkali cement and limitation of the alkali
content in concrete is not a sufficient safeguard in all cases.
Stark (1980) reported damaging AAR in highway structures
constructed using cements with alkalis in the range 0.45 to
0.57 percent Na2Oe. The reactivity of certain aggregates
with low-alkali cements was confirmed in laboratory mortar
bar expansion tests. Lane (1987) reported that some aggregates, classed as innocuous after 6 or 12 months in ASTM
C227 with low-alkali cement (0.54 percent Na2Oe), showed
delayed expansion and cracking after longer periods.
Thomas (1996) reported evidence of ASR in a number of
hydraulic structures with alkali contents below 4.0 lb/yd3
(2.4 kg/m3) Na2Oe.
Aggregates that are not normally reactive when used in
concrete with low-alkali cement may be deleteriously reactive in concrete of higher alkali content. This may occur
through alkali concentration caused by drying gradients,
alkali release from aggregates, or the ingress of alkalis from
external sources such as deicing salts or seawater. Stark
(1978) reported increases in soluble alkali from 1.85 to 6.07
lb/yd3 (1.1 to 3.6 kg/m3) Na2Oe close to the surface of some
highway structures. Migration of alkalis due to moisture,
temperature, and electrical gradients has also been demonstrated in laboratory studies (Nixon et al. 1979; Xu and
Hooton 1993).
There are many aggregates containing alkalis that may
be leached out into the concrete pore solution, thereby
increasing the risk of AAR (Stark 1980; Stark and Bhatty
1986; Way and Cole 1982; van Aardt and Visser 1977;
Thomas et al. 1991; Bérubé et al. 2002). Stark and Bhatty
(1986) reported that in extreme circumstances, some aggregates release alkalis equivalent to 10 percent of the portland
cement content.
Alkalis may penetrate concrete from external sources such
as brackish water, sulfate-bearing groundwater, seawater,
or deicing salts. Nixon et al. (1987) showed that seawater,
or NaCl solutions, present in the mixing water elevates
the hydroxyl-ion concentration and increases the amount
of expansion of concrete. Oberholster (1992) showed that
the expansion of large concrete blocks exposed to saltwater
spray may be doubled compared with the same blocks
exposed to tapwater spray. In addition, studies in Denmark
(Chatterji et al. 1987) have shown that exposure to NaCl
solution and other alkali salts can cause considerable expansion and cracking in concrete.
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
5.4.3 Use of SCMs—In his second published article on
ASR, Stanton (1940b) reported that expansion due to the
reaction could be reduced by the use of pozzolanic cement
containing finely ground shale or by replacement of 25
percent of high-alkali portland cement with pumicite. Subsequent tests by Stanton (1950) confirmed the beneficial effect
of a wide range of natural pozzolans and demonstrated that
partially replacing portland cement with a sufficient quantity of pozzolan (pumicite or calcined shale) eliminated
deleterious expansion, whereas replacement with similar
quantities of ground quartz (Ottawa) sand did not, indicating
that the beneficial action of the pozzolan extended beyond
merely diluting the cement alkalis. The first major use of
a natural pozzolan to control ASR dates back to the 1940s
when calcined siliceous shale, also called Puente shale, was
used in the Davis Dam, together with low-alkali cement
(Gilliland and Moran 1949). In the early 1950s, various
studies (Cox et al. 1950; Barona de la O 1951; Buck et al.
1953) showed that other SCMs (namely, fly ash and slag)
were also effective in reducing expansion. Later research
showed that silica fume was highly efficacious in controlling ASR with levels of 10 percent or less being sufficient
to suppress damaging expansion in mortar bars with reactive aggregate from Iceland (Asgeirsson and Gudmundsson
1979) and South Africa (Oberholster and Westra 1981).
All SCMs contain alkalis and some, like fly ash, may
contain substantially more than the portland cement they
replace. This has led to considerable controversy in the
past regarding whether the alkalis in SCMs are potentially
available for reaction and how they should be treated when
calculating the alkali content of the concrete. Recently,
however, it is generally accepted that the main mechanism
by which SCMs reduce the potential for damaging reaction is by reducing the availability of alkali in the concrete
pore solution. Alkalis released by portland cement and SCM
might be available in one of three forms: dissolved in the
pore solution, bound by the products of hydration, or fixed
by the products of alkali-silica gel. In systems free of reactive aggregate, the partition of the alkalis between the pore
solution and the hydrates is largely a function of the binder
composition (Thomas 2011).
There is a large body of data in the literature that shows
how various SCMs affect the composition of the pore solution extracted from hydrated cement pastes (Longuet 1976;
Diamond and López-Flores 1981; Diamond 1981, 1983a,b;
Glasser and Marr 1985; Canham 1987; Canham et al. 1987;
Kollek et al. 1986; Duchesne and Bérubé 1992, 1994;
Kawamura and Takemoto 1988; Page and Vennesland 1983;
Kawamura et al. 1987; Andersson et al. 1989; Durand et al.
1990; Yilmaz and Glasser 1990; Rasheeduzzafar and Hussain
1991; Shayan et al. 1993; Wiens et al. 1995; Nagataki and Wu
1995; Shehata et al. 1999; Ramlochan et al. 2000; Shehata
and Thomas 2002; Bleszynski 2002; Boddy et al. 2003).
Studies on the effect of fly ash and slag on the pore solution
of pastes have been reviewed by Thomas (1996) and studies
involving silica fume have been reviewed by Thomas and
Bleszynski (2001). These studies show that the incorporation of most SCMs leads to a reduction in the concentration
Fig. 5.4.3a‒–Evolution of the pore solution in pastes
containing SCM (Shehata et al. 1999; Ramlochan et al.
2000; Bleszynski 2002; Shehata and Thomas 2002).
Fig. 5.4.3b–‒Effect of SCM type and replacement level on
the pore solution hydroxyl ion concentration at 2 years
(1: Shehata and Thomas 2002; 2: Shehata et al. 1999; 3:
Bleszynski 2002; 4: Ramlochan et al. 2000).
of alkali hydroxides in the pore solution of pastes, mortar,
and concretes, with the amount of reduction increasing with
higher SCM contents. Figure 5.4.3a shows the evolution of
the hydroxyl ion concentration of the pore solution extracted
from sealed paste samples with w/cm = 0.50, and Fig. 5.4.3b
shows the OH– concentration at 2 years as a function of the
SCM content (from Thomas [2011], using data from Shehata
et al. [1999], Ramlochan et al. [2000], Bleszynski [2002],
and Shehata and Thomas [2002]). The most efficient SCM in
terms of reducing the pore solution alkalinity is silica fume,
followed closely by metkaolin and low-calcium fly ash. Slag
and high-calcium fly ash are less effective and have to be
used at higher cement replacement levels. Thomas (2011)
showed that the concentration of alkali in the pore solution is
a function of the composition of the binder (cement + SCM),
especially its alkali (Na2Oe), calcium (CaO), and silica
contents. Figure 5.4.3c shows a strong correlation between
the hydroxyl ion concentration in the pore solution and the
chemical parameter (Na2Oe·CaO)/(SiO2)2 calculated from
the chemical composition of the binder (cement + SCM).
Thus, SCMs with higher silica contents and lower calcium
and alkali contents will be more efficacious in terms of
controlling the alkali available for reaction.
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The ability of SCMs to reduce pore solution alkalinity is
linked to their effect on the composition and alkali-binding
capacity of the hydrates. The introduction of SCMs reduces
the Ca/Si ratio of the hydrates, which results in more alkali
being bound (Bhatty and Greening 1978; Rayment 1982;
Uchikawa et al. 1989; Thomas et al. 1991). Glasser and Marr
(1985) explain the differences in alkali absorption on the
basis of the surface charge on the C-S-H, which is dependent
on the Ca/Si ratio. At high Ca/Si ratios, the charge is positive and the C-S-H tends to repel cations. As the Ca/Si ratio
decreases, the positive charge decreases, becoming negative at Ca/Si ratios less than 1.3 (Glasser 1992). Negatively
charged C-S-H has an increased capacity to absorb cations,
especially alkalis. Hong and Glasser (1999) confirmed the
importance of the Ca/Si ratio on the alkali-binding capacity
of synthesized single-phase C-S-H, but subsequently showed
that the binding capacity could be greatly increased by introducing alumina into the C-S-H to form C-A-S-H (Hong and
Glasser 2002).
SCMs that are highly efficient at binding alkalis and
reducing the alkali concentration of the pore solution are
also found to be highly effective in controlling expansion of
concrete containing reactive aggregate. Figure 5.4.3d shows
the 2-year expansion of concrete prisms (ASTM C1293) as
a function of the type and amount of SCM used (Thomas
2011). Once again, it was found that the SCMs with the most
silica and the lowest Ca/Si ratio, silica fume, metakaolin,
and low-calcium fly ash are most effective and have to be
used at replacement rates of 10 to 30 percent, whereas slag
and high-calcium fly ash have to be used at higher replacement levels. With the exception of materials with very
high alkali contents, all SCMs can be used to control ASR,
provided that they are used at an adequate level of replacement. The amount of SCM required to control ASR depends
on (Thomas 2011):
a) Composition of SCM: Increasing amounts are required
as the alkali or calcium content of the SCM increase or as the
silica content decreases.
b) Alkali contributed by the portland cement: Generally increased amounts of SCM are required as the alkali
provided by the cement increases.
c) Reactivity of aggregate: The amount of SCM required
increases as the reactivity of the aggregate increases.
In most conditions, the following levels of replacement
are usually sufficient to control expansion due to ASR:
a) Silica fume: 10 to 15 percent
b) Metakaolin: 15 to 20 percent
c) Low-CaO fly ash: 20 to 30 percent
d) Slag: 35 to 50 percent
e) High-CaO fly ash: greater than or equal to 40 percent
As discussed previously, however, the amount of SCM
required should be determined on a case-by-case basis by
appropriate performance testing, or by reference to prescriptive guidelines developed from empirical data. AASHTO
PP065 and ASTM C1778 provide both performance-based
and prescription-based methodologies for determining the
required SCM content.
27
Fig. 5.4.3c–‒Relationship between pore solution composition and the chemical composition of the binder.
Fig. 5.4.3d–‒Effect of SCMs on 2-year expansion of concrete
containing siliceous limestone (1: Shehata et al. 1999; 2:
Shehata and Thomas 2002; 3: Bleszynski et al 2002; 4:
Ramlochan et al. 2000; 5: Thomas and Innis 1998).
5.4.4 Use of chemical admixtures—Chemical admixtures
to inhibit deleterious ASR have not been widely employed
by the construction industry. These include lithium salts,
barium salts, sodium silica fluoride, and alkyl alkoxy silane.
Interest in the use of lithium compounds, specifically lithium
nitrate, has resulted in significant research and testing in both
laboratory and field environments. These studies have investigated the ability for lithium compounds to control ASR in
new concrete construction as well as the potential of lithium
to reduce ongoing ASR in existing ASR-affected concrete
elements. Examples are pavements, highway barriers, and
bridge elements. A brief discussion of the findings regarding
lithium salts follows.
5.4.4.1 Lithium salts—Although McCoy and Caldwell
(1951) reported on the ability of lithium compounds (LiF,
LiCl, and Li2CO3) to control ASR, the use of lithium has not
been adopted by the construction industry, probably due to
its relatively high cost. Interest in the use of lithium has been
shown and a major research project, including field trials,
was conducted by Stark et al. (1993).
Initial work by McCoy and Caldwell (1951) and Lawrence
and Vivian (1961) indicated that a level of Li/(Na + K) molar
ratio of 0.74 was necessary to control ASR. Tremblay et al.
(2007) and Feng et al. (2008), however, have shown that
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
the dosage of lithium required to control ASR varies greatly
and is largely dependent on the aggregate type. Several
researchers have shown that molar ratios of (Li)/(Na + K) in
the range of 0.60 to 1.1 are sufficient to suppress expansion
for many aggregate types (Sakaguchi et al. 1989; Stark et
al. 1993; Tremblay et al. 2007). Tremblay et al. (2007) have
also shown that for certain aggregate types, such as granitic
gneisses and greywackes, doses as high as (Li)/(Na + K) =
1.1 might not be sufficient. Additional caution is advised
because insufficient lithium may actually increase expansion
(Stark et al. 1993). Other research has shown that lithium
salts such as LiOH and LiCO3 are less effective than LiNO3
in reducing or eliminating ASR (Collins et al. 2004). Several
documents provide more detailed review material and guidance for the use of lithium-based admixtures to control ASR
(AASHTO 2000; Folliard et al. 2006).
5.4.4.2 Other chemical admixtures—Other chemical
compounds were found to reduce expansion due to ASR;
these include various barium salts (Hansen 1960), sodium
silicofluoride, and alkyl alkoxy silane (Ohama et al. 1989).
A wide range of compounds was studied by Hudec and Larbi
(1989), but the results were largely inconclusive. Before
such admixtures are recommended for commercial applications, further research is required to confirm their efficacy
and to elucidate their role in controlling ASR.
5.5—Tests for evaluating preventive measures
5.5.1 ASTM C441/C441M—The pyrex mortar bar test, or
ASTM C441/C441M, has been commonly used for evaluating the efficacy of pozzolans and slag in controlling expansion due to ASR. This test method was developed in the
1940s as a method for assessing the suitability of pozzolans
for use in concrete containing reactive aggregate, such as
in connection with specifications for siliceous admixtures
for the Davis Dam (Gilliland and Moran 1949). Early tests
(Buck et al. 1953; Pepper 1964; Blanks 1950) indicated that
fly ash and slag were less effective than highly siliceous
natural pozzolans, and that they should be used in proportions exceeding 40 percent to be effective as defined by
ASTM C441/C441M. Since then, numerous other workers
have used this test to evaluate the performance of pozzolans
and slag.
In ASTM C441/C441M, the 14-day expansion of mortar
bars made with high-alkali cement (0.95 to 1.05 percent
Na2Oe) and 25 percent fly ash by volume, or 50 percent slag
and stored at 100°F (38°C), is compared with that of control
bars (cement only), and the percentage reduction due to the
pozzolan or slag is calculated. Alternatively, the materials
and mixture proportions to be used in the actual job may
be used. ASTM C618 requires that the expansion of the test
mixture, regardless of alkali content of cement used, be no
greater than the expansion of a low-alkali control. SCMs
meeting this requirement are considered to be as effective
as the low-alkali cement control for mitigating ASR. The
percentage of pozzolan used in practice is assumed to be
equal to or greater than that used in the test mixture, and also
assumed that the alkali content of the field cement used will
not exceed that of the test cement by more than 0.05 percent.
Although ASTM C989/C989M, a specification for slag,
does not include a requirement relating to ASR, a nonmandatory appendix suggests the use of ASTM C441/C441M
with 14-day expansion reduced by 75 percent of control or
kept below 0.02 percent when using project materials. Early
versions of the test required slag to be used at replacements
of 20 percent by volume. The criteria used to assess pozzolans or slag in this test have been criticized as too conservative (Klieger and Gebler 1987; Kennerley 1988; Kennerley
et al. 1981; Sturrup et al. 1983).
The potential for silica fume to reduce ASR expansion
is apparent if used at the specified 25 percent by volume
replacement in ASTM C441/C441M, with shrinkage often
being observed after the normal 14-day testing period
(Popovic et al. 1984). Perry and Gillott (1985) used various
silica fume contents and found that 10 percent was effective
in reducing the 14-day expansion by more than 75 percent
compared with control (this was the acceptance criterion at
the time). Other workers have confirmed the ability of 10
percent silica fume (Rasheeduzzafar and Hussain 1991;
Hooton 1993) or less (Bérubé and Duchesne 1992) to meet
this criterion. Perry and Gillott (1985), however, observed
continued expansion of the silica fume specimens beyond
14 days and questioned the reliability of short-term testing
by this method.
Results from tests with borosilicate glass (ASTM C441/
C441M) have shown the effects of fly ash and slag to vary
considerably between studies. The effect of fly ash has been
shown to depend on its alkali content, calcium content,
pozzolanicity, and fineness. The only consensus from the
literature is that the effectiveness of fly ash and slag increases
as the level of replacement increases, and that all fly ashes
and slags can be used to control reaction, provided that they
are used in sufficient quantity. Pepper and Mather (1959)
reported the effectiveness of a pozzolan or slag was related
to fineness, alkali release, and the amount of silica dissolved.
Borosilicate glass is extremely sensitive to test conditions
(surface area, alkali content, and temperature) and contains
significant quantities of alkalis that may be released into
the pore solution. Furthermore, borosilicate glass produces
damaging reaction in just a few days. Consequently, determining the role of pozzolans and slag, particularly their
alkali contributions, is complicated by the use of borosilicate
glass. There has been increased concern over the validity of
ASTM C441/C441M (Hobbs 1989) because the results do
not correlate well with data from concrete tests using natural
aggregates (Bérubé and Duchesne 1992). Generally, the
replacement level required to limit expansion in the borosilicate glass mortar bar test is significantly higher than that
required to limit expansion in concrete containing natural
reactive aggregates; few commercial aggregates are as reactive as borosilicate glass.
ASTM C441/C441M is not recommended to determine
the efficacy of lithium nitrate or other lithium compounds
to control ASR, as it does not consider the reactivity of the
aggregate, which is vital to determine the correct dosage
of lithium. In addition, excessive leaching during the test
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severely limits its ability to reliably predict the efficacy of
lithium nitrate or other lithium compounds.
5.5.2 Accelerated mortar bar test (ASTM C1567)—ASTM
C1567 was adopted in 2004 to assess the ability of SCMs to
control ASR in mortar bars. The test method is essentially
identical to ASTM C1260 (5.3.3.3) with the exception that
a portion of the cement is replaced by the SCMs under test.
The expansion limit of 0.10 percent at 14 days is specified
in ASTM C1567, as it has been correlated with field performance of concrete (Thomas et al. 2006). While several agencies specify different expansion criteria in terms of percent
expansion reached and test measurement age (14 versus
28 days), no additional information has yet been published
to show a stronger correlation to the more reliable ASTM
C1293 or field experience with similar combinations of
aggregates and SCMs.
ASTM C1567 is not suitable for testing the efficacy of
lithium nitrate, as a significant amount of the originally
incorporated LiNO3 is leached to the aggressive 1 N NaOH
host solution. Modified versions of ASTM C1567 for evaluating lithium have been proposed (Tremblay et al. 2010;
AASHTO PP065; USACE CRD-C 662).
5.5.3 Concrete prism test (ASTM C1293)—ASTM C1293
can be used to assess the efficacy of SCMs, chemical admixtures, or both. In this test, a portion of the cement is replaced
by the SCM under evaluation. If a chemical admixture such
as lithium nitrate is to be investigated, a range of dosages
may be necessary to determine the dosage to control ASR
(5.4.4.1). The remainder of the testing follows ASTM C1293
as outlined in 5.3.4.3, with the exception that the duration is
extended to 2 years. Measurements are typically taken at 3or 6-month intervals after the first year of testing following
the standard recommendation for concrete prism testing.
An expansion limit of 0.04 percent at 2 years is specified
by CSA A23.2-27A and is recommended in the appendix
to ASTM C1293. This expansion limit has shown a strong
correlation to field structures cast with similar preventive
measures. This method can also be used to effectively determine the required dosage of lithium salts—namely, lithium
nitrate—to mitigate deleterious ASR.
5.6—Protocols for minimizing the risk of alkaliaggregate reactivity
Numerous protocols have been developed for minimizing
the risk of alkali-aggregate reactivity in concrete. Many
of these essentially take a two-stage approach. First, the
aggregate is evaluated to determine whether it is potentially alkali-silica reactive or alkali-carbonate reactive.
Second, appropriate measures are selected if the aggregate
is alkali-silica reactive; the aggregate is typically rejected
for use in concrete if it is determined to be alkali-carbonate
reactive. In Canada, both prescriptive (CSA A23.2-27A)
and performance (CSA A23.2-28A) approaches are available for selecting preventive measures. In 2010, the Canadian protocol was modified and adopted by AASHTO as
AASHTO PP065. In 2014, AASHTO PP065 was modified
and adopted as ASTM C1778.
29
The development of AASHTO PP065 is described by
Thomas et al. (2008a). A flowchart describing the protocol
for determining aggregate reactivity is presented in Fig. 5.6.
Although aggregates can be accepted solely on the basis of
satisfactory field performance petrographic examination, or
both, the protocol warns that a certain level of risk is assumed
by the owner, as either of these approaches may fail to identify a reactive aggregate. Laboratory expansion testing is
recommended and the preferred test is ASTM C1293, which
is generally believed to be the most reliable for identifying
aggregate reactivity. Because its 1-year duration renders it
impractical in many situations, however, the protocol does
allow aggregate reactivity to be determined using ASTM
C1260, recognizing that it frequently results in false positives
in that it identifies some nonreactive aggregates as reactive,
and occasionally results in false negatives in that it fails to
correctly identify some reactive aggregates. Quarried carbonates are evaluated on the basis of their chemical composition
(MgO, CaO, and Al2O3) to determine the potential for ACR
(CSA A23.2-26A). If the rock is determined to be potentially
alkali-carbonate reactive, it must be tested in concrete, as the
accelerated mortar bar test is not suitable for determining the
risk of ACR. There are three outcomes resulting from the test
protocol shown in Fig. 5.6, and the recommendations for each
outcome are shown in Table 5.6 (CSA A23.2-26A).
Allowable preventive measures include limiting the alkali
content of the concrete, using SCMs, or using lithium-based
admixtures. The level of SCM required can be determined
using either ASTM C1293 or ASTM C1567. The dosage
of lithium can also be determined using ASTM C1293
or a modified version of ASTM C1567. ASTM C1293
is preferred, but due to its 2-year duration, use of ASTM
C1567 is also acceptable.
The practice also contains a protocol for determining the
appropriate alkali limit or the level of SCM using a prescriptive approach that has been developed from empirical data.
The limits (maximum alkali content and minimum SCM
level) are based on the criteria:
a) Aggregate reactivity, based on the amount of expansion
in the concrete prism test or accelerated mortar bar test
b) Exposure condition (availability of moisture and
external alkalis) and size of the element
c) Class of structure, based on the required service life and
the consequences should ASR occur
d) Type and composition of SCM
The maximum alkali content of the concrete varies within
the range from 3 to 5 lb/yd3 (1.8 to 3.0 kg/m3) Na2Oe,
depending on the risk of ASR and the level of prevention
required. Similarly, the minimum SCM content ranges from
15 percent fly ash or 25 percent slag where the risk of ASR
is low and only mild mitigation measures are required, to
35 percent fly ash or 65 percent slag where there is a high
risk of ASR and more stringent measures are required. In
extreme cases—for example, where a critical structure with
a 100-year life is to be built with a highly reactive aggregate
and exposed to alkalis in service—it is necessary to both
limit the alkali content of the concrete and to use high SCM
contents.
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
Fig. 5.6‒–Sequence of test to determine aggregate reactivity (CSA A23.2-27A).
Table 5.6—Testing outcomes and recommended
actions
Outcome of testing
Recommended action
Aggregate is not deleteriously
reactive
Accept aggregate for use – no
prevention required
Aggregate is alkali-carbonate
reactive
Avoid reactive material
Aggregate is alkali-silica reactive
Reject aggregate or select appropriate preventive measures
CHAPTER 6—SULFATE ATTACK
Sulfate attack can take many forms, although it most often
occurs in concrete exposed to external sources of sulfates
(6.1). Less commonly, internal sources of sulfate can also
result in damage, particularly when the concrete is exposed to
excessive temperatures at early ages (6.2.2). The attack can be
in the form of chemical attack on the cement paste or physical
attack due to crystallization of sulfate salts (Chapter 8).
6.1—External sulfate attack
6.1.1 Occurrence—Naturally-occurring sulfates of
sodium, potassium, calcium, or magnesium that can attack
hardened concrete are sometimes found in soil or dissolved
in groundwater adjacent to concrete structures (Table 6.1.1).
These sulfates have their source from ancient seabed deposits
or a breakdown of sulfide or sulfate-bearing minerals. Industrial and agricultural effluents, as well as municipal wastewater, can also supply sulfates. Other sources of sulfates are
water used in concrete cooling towers, where the sulfate ions
gradually build up due to evaporation. Soil fills containing
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industrial waste products, such as slag from iron processing,
can leach sulfate ions.
There are other environments where multiple deterioration mechanisms may be involved. Seawater, brackish
water, and coastal soils constitute a special type of exposure.
Recommendations for these environments are addressed
in 6.3. Portland-cement concrete can also be attacked
by sulfuric acid solutions, which result from oxidation of
sulfur-containing minerals or from decay of organic matter
by bacterial action (Chapter 7).
When water evaporates from a concrete surface, especially
in arid regions, an accumulation of sulfate salts can occur,
resulting in physical salt attack. The general topic of physical salt attack or salt weathering is addressed in Chapter 8.
Sulfate attack has occurred at various locations throughout
the world and is a particular problem in arid areas, such as
the northern Great Plains and parts of the western United
States (Bellport 1968; Harboe 1982; Reading 1975, 1982;
U.S. Bureau of Reclamation 1975; Verbeck 1968); the
prairie provinces of Canada (Hamilton and Handegord 1968;
Hurst 1968; Price and Peterson 1968); and the Middle East
(French and Poole 1976). Other non-arid countries, such as
England (Bessey and Lea 1953) and Norway (Bastiensen et
al. 1957), have also experienced sulfate attack on concrete.
6.1.2 Historical background—Sulfate attack has been
recognized for nearly 250 years. Smeaton (1791), requiring
a material that would harden under water, developed the
first hydraulic lime formulation of the industrial age, the
precursor of modern portland cements. While developing a
cement formulation for use in the construction of the Eddystone Lighthouse, Smeaton described its attack by sulfatecontaining solutions in 1756. During the nineteenth century,
various aspects of sulfate attack of concrete were studied
(Bogue 1955).
Traditionally, sulfate attack has been thought to occur as
a consequence of a sulfate-containing solution entering the
pore structure of concrete and reacting with hydrating cement
compounds such as tricalcium aluminate (C3A) to form various
sulfate-containing phases that adversely affect concrete durability. Bates et al. (1913) state that “It is almost universally
believed that it is the reaction of sulphate of magnesia of
the sea water with the lime of the cement and the alumina
of the aluminates of the cement, resulting in the formation
of hydrated magnesia and calcium sulpho-aluminate, which
crystallizes with a large number of molecules of water.”
This understanding has been the basis for the development
of sulfate-resisting cements, and of mixture proportions for
concretes to be placed in sulfate environments.
While there was an understanding that sulfate attack was
associated with a compound first described by Candlot in
1880, the precise compositions of calcium sulfoaluminate
hydrates were not established until 1929. Lerch et al. (1929)
established the composition of ettringite, known as Candlot’s salt, as 3CaO·Al2O3·3CaSO4·31H2O and that of monosulfate as 3CaO·Al2O3·CaSO4·12H2O. In addition to establishing the compositions of these salts, Lerch et al. (1929)
also showed that monosulfate converts to ettringite when
an external source of sulfate is provided. This appears to be
31
Table 6.1.1––Mineral names and general
composition often used in reports of sulfate attack
Anhydrite
CaSO4
Thenardite
Na2SO4
Bassanite
CaSO4 ≈ 0.5 H2O
Mirabilite
Na2SO4 ≈ 10H2O
Gypsum
CaSO4 ≈ 2H2O
Arcanite
K2SO4
Kieserite
MgSO4 ≈ H2O
Glauberite
Na2Ca(SO4)2
Epsomite
MgSO4 ≈ 7H2O
Langbeinite
K2Mg2(SO4)3
the first citation in the literature that describes this chemical
mechanism of sulfate attack.
Although the precise composition of ettringite was not
established until the work of Lerch et al. (1929), the role of
C3A in sulfate attack had been recognized earlier. Concern
about the C3A content of cements had stimulated a variety
of studies of cement compositions over a period of 20 years
or more. Bates and Klein (1917), who studied the properties of calcium silicates and C3A, reached the conclusion that
it would be impossible to commercially produce a portland
cement containing less than approximately 1 percent alumina
because of the high firing temperatures required. Work in
Canada by Thorvaldsen, however, identified the means to
reduce C3A contents by changing the proportions of C3A and
C4AF while avoiding excessively high kiln temperatures.
Thorvaldsen observed that cements with high iron contents
also exhibited improved sulfate resistance. This led to the
eventual development of Type V cement (Fleming 1933).
Today, low-C3A cements are routinely produced. Sulfate
attack, however, involves phenomena in addition to the
formation of ettringite. Consequently, even so-called zeroC3A cements might not be immune to sulfate attack.
The basis of the most commonly used method for establishing the sulfate resistance of cements by measurement
of expansion should also be credited to Thorvaldson et al.
(1927, 1929). From observations of deterioration due to
warping and expansion of cement-containing materials, such
as tiles, mortar and concrete, they developed an expansion
test, which is the basis of ASTM C1012/C1012M.
Resistance to sulfate attack is increased by controlling
both cement composition and concrete permeability. The
importance of this was demonstrated in studies by Verbeck
(1968) and Stark (1989b) that showed that reduction of
permeability was of greater importance in limiting sulfate
attack than was using a sulfate-resistant cement composition. In a 40-year summary of U.S. Bureau of Reclamation data, it was found that a w/cm of 0.45 or lower helps
in avoiding damage from sodium sulfate attack on portland
cements having C3A contents less than 8 percent (Monteiro
and Kurtis 2003). In some cases, failure was avoided with a
w/cm as high as 0.53, but significant damage can occur in the
w/cm range from 0.45 to 0.53.
An appreciation that limitations on C3A content and
on w/cm are both needed to produce sulfate-resistant
concrete have been embodied in various codes and standards governing the selection of concrete for use in sulfate
environments. The U.S. Bureau of Reclamation (1975) has
formally recognized these requirements since 1949.
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
6.1.3 Mechanisms—Phenomena characterized as external
sulfate attack can occur in a number of ways involving the
formation of a variety of compounds. Origins of sulfates can
be both external and internal, and include the oxidation of
pyrites and generation by bacterial action.
These forms of external sulfate attack have been recognized in Brown (2002):
a) Ettringite (AFt), monosulfate (AFm), and gypsum
formation
b) Sulfate-containing salt formation at or near an evaporative surface
c) Thaumasite formation
Mechanisms for these three forms of external sulfate attack
are described as follows, and publications discussing them
in detail include Lea (1971), Mehta (1976, 1992), Mehta
and Monteiro (2006), DePuy (1994), Taylor (1997), Hewlett
(1998), and Skalny et al. (1998). Publications with particular
emphasis on permeability and the ability of concrete to resist
ingress and movement of water include Reinhardt (1997),
Hearn et al. (2006), and Diamond (1998).
6.1.3.1 Sulfate attack associated with ettringite and
gypsum formation—These forms of sulfate attack occur
when external sulfate-containing solutions infiltrate the
pores of concrete. This increases the concentration of
sulfates in the concrete pore solution available to react with
sources of calcium and alumina to form ettringite and with
sources of calcium to form gypsum. These reactions are also
influenced by pH because ettringite is not stable below a
pH of approximately 10 and gypsum is not stable above a
pH of approximately 10.5 (Taylor 1997). This explains the
common observations in both laboratory studies and in field
concrete of gypsum near exposed concrete surfaces, where
carbonation has lowered the pH, and of ettringite in the interior of gypsum deposits.
In mature concrete, ettringite typically forms directly
from the monosulfate that had formed during the hydration of the cement. Some supplementary cementitious
materials (SCMs) serve as additional sources of reactive
alumina and thus may increase the potential for ettringite
formation. Low-w/cm concrete mixtures containing SCMs,
however, are generally more resistant to sulfate attack due to
a reduced rate of ingress of the sulfate solution and reduced
CH contents.
Ettringite is responsible for internal cracking and expansion; this damage mechanism is described in early studies
of sulfate attack. However, ettringite may also form as
secondary deposits in voids and cracks in mature field
concrete exposed to wetting and drying. Consequently, the
presence of ettringite in concrete does not necessarily indicate sulfate attack (ASTM C856).
External sulfate attack produces two damage mechanisms
in concrete. Cracking due to expansion is probably the most
widely reported form of damage. Expansion occurs because
the volumes of ettringite and gypsum are greater than those
of the reactants from which they form. An increase in the
volume of solid phase in the hardened cementitious matrix
results in tensile stresses due to crystallization pressures,
and cracks develop once the tensile strength of the paste is
locally exceeded. A second damage mechanism associated
with external sulfate attack involves softening and loss of
cohesion. As discussed in the following, this damage mechanism involves chemical alterations that destabilize the C-S-H
and calcium hydroxide (CH), and can result in the formation
of microcracks without significant expansion.
The damage that results from external sulfate attack also
depends on the cation associated with sulfate. The most
common naturally-occurring sulfates that attack concrete
are calcium, sodium, and magnesium sulfate, which are
listed in order of increasing aggressiveness. Calcium sulfate
(gypsum, CaSO4·2H2O) is generally the least aggressive
because its solubility is significantly lower than that of
sodium and magnesium sulfate. Calcium sulfate solutions
can, however, attack concrete (Thorvaldson 1954; Taylor
1997; Drimalis 2007). In addition, after calcium and sulfate
ions enter concrete pores, the high alkalinity of the pore
solution increases their solubility compared to that in natural
waters (Hansen and Pressler 1947), thus allowing the development of higher concentrations of sulfate that can increase
the severity of the attack. The calcium sulfate attack damage
mechanism involves internal expansion and cracking due to
ettringite formation.
A sodium sulfate solution can be more aggressive than
calcium sulfate because sodium sulfate is more soluble.
Consequently, concrete can be exposed to higher sulfate
concentrations. Sodium sulfate attack can lead to the formation of gypsum and ettringite within the cement paste at the
expense of the normal cement hydration products of CH,
monosulfate, hydrated C3As, and in severe cases, the C-S-H
binder. Both cracking and softening are associated with
sodium sulfate. In addition, salt deposits can form on evaporative surfaces of concrete elements subjected to sodium
sulfate attack, causing scaling (physical salt attack). Chapter
8 discusses the mechanisms of salt crystallization and its
associated damage mechanisms in more detail; understand,
though, that the presence of salt deposits on scaled evaporative surfaces may indicate the occurrence of external chemical sulfate attack, physical salt attack, or both.
Although magnesium sulfate and sodium sulfate share
similar solubility, magnesium sulfate attack can be more
damaging because both magnesium and sulfate ions participate in the attack. The reaction products of magnesium sulfate
attack include ettringite, gypsum, magnesium hydroxide, and
a silica gel, and may produce a matrix with very low strength
or binding capacity (Gollop and Taylor 1995; Taylor 1997).
The magnesium ion undergoes a base exchange process
with CH or with C-S-H that forms brucite (magnesium
hydroxide). This lowers the pH of the concrete pore solution and provides a source of calcium to react with sulfate
and produce gypsum. The reactions will continue until they
exhaust the CH and the C-S-H from the paste. Consequently,
softening and loss of cohesion is the end-state damage mechanism associated with magnesium sulfate attack.
6.1.3.2 Physical salt attack by sulfate salts—Deterioration
due to physical salt attack starts at the surface of concrete.
Initially, the deterioration, which has an appearance similar
to scaling caused by freezing and thawing, can be induced
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by any of several different salts, most commonly sodium
sulfate. Damage is due to crystallization pressure from
the precipitation of crystals within the pore structure of
concrete. Physical salt attack is considered a physical form
of attack because damage is not related to chemical interaction between sulfate ions and the hydrated phases of portland cement. Physical salt attack can occur together with
chemical sulfate attack.
If sulfate ions are identified in the soil, groundwater, or as
part of the efflorescence on concrete surfaces, any damage of
concrete by physical salt attack, including damage to architectural surfaces, requires serious consideration and evaluation. Chapter 8 discusses physical salt attack.
6.1.3.3 Thaumasite formation—Although it has a different
chemical composition, thaumasite has a similar crystal structure to ettringite. While it is typical to denote ettringite by
its oxide composition as 3CaO·Al2O3·3CaSO4·32H2O, the
appropriate representation of the molecular structure of ettringite is (Ca3Al(OH)6.12H2O3+)2(SO42–)3·2H2O·Ca3Al(OH)6;
units form four heavily hydrated columns per unit cell.
There are also four interstices between the columns per unit
cell. Three of these are occupied by sulfate anions and the
fourth by two molecules of water. This structure permits a
broad range of substitutions: divalent cations can be substituted for calcium; trivalent cations, such as transition metal
ions, can substitute for aluminum; tetravalent ions, such
as Si4+, can also substitute for Al3+. One such silica-based
compound is thaumasite (CaO.SiO2·CaSO4·CaCO3·15H2O
or CaSiO3·CaCO3·CaSO4·15H2O).
Thaumasite was first identified as occurring in deteriorating concrete in 1965 by Erlin and Stark (1966), and later
by Bickley et al. (1994), and has been extensively investigated by Matthews (1994), the Thaumasite Expert Group
(1999) in the UK, and Crammond (2002a,b). Ettringite and
thaumasite are frequently found together in deteriorating
concrete. Whether they are present as intimate mixtures
or exhibit solid solution behavior when formed under
these conditions has been debated. Erlin and Stark (1966),
however, found lath-shaped crystals where there were petal
overgrowths of thaumasite on ettringite, finely banded crystals of alternating ettringite and thaumasite, and continuous
lath-shaped crystals of which one-half was ettringite and the
other half thaumasite.
Although thaumasite can form in the absence of external
sulfate due to carbonation, which can decompose both ettringite
(releasing sulfates) and the calcium silicate hydrate (C-S-H)
binder (releasing hydrous silica), it is unlikely. External sulfate,
or an inadvertent, gross excess of internal sulfate can also
contribute to thaumasite formation and are almost exclusively associated with its formation. Similarly, if a source of
readily soluble CaCO3 is present within the concrete, thaumasite formation can occur. Thus, thaumasite can form as a
consequence of sulfate ingress, carbonation, or both. While
originally thought to occur only in concretes exposed to cool
temperatures, thaumasite formation has also been observed
in concrete in temperate climates (Crammond 2002a).
Thaumasite preferentially forms under the cold, wet, alkaline conditions typically experienced by buried concrete
33
structures. The occurrence of thaumasite in deteriorated
building materials has been identified in a number of countries worldwide, including the United Kingdom, United
States, Canada, South Africa, France, Germany, Norway,
Denmark, Switzerland, Italy, and Slovenia. Probably the
most severe case of thaumasite-damaged concrete encountered so far was in the Canadian Arctic (Bickley et al. 1994).
The use of sulfate-resisting concrete does not necessarily
prevent the formation of thaumasite, because it is the C-S-H
and not aluminate phases that are attacked by external
sulfates. The replacement of C-S-H by thaumasite transforms the cement paste matrix into a white, soft, noncohesive mass. From the formula for thaumasite, it is seen that
carbonate ions are also necessary.
A positive identification of thaumasite in a cement-based
building material does not automatically indicate that a
problem has occurred or, if it has, that thaumasite was the
cause. There are two distinct ways in which thaumasite
can precipitate as a reaction product within concretes and
mortars (Thaumasite Expert Group 1999) and the following
characteristics should be considered during diagnosis.
6.1.3.3.1 Thaumasite form of sulfate attack is visually
very distinctive, characterized by significant damage to the
cement paste matrix of the concrete or mortar. The main hallmark of thaumasite sulfate attack is that hardened cement
paste becomes partially or totally replaced by thaumasite. As
thaumasite does not possess any binding ability, the affected
cement paste is ultimately transformed into a noncohesive
mass loosely holding the aggregate particles together. Other
distinguishing features include subparallel cracks filled with
thaumasite and white haloes of thaumasite occurring around
aggregate particles. Thaumasite sulfate attack, which causes
gradual softening of the matrix of a buried concrete starting
from the concrete-ground interface and progressing inward,
can sometimes be accompanied by expansive disruption.
6.1.3.3.2 Thaumasite, like ettringite, can precipitate harmlessly in voids and cracks. This phenomenon has been termed
“thaumasite formation” and can be found in concretes or
mortars showing no obvious visual signs of sulfate attack.
Thaumasite formation also occurs in concretes already
damaged by other deterioration mechanisms such as ASR
(French 1986; Regourd and Hornain 1986). Although the
presence of thaumasite is more often innocuous, it can be a
precursor to thaumasite sulfate attack.
Thaumasite sulfate attack is typically associated with
ingress of external sulfates. The carbonate ions necessary for
thaumasite sulfate attack can be supplied by limestone aggregates or limestone in cement, or externally by carbonate or
bicarbonate ions dissolved in sulfate-bearing water. Hooton
and Thomas (2002) considered 5 percent limestone additions
to cement not to be a risk. Thaumasite preferentially forms
at temperatures below 59°F (15°C). Although it can form
at temperatures up to 77°F (25°C), the rate is much slower
(Thaumasite Expert Group 1999; Alksnis and Alksne 1986).
6.1.4 Recommendations
6.1.4.1 Sulfate attack associated with ettringite, monosulfate, and gypsum formation—Protection against the
various forms of sulfate attack is obtained by proportioning
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34
GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
Table 6.1.4.1a—Severity of exposure conditions
determined from sulfates in soil or water
Exposure class
Water-soluble sulfate
(SO42–)* in soil, %
Sulfate (SO42–)* in
water, ppm
S0 (not applicable)
SO42– < 0.10
SO42– < 150
S1 (moderate)
0.10 ≤ SO42– < 0.20
150 ≤ SO42– < 1500,
or seawater
S2 (severe)
0.20 ≤ SO42– ≤ 2.00
1500 ≤ SO42– ≤ 10,000
S3 (very severe)
SO42– > 2.00
SO42– > 10,000
*
Sulfate expressed as SO4 is related to sulfate expressed as SO3, as given in reports of
chemical analysis of portland cements as follows: SO3 × 1.2 = SO4.
concrete mixtures to minimize the ingress and movement of water using appropriate ingredients. The sulfate
resistance of portland cement generally decreases with an
increase in its calculated tricalcium-aluminate (C3A) content
(Mather 1968; Stark 2002). Accordingly, ASTM C150/
C150M includes Type V sulfate-resisting cement for which
a maximum of 5 percent calculated C3A is permitted, and
Type II moderately sulfate-resisting cement for which the
calculated C3A is limited to a maximum of 8 percent. There
is also evidence that the alumina in the aluminoferrite phase
of portland cement can participate in sulfate attack (Brown
et al. 1986). Therefore, ASTM C150/C150M provides that in
Type V cement, the C4AF + 2C3A content should not exceed
25 percent unless the alternate requirement based on the use
of the performance test (ASTM C452/C452M) is invoked.
In the case of Type V cement, the optional sulfate-expansion test (ASTM C452/C452M) can be used in place of the
chemical requirements (Mather 1978). In CAN/CSA A3000,
ASTM C452/C452M expansion limits are used to qualify
both moderate- and high-sulfate-resisting portland cements.
The use of ASTM C1012/C1012M is discussed by Patzias
(1991). ASTM C1157/C1157M, ASTM C595/C595M, and
CAN/CSA A3000 blended cements also use ASTM C1012/
C1012M expansion limits to qualify moderate (MS) or
high (HS) resistant performance. Both ACI 318 and CSA
A23.1-14/CSA A23.2 allow the use of other combinations
of cementing materials in sulfate exposure, provided that
performance testing using ASTM C1012/C1012M demonstrates that the expansion limit for the appropriate exposure
class (Table 6.1.4.1a) is not exceeded. Note that in Table
6.1.4.1b, performance testing is not required if ASTM C150/
C150M Type II cement is selected for S1 exposure and
Type V cement is selected for S2 exposure.
One strategy for reducing the ingress and movement of
dissolved sulfates and water is to lower the w/cm. The use
of acceptable SCMs is another, complementary strategy to
reduce the ingress and movement of water into the concrete
(refer to Chapter 3 regarding limiting fluid ingress). Care
should be taken to ensure that the concrete is designed and
constructed to minimize shrinkage cracking. Proper placement, compaction, finishing, and curing of concrete are
essential to minimize the ingress and movement of fluids
that carry aggressive salts. Recommended procedures for
these are found in ACI 304R, ACI 302.1R, ACI 308R, ACI
305R, ACI 306R, and in Chapter 3 of this guide.
Recommendations for the maximum w/cm and the type
of cementitious material for concrete that will be exposed
to sulfates in soil or groundwater are given in Table 6.1.4.1b
for exposures defined in Table 6.1.4.1a. Both recommendations are important, as limiting only the type of cementitious material is not adequate for satisfactory resistance to
sulfate attack (Kalousek et al. 1976; Stark 2002; Monteiro
and Kurtis 2003).
Table 6.1.4.1b provides recommendations for various
degrees of potential exposure. These recommendations are
designed to protect against distress from sulfate sources
external to the concrete, such as may be in adjacent soil,
groundwater, and effluents carried in concrete pipes.
The field conditions of concrete exposed to sulfate are
numerous and variable. The aggressiveness of the conditions depends on, among other things, soil saturation, water
movement, ambient temperature and humidity, concentration of sulfate, and type of sulfate or combination of sulfates
involved. Table 6.1.4.1b provides criteria that should maximize the service life of concrete subjected to aggressive
sulfate exposure conditions.
6.1.4.2 Physical sulfate attack: physical salt attack by
sulfate salts—Chapter 8 provides recommendations for the
more general case of physical salt attack. In the presence of
sulfates, however, the code requirements for prevention of
chemical sulfate attack must be followed (Tables 6.1.4.1a
and 6.1.4.1b).
6.1.4.3 Thaumasite formation—Use of Types II or V
sulfate-resisting cements does not prevent thaumasite sulfate
attack because thaumasite does not consume aluminate
phases. Ettringite, however, does appear to be a precursor
to thaumasite formation in many cases. Thaumasite attack
is sometimes associated with: 1) excess carbonate fines in
aggregates or cements that are well above the concentrations allowed by ASTM C150/C150M; and 2) carbonation
from air or water exposures. The conditions under which
thaumasite attack occurs are not fully known and, as yet,
there are no standards that address its prevention specifically. Very wet exposure conditions, however, appear to
be common with thaumasite sulfate attack, so provision of
low-w/cm concretes would reduce the ingress of sulfate and
carbonate ions as well as reducing the rate of carbonation
(Hooton 2007). ASTM C1012/C1012M, when modified
such that mortar bars are exposed to the sulfate solution at
40°F (5°C), has been found to be suitable for determining
whether cementitious binders are resistant to thaumasite
sulfate attack (Hooton and Brown 2009; Hooton et al. 2010);
this test was adopted in CSA A3000 in 2010. In addition to
measuring length and mass change, X-ray diffraction can
determine whether thaumasite is present.
The use of slag cement in concrete appears to help resist
thaumasite sulfate attack (Hill et al. 2003). Nobst and Stark
(2003) found that concrete with cement containing at least
66 percent slag cement was resistant to thaumasite. This was
confirmed by Bellmann and Stark (2008) where mortar bars
made with a European CEM IIIBi cement with 65 percent
slag cement was resistant to thaumasite after exposure to
sulfate solutions at 46°F (8°C). They attributed the good
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
35
Table 6.1.4.1b—Requirements to protect against damage to concrete by sulfate attack from external
sources of sulfate
Severity of
potential
exposure
Prescriptive cementitious material requirements
Performance cementitious material requirements
Cement types*
Maximum expansion when tested using ASTM
C1012/C1012M
w/cm by mass,
maximum
ASTM
C150/C150M
ASTM
C595/C595M
ASTM
C1157/C1157M
S0
No w/cm
restriction
No type
restriction
No type restriction
S1
0.50†
Type II‡§
S2
0.45†
S3
0.40†
At 6 months
At 12 months
At 18 months
No type
restriction
—
—
—
IP (MS), IS (<70) (MS),
IT (P<S<70) (MS), or IT
(P≥S) (MS)
MS
0.10%
—
—
Type V#
IP (HS), IS (<70) (HS),
IT (P<S<70) (HS), or IT
(P≥S) (HS)
HS
0.05%
0.10%||
—
Type V plus
pozzolan or
slag cement**
IP (HS), IS (<70) (HS),
IT (P<S<70) (HS), or IT
(P≥S) (HS)
HS††
—
—
0.10%
*
Alternative combinations of cementitious materials to those listed in Table 6.1.4.1b can be permitted when tested for sulfate resistance and meeting the ASTM C1012/C1012M
expansion criteria for the severity of potential exposure.
‡
Other available types of cement, such as ASTM C150/C150M Type I or Type III can be permitted in Exposure Classes S1 if the C3A content is less than 8 percent.
For seawater exposure, other ASTM C150/C150M cement types with C3A contents up to 10 percent are permitted if w/cm does not exceed 0.40. (Refer to Section 6.3 on seawater
exposure.)
§
#
An ASTM C150/C150M Type III cement with the optional limit of 5 percent can be permitted or ASTM C150/C150M cement of any type having expansion at 14 days no greater
than 0.040 percent when tested by ASTM C452/C452M.
||
The 12-month expansion limit can be used if the 6-month limit is not met, but is not required if the 6-month limit is met.
Values applicable to normalweight concrete. They are also applicable to structural lightweight concrete except that the maximum w/cm of 0.50, 0.45, and 0.40 should be replaced
by specified 28-day compressive strengths of 26, 29, and 33 MPa (3750, 4250, and 4750 psi), respectively.
†
As stated in ACI 318, the amount of the specific source of the pozzolan or slag cement to be used shall be at least the amount that has been determined by service record to
improve sulfate resistance when used in concrete containing Type V cement. Alternatively, the amount of the specific source of the pozzolan or slag cement to be used shall be at
least the amount tested in accordance with ASTM C1012 and meeting the criteria shown in the table.
**
††
For Exposure Class S3, ASTM C1157/C1157M HS cement must contain pozzolan cement, slag cement, or both.
resistance to the reduction in CH content of the matrix, and
when CH was added to mortar bars made with the same
cement, damage due to thaumasite sulfate attack occurred.
Fifty percent slag will resist the formation of thaumasite
in mortar bars stored at 40°F (5°C) (Hooton et al. 2010).
Bellmann and Stark (2007) also found that when a CEMIIAL cement, which is portland cement with up to 20 percent
interground limestone, was replaced with either 20 or 40
percent Class F fly ash, there was no damage after 4.5 years
of storage in 1500 mg/L SO4= (sodium sulfate) solution at
46°F (8°C).
6.1.5 Sampling and testing to determine potential sulfate
exposure—To assess the severity of the potential exposure
of concrete to detrimental amounts of sulfate, representative samples should be obtained of both the fluid and sulfate
compound(s) that might reach the concrete or of soil that
might be leached by water moving to the concrete. The
procedure for making a water extract of soil samples for
sulfate analysis that is given in ASTM C1580 is recommended (Hayes 2007). Although other methods have been
used, the results are often affected by the test method, especially the extraction ratios.
6.1.6 Establishing equivalent performance for cementitious materials—The use of alternative combinations of
cementitious materials to those listed in Table 6.1.4.1b is
permitted for any class of exposure. Any binary or ternary
blend of portland cement of any type meeting ASTM C150/
C150M, ASTM C595/C595M, or ASTM C1157/C1157M
with fly ash or natural pozzolan meeting ASTM C618, silica
fume meeting ASTM C1240, or slag cement meeting ASTM
C989/C989M is permitted if it meets the expansion limits
in Table 6.1.4.1b when tested in accordance with ASTM
C1012/C1012M.
The portland-cement portion of the test mixture should
always consist of cement with Bogue-calculated C3A content
of not less than that being proposed for use. Material qualification tests using the expansion limits in Table 6.1.4.1b
should be based on passing results from two samples taken
at times a few weeks apart. The qualifying test data should
be no older than 1 year from the date of test completion.
6.1.7 Proportions and uniformity of pozzolans and slag
cement—The proportion of fly ash, natural pozzolan, silica
fume, or slag cement used in the project mixture (in relation
to the amount of portland cement) should be the same as
that used in the test mixture prepared to meet the recommendations of Table 6.1.4.1b and Section 6.1.6. In blends
or mixtures with portland cement containing only one SCM,
such as fly ash, natural pozzolan, silica fume, or slag cement,
the proportion of fly ash or natural pozzolan can generally
be expected to range between 15 and 50 percent by mass
of the total cementitious material, depending on the severity
of exposure. Similarly, the proportion of silica fume can be
expected to range between 5 and 12 percent by mass of the
total cementitious material, and the proportion of cement
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36
GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
can be expected to range between 35 and 70 percent by
mass of the total cementitious material. When more than one
supplementary cementitious material is used, the individual
proportion of each may be less than these values.
The uniformity of the fly ash or slag cement used in the
project should be within the following limits compared to
that used in the mixtures tested to meet the recommendations of Table 6.1.4.1b and Section 6.1.6:
a) Fly ash: reported calcium-oxide content (analyzed in
accordance with ASTM C114) no more than 2.0 percentage
points higher than that of the fly ash used in the test mixture
b) Slag cement: reported aluminum-oxide content
(analyzed in accordance with ASTM C114) no more than 2.0
percentage points higher than that of the slag cement used in
the test mixture.
The portland cement used in the project should have a
Bogue-calculated C3A content no higher than that used in
the mixtures tested to meet the recommendations of Table
6.1.4.1b and Section 6.1.6.
Studies have shown that some pozzolans and slag cements
used, either in blended cement or added separately to the
concrete in the mixer, increase the life expectancy of concrete
considerably in sulfate exposure. Many slag cements and
pozzolans significantly reduce the permeability of concrete
(Bakker 1980; Mehta 1981). They also combine with the
alkalis and CH released during the hydration of the cement
(Vanden Bosch 1980; Roy and Idorn 1982; Idorn and Roy
1986), reducing the potential for gypsum formation (Lea
1971; Biczok 1972; Kalousek et al. 1972; Mehta 1976).
Table 6.1.4.1b requires a suitable pozzolan or slag cement
along with Type V cement or equivalent in S3 exposures.
Research indicates that some pozzolans and slag cements
are effective in improving the sulfate resistance of concrete
made with Type I and Type II cement; this option is allowed
if the 18-month ASTM C1012/C1012M expansion limit in
Table 6.1.4.1b is met. Some pozzolans, especially Class C
fly ashes, decrease the sulfate resistance of mortars in which
they are used (Mather 1981b; Mather 1982). Good results
were obtained when the pozzolan was a fly ash meeting the
requirements of ASTM C618 Class F (Dikeou 1975; Dunstan
1976). Slag cement should meet ASTM C989/C989M and
silica fume should meet ASTM C1240.
In concrete that is made with non-sulfate-resisting
cements, calcium chloride reduces resistance to attack by
sulfate (U.S. Bureau of Reclamation 1975) and, therefore, its
use should be prohibited in concrete exposed to sulfate (S-1
or greater exposure). If Type V cement is used, however, it
is not harmful to use calcium chloride in normally acceptable amounts as an accelerating admixture (Mather 1992).
Calcium chloride, however, can induce and accelerate corrosion of reinforcing steel and aluminum conduit.
6.2—Internal sulfate attack
6.2.1 Concrete materials—Cementitious materials
meeting current ASTM specifications will not have sulfate
contents that are deleterious to concrete. Allowable sulfate
contents in cements meeting ASTM C150/C150M were
increased several times from 1941 to 1971 (Hooton 2008), as
cement compositions and finenesses changed to allow better
optimization of sulfate contents. ASTM C150/C150M now
allows SO3 limits to be exceeded if it can be demonstrated
(typically using ASTM C563) that the optimum SO3 content
is above the stated limit. In this case, ASTM C1038/C1038M
must show that the SO3 content of the cement will not result
in adverse expansions. These results are considered satisfactory when a 14-day expansion limit of 0.020 percent is specified. This test and expansion limit has also been adopted in
ASTM C1157/C1157M and is used for Canadian portland
cements, blended cements, and combinations of cementitious materials in CSA A3001 (Hooton and Brown 2009).
This test and expansion limit has not been adopted in ASTM
C595/C595M because it still relies on ASTM C265.
Although ASTM C33/C33M does not limit sulfate content
in aggregates for use in concrete, they should not contain
appreciable levels of sulfate-bearing minerals, such as
calcium sulfate inclusions, or be contaminated with sulfates
or sulfides such as pyrite. Limits are placed on sulfides in
air-cooled, blast-furnace slag aggregate. The sulfides in
slag aggregate include iron sulfide and calcium sulfide.
Aggregate sampling and testing for sulfate content should
be completed in advance of use. Sulfate concentrations in
mixing water are not normally deleterious to concrete and
should meet the limits in ASTM C1602/C1602M. Sulfates in
chemical admixtures meeting ASTM C260/C260M, ASTM
C494/C494M, and ASTM C1017/C1017M will not be deleterious to concrete.
6.2.2 Delayed ettringite formation (DEF)
6.2.2.1 Occurrence—Under certain conditions, heatcured concrete elements can suffer expansion and cracking
on subsequent exposure to moisture. This form of deterioration has been commonly referred to as delayed-ettringite
formation (DEF). The normal early formation of ettringite
that occurs in concrete cured at ambient temperature can be
delayed as a result of exposure to excessive temperatures
during manufacture. The ettringite then forms at later ages
when the concrete is exposed to moisture in service. This
delayed formation of ettringite can lead to internal expansion
and damage in hardened concrete. In the 1990s, it appeared
that heat-cured railway ties were particularly susceptible to
this form of deterioration with cases involving ties being
reported in Germany, Finland, former Czechoslovakia,
Canada, the United States, South Africa, and Australia (Heinz
and Ludwig 1987; Tepponen and Eriksson 1987; Vitousova
1991; Mielenz et al. 1995; Oberholster et al. 1992; Shayan
and Quick 1992). In most, if not all, of the alleged cases of
DEF, other mechanisms of deterioration, especially alkalisilica reaction (ASR) and freezing-and-thawing damage,
have also been implicated, making it difficult to identify the
precise role of DEF in the deterioration.
The occurrence of these problems and apparent role of
elevated-temperature curing has led many countries to
impose restrictions on heat curing of precast concrete. These
restrictions include limits placed on preset times, rates of
heating and cooling, and maximum temperature. There is
evidence that these and similar practices in Europe have
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
been successful in eliminating damage due to DEF (Skalny
and Locher 1997; Stark and Bollmann 1999).
Note that the risk of damage due to DEF is not restricted
to heat-cured precast concrete elements. Internal concrete
temperatures may increase sufficiently, due to the heat
released during the hydration of the cementitious component
of the concrete, to promote DEF in non-heat-cured precast
concrete or in massive cast-in-place concrete elements
(Thomas et al. 2008b). Indeed, the new European practices
impose similar limits on the maximum internal concrete
temperature for both precast and cast-in-place concrete.
6.2.2.2 Mechanisms—As reviewed by Thomas and Skalny
(2006), the solubility of ettringite increases with temperature
and pH, and elevated temperatures increase both the sulfate
and alumina concentrations in the pore solution of concrete.
Much of this sulfate and alumina is taken in by the rapidly
forming calcium-silicate hydrates (C-S-H). Immediately
after the early exposure to elevated temperature, little or no
ettringite is detected in the concrete, and poorly crystalline
monosulfate appears to be the main sulfate-bearing phase
(Taylor et al. 2001). During subsequent exposure to moisture at ambient temperatures, most of the sulfate, but only
a small amount of the alumina, is released by the C-S-H.
The increased availability of sulfate results in a conversion
of the monosulfate into ettringite, which could occur many
months or years following the exposure to elevated temperatures. This delayed formation of ettringite may, under some
circumstances, result in expansion of the paste and consequent cracking of concrete. The cement paste expansion is
generally believed to be a result of the growth of ettringite
crystals in the very small pores (approximately 100 nm) of
the cement paste (Taylor et al. 2001; Lawrence 1995). The
expansion of cement paste results in the formation of gaps
around aggregate particles and cracking of the cement paste
(Johansen et al. 1993). Ettringite eventually reprecipitates
into these gaps and the cracks, but this is not a cause of
damage (Johansen et al. 1993).
Reviews of laboratory studies (Day 1992; Thomas 2001)
and recent data (Thomas et al. 2008b) indicate that expansion due to DEF is unlikely to occur unless mortar or
concrete specimens are subjected to elevated temperatures
in excess of approximately 160°F (70°C) and that the risk of
expansion increases with increasing temperature above this
threshold value. For mortar or concrete that has been exposed
to higher temperatures, the risk of expansion appears to be
a function of many parameters, including both physical
and chemical characteristics of the cementitious binder
(Shimada 2005; Shimada et al. 2007). A review of published
results (Thomas 2001) appears to indicate that high-fineness
cements produced from clinker with high concentrations of
C3A, C3S, and Na2Oeq, and consequently having a high SO3
content, have the greatest susceptibility to DEF expansion
when heat cured. Kelham (1996) showed a clear correlation
between the 2-day compressive strength of mortar and the
expansion of the same mortar subsequent to heat curing at
194°F (90°C). Although studies have shown relationships
between DEF expansion and various compositional parameters of the cement, there is no single parameter that can
37
be used to reliably predict the performance of a particular
cement. Thus, it is not possible to impose a single limit on
the chemical composition of the cement to eliminate the risk
of expansion in concrete that may be exposed to excessive
temperature during curing. However, it is apparent from
laboratory studies that portland cements having high C3A,
C3S, Na2Oeq, and SO3 contents generally have the highest
propensity for expansion when cured at high temperatures.
Cements that exhibit lower early-age strength development
generally present a lower risk of expansion (Kelham 1996;
Ramlochan 2002).
The risk of expansion of heat-cured mortars and concrete
can be effectively eliminated by the incorporation of enough
of the appropriate SCMs (Ghorab et al. 1980; Ramlochan et
al. 2003). Silica fume, when used as the sole SCM, reduces
but does not fully mitigate DEF, apparently due to the lack of
aluminates in the hydrates (Ramlochan et al. 2003). This is
the reason that the recommendations shown in Table 6.2.2.2
require silica fume to be used in a ternary system with fly ash
or slag cement.
6.2.2.3 Recommendations—To minimize the risk of
poor durability due to deleterious DEF reactions associated with exposure to elevated temperatures at early ages,
the maximum internal temperature of concrete should be
controlled such that it does not exceed 158°F (70oC) at any
time. If temperatures in the range of 158°F < T ≤ 185°F (70
to 85°C) are unavoidable, the measures in Table 6.2.2.2
should be adopted.
6.3—Seawater and brine exposure
6.3.1 Occurrence—Seawater throughout the world varies
in the concentration of total salts. The proportions of the
constituents of seawater salts, however, are essentially
constant. More concentrated brines are contained in some
land-locked bodies of water, such as the Great Salt Lake and
the Dead Sea. Brackish water is also aggressive to reinforced
concrete.
The concentration is lower in the colder and temperate
regions than in warm seas and is especially high in shallow
coastal areas with high evaporation rates. The concentration
of salts in land-locked seas also depends on the amount of
fresh water flowing in from rivers (Hewlett 1998).
Where concrete structures are placed on reclaimed coastal
areas with the foundations below saline groundwater levels,
capillarity and evaporation may cause supersaturation and
crystallization of salts in the concrete above ground, resulting
both in chemical and physical attack on concrete from
sulfates, and in aggravated corrosion of steel from accompanying chlorides. Where concrete structures are immersed in
sea water, the portions above water are usually affected the
most by sulfate attack, both physical and chemical, while
portions that are totally immersed often suffer considerably
less damage (Hewlett 1998).
These combined deleterious effects can cause severe
defects in concrete in the course of a very few years, especially in tropical climates where high temperature increases
the rate of deterioration. This section focuses on the influ-
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38
GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
Table 6.2.2.2—Recommended measures for reducing potential for DEF in concrete exposed to elevated
temperatures at early ages*
Maximum concrete temperature T
*
Prevention required
T ≤ 158°F (70°C)
No prevention required
158°F < T ≤ 185°F
(70°C < T ≤ 85°C)
Use one of the following approaches to minimize the risk of expansion:
1. Portland cement meeting requirements of ASTM C150/C150M moderate or high sulfate-resisting and low-alkali
cement with a fineness value less than or equal to 430 m2/kg
2. Portland cement with a 1-day mortar strength (ASTM C109/C109M) less than or equal to 2850 psi (20 MPa)
3. Any ASTM C150/C150M portland cement in combination with the following proportions of pozzolan or slag
cement:
a) Greater than or equal to 25 percent fly ash meeting the requirements of ASTM C618 for Class F fly ash
b) Greater than or equal to 35 percent fly ash meeting the requirements of ASTM C618 for Class C fly ash
c) Greater than or equal to 35percent slag cement meeting the requirements of ASTM C989/C989M
d) Greater than or equal to 5 percent silica fume (meeting ASTM C1240) in combination with at least 25 percent
slag cement
e) Greater than or equal to 5 percent silica fume (meeting ASTM C1240) in combination with at least 20 percent
Class F fly ash
f) Greater than or equal to 10 percent metakaolin meeting ASTM C618
4. An ASTM C595/C595M or ASTM C1157/C1157M blended hydraulic cement with the same pozzolan or slag
cement content as listed in Item 3
T > 185°F (85°C)
The internal concrete temperature should not exceed 185°F (85°C) under any circumstances.
Assembled from Ghorab et al. (1980), Ramlochan et al. (2003), Thomas (2001), Thomas et al. (2008b).
ence of sulfates in seawater. Section 7.2 describes the influence of chloride and magnesium ions in seawater.
6.3.2 Mechanisms—The reaction of mature concrete with
sulfate ions in seawater is similar to that with sulfate ions in
fresh water or leached from soils, but the effects are different
(Mather 1966). Concrete in seawater often exhibits erosion,
softening, or loss in mass as a result of sulfate attack as
opposed to the expansion, which may also occur in nonsaline sulfate environments.
The presence of chloride ions, however, alters the extent
and nature of the chemical reaction so that less expansion
is produced by a cement of given Bogue-calculated C3A
content than would be expected of the same cement in a
freshwater exposure where the water has the same sulfate
ion content. To an extent, this can be explained by the ability
of chlorides to bind with C3A in the cement to form chloroaluminates, such as Friedel’s salt (Verbeck 1975). Formation
of chloroaluminates does not result in undesirable expansion, and it also lowers the amount of C3A available to react,
reducing the damage caused by sulfate attack. In the tidal
and splash zones, however, the concentration of sulfate and
chloride ions in concrete can be increased by capillary action
and evaporation.
It has been suggested that the magnesium sulfate in
seawater is primarily responsible for the chemical reactions occurring in concrete. Because CH and calcium sulfate
are both more soluble in seawater than in fresh water, they
are more easily removed by leaching, while magnesium
sulfate forms gypsum, silica gel, and magnesium hydroxide
(Hewlett 1998). Sulfate attack in seawater can also lead
to decalcification of C-S-H. The formation of magnesium
hydroxide in concrete pores can act as a barrier to the ingress
of sulfate ions, but this effect is not as pronounced in more
permeable concrete (Hewlett 1998; Santhanam et al. 2006).
6.3.3 Recommendations relative to seawater or brine—
The rate of deterioration depends on the concentration of
aggressive ions, duration of exposure, and permeability and
chemical resistance of concrete. Low permeability plays an
important role in hindering the ingress of aggressive ions
in seawater or brine. As with conventional sulfate attack,
permeability is more important than chemical composition
of cement in avoiding damage from seawater exposure.
Recommended w/cm and cement types for seawater exposure are provided in Table 6.1.4.1b under S1 exposure.
The performance of concretes continuously immersed in
seawater made with ASTM C150/C150M cements having
C3A contents as high as 10 percent has proven satisfactory,
provided the permeability of the concrete is low (Browne
1980). The U.S. Army Corps of Engineers (USACE) (1984)
and the Portland Cement Association recommend up to
10 percent Bogue-calculated C3A for concrete that will
be permanently submerged in seawater if the w/cm is kept
below 0.45 by mass.
CSA A23.1-14/CSA A23.2 allows w/cm of up to 0.50 and
recommends 8 percent maximum C3A content. With reinforced concrete construction exposed to seawater, however,
the maximum w/cm would be limited to 0.40 due to the
additional chloride exposure. In addition, low C3A content
decreases resistance to chloride penetration; therefore, CSA
A23.1-14/CSA A23.2 suggests that C3A contents of portland
cement be kept between 4 and 8 percent to help protect the
reinforcement by increasing chloride binding.
Verbeck (1968) and Regourd et al. (1980) showed,
however, that there may be a considerable difference
between the calculated and the measured phase composition
of cement, especially as far as C3A and C4AF are concerned.
Therefore, the interrelation between the measured C3A
content and seawater resistance may be equally uncertain.
The requirement for low permeability is essential not
only to delay the effects of sulfate attack but also to afford
adequate protection to reinforcement with the minimum
concrete cover as recommended by ACI 357.1R for expo-
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39
Table 7.1a—Effect of common chemicals on concrete*
Rate of attack at
ambient temperature
Rapid
Inorganic acids
Organic acids
Alkaline solutions
Salt solutions
Miscellaneous
Hydrochloric Nitric
Sulfuric
Acetic
Formic
Lactic
—
Aluminum chloride
—
Bromine (gas)
Sulfite liquor
Moderate
Phosphoric
Tannic
Sodium hydroxide
greater than 20 percent
Ammonium nitrate
Ammonium sulfate
Sodium sulfate
Magnesium sulfate
Calcium sulfate
Slow
Carbonic
—
Sodium† hydroxide
10 to 20 percent
sodium hypochlorite
Ammonium chloride
Magnesium chloride
Sodium cyanide
Chlorine (gas)
Seawater
Soft water
—
Oxalic
Tartaric
Sodium hydroxide
Less than 10 percent
sodium hypochlorite
ammonium hydroxide
Calcium chloride
Sodium chloride
Zinc nitrate
Sodium chromate
Ammonia (liquid)
†
Negligible
*
Refer to Portland Cement Association (2001) for a more complete list of chemicals and their potential effects on concrete.
†
The effect of potassium hydroxide is similar to that of sodium hydroxide.
sure to seawater. The required low permeability is attained
by using concrete with a low w/cm and is well consolidated
and adequately cured. ACI 357.1R recommends a maximum
w/cm of 0.45 for the submerged zone and 0.40 for the splash
zone.
The permeability of concrete made with appropriate
amounts of suitable slag cement or pozzolan can be as low as
1/10 or 1/100 that of comparable concrete of equal strength
made only with portland cement (Bakker 1980). The satisfactory performance of concretes containing slag cement in
a marine environment has been described (Lea 1971; Vanden
Bosch 1980; Mather 1981a).
Concrete should be designed and constructed to minimize
crack widths, therefore limiting chloride penetration to reinforcement and avoiding the concentration of sulfates. Additionally, concrete should reach a maturity equivalent of not
less than 5000 psi (35 MPa) at 28 days when fully exposed
to seawater.
CHAPTER 7—CHEMICAL ATTACK
7.1—General
Concrete is rarely attacked by chemicals in their solid
form. To produce a significant attack on concrete, aggressive
chemicals must be in solution and above some minimum
threshold concentration to drive the chemical reactions
that diminish its engineering properties. Although concrete
may perform satisfactorily in a variety of exposure conditions where aggressive chemicals are present, some kinds of
chemical environments will significantly shorten the service
life of even the best concrete unless specific measures are
taken.
An understanding of these exposure conditions permits
measures to be taken to prevent or slow deterioration. Lower
permeability will hinder the infiltration of aggressive chemicals. Less-reactive paste can be effective in mitigating deterioration. Concrete members exposed to aggressive solutions
that are under hydraulic pressure from one side may be more
vulnerable because the hydraulic gradient can accelerate the
infiltration of the aggressive solution into the concrete.
This chapter discusses aggressive chemical exposures,
including: seawater, acids, fresh water, carbonation, industrial chemicals, deicing chemicals, and environmental
structures. Some useful summaries of the potential effects
of chemical exposures include Lea (1971), Biczok (1967),
Scrivener and Young (1997), Hewlett (1998), Eglinton
(1998), Portland Cement Association (2001), ACI 515.2R,
and ACI 350.1. Table 7.1a summarizes the effects of the
more common chemicals that lead to the deterioration of
concrete. Table 7.1b summarizes factors that may affect the
rate of chemical attack.
7.2—Seawater
7.2.1 Occurrence—Seawater contains dissolved salts that
are potentially aggressive to concrete. The major chemical
components include, in approximate order of decreasing
concentration: chloride, sodium, sulfate, magnesium,
calcium, and potassium. The concentration of total salts
in seawater varies; warmer climates generally have higher
concentrations. The severity of marine exposures can vary
greatly within a given concrete structure. In general, continuous submersion is the least aggressive exposure. Areas
where capillary suction and evaporation are prevalent are
the most aggressive because these processes tend to increase
the concentration of salts. Examples of such exposures
include reclaimed coastal areas with foundations below
saline groundwater level, intertidal zones, and splash zones.
This section mainly focuses on the influence of chloride and
magnesium ions in seawater. Section 6.3 describes the influence of sulfates in seawater.
7.2.2 Reaction mechanisms—The chemical reactions
that affect concrete exposed to seawater can be complex.
The most significant aggressive species include magnesium, sulfate, and chloride. Magnesium ions may react with
calcium hydroxide (CH) and form magnesium hydroxide
or brucite (Lea 1971). This phase is highly insoluble and
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
Table 7.1b—Factors influencing chemical attack on concrete
Factors that accelerate or aggravate attack
Factors that mitigate or delay attack
1. High permeability due to:
a) High water absorption
b) High w/cm
c) Poor consolidation
d) Poor curing
e) Cracking and microcracking
1. Low-permeability concrete* achieved by:
a) Proper mixture proportioning†
b) Reduced unit water content
c) Increased cementitious material content
d) Appropriate use of supplementary cementitious materials (SCMs)
e) Air entrainment
f) Adequate consolidation
g) Effective curing‡
2. Cracks and separations due to:
a) Loading/stress concentrations
b) Thermal stress
c) Shrinkage
2. Reduced tensile stress in concrete by:#
a) Using tensile reinforcement of adequate size, correctly located
b) Inclusion of pozzolan to reduce temperature rise
c) Provision of adequate contraction joints
d) Effective curing
3. Leaching and liquid penetration due to:
a) Flowing liquid§
b) Ponding
c) Hydraulic pressure
3. Structural design
a) Minimize areas of contact and turbulence
b) Provision of membranes and protective-barrier system(s)||
c) Provision of adequate drainage and through-flow
*
Factors that control permeability are discussed in more detail in Chapter 3.
†
The mixture proportions and initial mixing and processing of fresh concrete determine its homogeneity and density.
‡
Poor curing procedures result in flaws and cracks.
#
Resistance to cracking depends on strength and strain capacity.
§
Movement of water-carrying deleterious substances increases reactions that depend on both quantity and velocity of flow.
||
Concrete that will be frequently exposed to chemicals known to produce rapid deterioration should be protected with a chemically-resistant protective-barrier system.
may fill pores on the outer surface of the concrete, which can
reduce permeability and limit further infiltration of aggressive
waters. However, brucite is not always observed in seawater
exposures. Magnesium can react with any hydrate normally
present in concrete, including calcium silicate hydrate (C-S-H)
and hydrated magnesium silicate (M-S-H) phases, forming
phases such as hydrotalcite. Such phases can be detrimental
to concrete because they can reduce the binding capacity of
the cement paste. Sulfates in seawater may lead to the formation of secondary reaction products typically associated with
sulfate attack such as ettringite, gypsum, and thaumasite.
Chlorides may react with calcium aluminate phases to form
chloroaluminate phases such as Kuzel’s salt and Freidel’s salt.
Chlorides are also a particular concern because of the potential for corrosion of embedded steel.
Mather (1966) noted that reactions of mature concrete with
sulfates in seawater is similar to that with sulfates in fresh
water or leached from soils, but the effects differ because
the presence of chloride ions alters the extent and nature
of the chemical reaction. Less expansion is found with a
cement of a given C3A content than would be expected of
the same cement in a nonmarine exposure where the water
has the same sulfate ion content but lacks significant chloride concentration. Concretes made with portland cements
having C3A contents as high as 10 percent and subject to
continuous immersion in seawater have performed satisfactorily, provided that the permeability of the concrete is
low (Browne 1980). USACE EM 1110-2-2000:1994 and
the Portland Cement Association recommend up to 10
percent calculated C3A for concrete that will be permanently
submerged in seawater if the w/cm is kept below 0.45 by
mass. Verbeck (1968) and Regourd et al. (1980) showed that
there may be a considerable difference between the calculated and measured clinker composition of cement, particu-
larly with respect to the proportions of C3A and C3AF. Therefore, the interrelation between the measured C3A content and
the resistance to seawater may be uncertain. CSA A23.1-14/
CSA A23.2 recommends the use of cements with between 4
and 8 percent C3A unless slag or pozzolans are used.
7.2.3 Mitigation—The mitigation of seawater exposure requires the use of mixture proportions that minimize
permeability and tendency to microcracking, structural
designs that minimize the number and width of cracks, and
possibly the application of coatings that provide cathodic
protection or reduce permeability. The requirement for low
permeability is essential not only to delay the effects of
chemical attack on the cementitious phases in the concrete,
but also to afford adequate protection to reinforcement with
the minimum concrete cover recommended by ACI 357.1R
for exposure to seawater. Steps to achieve low-permeability
concrete are discussed in more detail in Chapter 3. Factors
such as w/cm, types of cementitious materials, appropriate
use of pozzolans and slag, aggregate grading, good consolidation, and adequate curing are important to achieve satisfactory performance.
The use of slag cement and silica fume may reduce the
permeability and increase the performance of concrete in
a marine environment (Lea 1971; Fidjestøl and Frearson
1994; Mather 1981a; Vanden Bosch 1980). The permeability
of concrete made with appropriate amounts of suitable slag
cement or pozzolan can be lowered by orders of magnitude
compared to concrete of equal strength made with portland
cement only (Bakker 1980). Concretes made with combinations of cement and silica fume, and with combinations of
cement, slag cement, and silica fume also have lower permeability and good performance in seawater exposure.
Concrete should be designed and constructed to minimize
the length, width, and number of cracks to limit seawater
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
access to the reinforcement. Additionally, concretes should
achieve an in-place strength of at least 4000 psi (27.5 MPa)
before being exposed to seawater. Marine structures often
involve thick sections and rather high cement factors. Such
concrete may need to be treated as mass concrete in which
the effect of the heat of hydration is considered. When this is
the case, the recommendations of ACI 207.1R, ACI 207.2R,
and ACI 224R apply.
Conductive coatings applied at the time of construction
as part of a cathodic protection system may provide additional protection for concrete exposed to saline groundwater.
Coatings that significantly restrict the evaporation of free
water from the interior of concrete can reduce resistance to
freezing and thawing.
7.3—Acid attack
In general, portland-cement concrete does not have good
resistance to acids, although some weak acids can be tolerated, particularly if the exposure is occasional.
7.3.1 Occurrence—The products of combustion of many
fossil fuels contain gases that can combine with moisture
to form acids. Sewage that is not sufficiently aerated can
form sulfuric acid (Flemming 1995; Sydney et al. 1996).
Acids may occur in runoff from some mines and in some
industrial waters. Peat soils, clay soils, and alum shales may
contain sulfide-bearing minerals such as pyrite that produce
sulfuric acid on oxidation. Further reactions can produce
sulfate salts, which may lead to sulfate attack (Hagerman
and Roosaar 1955; Lossing 1966; Bastiensen et al. 1957;
Mourn and Rosenquist 1959). Some mineral waters may
contain high concentrations of dissolved carbon dioxide and
hydrogen sulfide; such solutions can be acidic and highly
aggressive to concrete (RILEM 1962; Thornton 1978).
Organic acids may come from farm silage or from manufacturing and processing facilities such as breweries, dairies,
canneries, and wood-pulp mills. Animal feed and manure
may also contain acids that corrode concrete (De Belie et
al. 1996). This can be of considerable concern in the case of
floors, even where structural integrity is not impaired.
7.3.2 Reaction mechanisms—The deterioration of
concrete by acids is primarily the result of decomposition of
the hydration products of the cementitious paste. Different
cement hydration products start to decompose at different pH
values. Portlandite (CH, Ca(OH)2) is the most soluble hydration product. At room temperature, portlandite decomposes
at pH below 12.4, ettringite decomposes at pH below 10.4
(Warren and Reardon 1994), and C-S-H starts to decompose
when pH drops to around 10 (Beaudoin and Brown 1992).
Aggregates made from limestone and dolomitic rocks are
susceptible to acid attack, while most siliceous aggregates
are resistant to acids.
The degree to which an acid is aggressive toward concrete
depends on the type of anion, its concentration, and its degree
of dissociation in the solution (Zivica and Bajza 2001). For
a given pH, acetic acid is more aggressive than nitric acid
(Shi and Stegemann 2000; Shi 2003; Bakharev et al. 2003).
Oxalic and phosphoric acids are less aggressive, primarily
because they react with concrete to form precipitates of
41
calcium phosphate and calcium oxalate, respectively. These
deposits are insoluble in water and tend to coat the concrete
surface, protecting against further deterioration. Exposure to
sulfuric acid may lead to rapid deterioration because of its
low pH. In addition, these reactions may produce calcium
sulfate, which may then drive sulfate attack of adjacent
concrete that was unaffected by the initial acid attack.
The decomposition of C-S-H by acid attack will typically produce a silica gel that has little binding capacity. The
decomposition products resulting from the decalcification
of the original C-S-H have low solubility and can provide
some protection from further corrosion (Shi and Stegemann
2000; Shi 2003). If the original C-S-H has a high calcium/
silica ratio, more calcium will dissolve and the concrete will
corrode more quickly. Shi and Stegemann (2000) found that
a lime-fly ash paste corroded more slowly than portland
cement paste, although the former was more porous.
7.3.2.1 Carbonic acid attack—Carbon dioxide can
dissolve in rain to form carbonic acid, which may then enter
the ground. The decay of organic matter liberates carbon
dioxide that may also form carbonic acid in groundwater. The
concentration of carbonic acid in groundwater can become
high enough to attack concrete. Calcium carbonate dissolves
in the presence of carbonic acid to form free calcium (Ca2+)
and bicarbonate (HCO3–) ions. Detailed discussions of
carbonate equilibria in natural systems can be found elsewhere (Stumm and Morgan 1995; Krauskopf and Bird 1995;
Butler 1998). If the alkalinity of the soil is high enough, the
soil will neutralize or buffer the carbonic acid component of
the water, preventing carbonic acid attack of the concrete. If
the acid is not neutralized, it can attack concrete to varying
degrees, ranging from mild to significant. This type of attack
has been referred to as aggressive CO2 attack in the literature
when, in fact, it was not CO2 but carbonic acid attack. Test
criteria relating to carbonic acid attack have used the term
“aggressive CO2” (reported as milligrams of CO2 per liter)
for what was really carbonic acid in sufficient concentration
to attack concrete. Good discussions of this topic may be
found in Lea (1971) and Hewlett (1998).
Waters that have potentially harmful concentrations of
carbonic acid tend to have pH values ranging from approximately neutral to slightly acidic. Because the rate of attack
depends on both the properties of the concrete and concentration of the carbonic acid, neither the pH nor the amount
of free CO2 in water is a reliable indicator of the degree
of potential harm. In addition, there is no consensus as to
the limiting values of pH or CO2 concentration in water, in
part because of widely varying conditions in underground
construction. Studies have shown that water containing more
than 20 mg/L of carbonic acid (reported as mg/L of CO2) can
result in rapid carbonation and attack of the hydrated cement
paste. However, freely moving waters with 10 mg/L or less
of carbonic acid (reported as mg/L of CO2) can also result
in significant carbonation (Terzaghi 1948, 1949; Hewlett
1998). DIN 4030 includes both criteria and a test method
for assessing the potential of damage from carbonic-acidbearing water.
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
7.3.3 Mitigation—The use of pozzolanic materials such
as fly ash and silica fume will decrease the CH content and
increase the resistance of concrete to acids (Sellevold and
Nilson 1987). In all cases, however, exposure time to acids
should be minimized, if possible, and immersion should be
avoided. No hydraulic-cement concrete, regardless of its
composition, will long withstand highly acid water (pH of 3
or lower). Such cases typically call for the use of an appropriate protective-barrier system.
7.4—Fresh water
Fresh water refers to aqueous solutions with nearly neutral
pH, very low ionic strength, and low dissolved solids content.
7.4.1 Occurrence—Fresh waters include rainwater; waters
in most streams, rivers, and lakes; and domestic water that
is chlorinated and fluorinated. Fresh waters can also occur in
industrial, manufacturing, and other facilities where distilled
waters are produced or used in various processes.
In nature, lightning produces weak nitrous, nitric, sulfurous, and sulfuric acids in natural waters that can cause some
surface deterioration of concrete, especially in areas that
experience frequent thunderstorms. Some fresh waters may
be somewhat acidic due to exposure to acid rain, or they may
contain small concentrations of sulfates, nitrates, and other
salts that, in higher concentrations, could attack concrete.
Significant chemical attack by fresh water, however, is
virtually unreported. That concrete is not significantly deteriorated by fresh water is evidenced by highways, culverts,
pipes, and buildings that are built with the full expectation
that their function will not be significantly affected by such
exposure during their expected service life.
7.4.2 Reaction mechanisms—Very pure waters are aggressive because they are undersaturated with respect to the CH,
C-S-H, and calcium carbonate components of the cementitious paste. In general, the reaction mechanisms associated with freshwater exposures are similar to those of acid
attack because the attack involves preferential dissolution
or leaching of soluble cement hydration products. Consequently, constant replenishment of fresh water may accelerate deterioration and flowing water may be aggressive as
well. Falling water from devices such as gutters and downspouts may be particularly aggressive because it introduces
a physical or erosive component to the attack.
Freshly cast concrete is highly alkaline and its surfaces
may be affected by exposure to fresh water. Most surfaces
carbonate readily when exposed to air, however, rendering
them largely stable. Unfortunately, many publications report
laboratory studies of fresh concrete samples with exposure
to chemicals before significant carbonation. Such studies are
often inapplicable to field concrete. Some fresh water that is
undersaturated with respect to carbonates may attack carbonated concrete. Even under decades of such exposure, the deterioration of carbonated concrete is typically quite slow.
7.4.3 Mitigation—Strategies to minimize the effects of
fresh water exposure include minimizing permeability and
reducing the portlandite content of the cement paste. Design
considerations that provide adequate drainage, limit replenishment, and shelter against falling water are also important.
7.5—Carbonation
Carbonation occurs when hydrated cementitious
compounds react with atmospheric carbon dioxide or
carbonate ions in solution. The pore structure of concrete
largely determines the rate and depth to which carbonation
occurs (Bier 1987). Carbonation begins at the exposed
surface of concrete to form an outer layer of carbonatebearing compounds, reducing the porosity of the surface
(Parrott 1987). The reduction in porosity is directly related
to the conversion of CH to calcium carbonate, thus resulting
in an 11 percent increase in solid volume as compared with
the initial volume of CH within the upper layer (Bier 1987).
Prolonged moist curing may delay carbonation (Parrott
1987). Carbonation can be either beneficial or harmful,
depending on the age of concrete and the environment.
Although carbonation can improve the strength of concrete
and decrease permeability, it can also increase the rate of
corrosion of steel reinforcement.
7.5.1 Occurrence
7.5.1.1 Carbonation of fresh concrete—The carbonation of fresh concrete occurs typically from exposure to
atmospheric carbon dioxide during the hardening process.
Carbonation may also occur due to use of unvented heaters
or the exhaust fumes of equipment. The result may be excessive surface cracking or a weak powdery residue (laitance)
on the surface. Severe carbonation prior to final set can result
in a less wear-resistant surface.
7.5.1.2 Carbonation of early-age concrete—Immature
concrete is more susceptible to carbonation than mature
concrete because its matrix has not hydrated sufficiently
to limit permeability. Carbonation will, therefore, progress faster in the early history of a given concrete member.
Initially, CH will react with carbon dioxide to form calcium
carbonate; tetracalcium aluminate hydrate will react with
carbon dioxide to form monocarboaluminate hydrate (Bier
et al. 1988).
7.5.1.3 Carbonation of mature concrete—Aged concrete is
carbonated to some degree. Carbonation of concrete occurs
at exposed surfaces. More-permeable, coarse, open-finished
concrete, and certain environments, however, can increase
the rate and depth of carbonation. Continued carbonation
can decrease the pH of the cementitious matrix, which leads
to an increased rate of corrosion of reinforcement. Refer to
ACI 222R for detailed discussion on corrosion.
7.5.2 Reaction mechanisms
7.5.2.1 Atmospheric carbonation—The reaction of hydrated
portland cement with atmospheric carbon dioxide is generally
slow. The rate is highly dependent on the relative humidity of
the environment, temperature, permeability of the concrete,
and the concentration of CO2. Keeping all other factors equal,
carbonation occurs most rapidly when the relative humidity of
the concrete is between 50 and 75 percent. Below 25 percent
relative humidity, the degree of carbonation that takes place is
insignificant, and above 75 percent relative humidity, moisture in the pores restricts CO2 penetration (Verbeck 1958).
7.5.2.2 Carbonation by carbon species in water—Carbonation can also take place when concrete is exposed to water
containing sufficient concentrations of carbonate or bicar-
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bonate ions. Carbon species are found in most water. Carbon
species enter water when carbon dioxide in the air dissolves
in rain, as organic matter decays in soil, or as certain minerals
containing carbonates are weathered. Carbonate (CO32–) and
bicarbonate (HCO3–) ions are usually the most abundant
carbon species found in natural waters. Detailed discussions
of carbonate equilibria in natural systems can be found in
Stumm and Morgan (1995), Krauskopf and Bird (1995), and
Butler (1998).
7.5.2.2.1 Alkalinity of water (carbonate alkalinity buffer)—
Carbonate alkalinity can play a critical role in the service life
of concrete. In general, when carbonate alkalinity, which is a
concentration of carbonate and bicarbonate ions, is sufficient
to neutralize or buffer the acidic component of water, both
acid attack and leaching may be prevented. Leaching attack
can occur when water is consistently low in carbonate alkalinity, causing the selective dissolution of carbonate-bearing
compounds of cement. This leaching attack, referred to as
aggressive CO2 attack in past literature (Biczok 1967; Lea
1971; Ibrahim et al. 1997), is better termed “low-carbonate
alkalinity attack”.
Where free CO2 attack, or carbonic acid in solution,
relates to acidic attack that dissolves concrete from the
surface inward (4.3), low-carbonate alkalinity attack relates
to leaching attack from within concrete. Simply stated, lowcarbonate alkalinity is an imbalance in water concerning the
lack of carbonate alkalinity needed to neutralize or buffer
the acidic component, and low-carbonate alkalinity attack,
or the term aggressive CO2 attack, is the leaching of the
cementitious carbonate-bearing compounds until an equilibrium has been reestablished. This guide does not detail this
complicated chemistry, but rather explains the importance of
carbonate alkalinity in most placement environments where
water is in direct contact with concrete. Discussions of the
carbonate alkalinity buffer system in concrete are found in
Lea (1971), Biczok (1967), and Hewlett (1998).
7.6—Industrial chemicals
Industrial chemicals may attack concrete in different
ways, depending on the type or classification of the chemical, its concentration, the duration of exposure, and interactions with the components in the concrete. Industrial
processes usually result in exposure conditions ranging from
incidental to continuous, and even extreme, when industrial
process factors such as high temperatures, high humidity,
and equipment vibration exacerbate the situation.
7.6.1 Occurrence—Industrial chemicals may include
acids, bases, alkalis, corrosives, oxidizers, combustibles,
flammables, explosives, cryogenic, and other processspecific conditions or combinations of these conditions.
Refer to PCA (2001) and ACI 515.2R for summaries of the
effects of many industrial chemicals on concrete.
Concrete in industrial settings may experience direct
exposure to chemicals, indirect exposure to chemicals, and
exposures where abrasion and erosion are a concern. Direct
exposures occur in structures such as those handling cooling
water and primary containment structures. Abrasion and
erosion must be considered in structures subject to moving
43
liquids. Foundations, equipment supports, and structural
framing may experience direct contact with chemicals but
are more frequently subject to indirect chemical exposure.
Indirect exposure includes fumes or vapors and precipitants
from these sources.
Anhydrous chemicals may or may not be aggressive
to concrete in their pure form but can become aggressive
when exposed to moisture, such as humidity in the air. Upon
contact with humidity, these compounds may deliquesce and
then infiltrate the surrounding concrete.
7.6.2 Reaction mechanisms—Because industrial chemicals range widely from highly alkaline to very acidic, a
detailed discussion of the potential reaction mechanisms is
beyond the scope of this guide. Biczok (1967) gives useful
background information on a wide range of these chemicals.
Alkaline solutions have a pH greater than 7. Because
concrete is highly alkaline itself, the interaction of alkaline
chemicals from industrial processes may not affect the durability of the concrete directly unless there is a component
or characteristic in the concrete structure that would react
with the penetrating chemical. An example is the corrosive
action of salt solutions on reinforcing steel in the concrete.
Leaching by more neutral solutions can increase the porosity
of concrete, making it more susceptible to freezing-andthawing damage, spalling, and further attack. Precipitant
deposits could plug air entrainment spaces, making the
concrete more susceptible to freezing-and-thawing damage
and spalling. Acids and corrosives do not penetrate the
concrete to any appreciable depth but react on contact with
the concrete surface via the mechanisms described in 7.3.
7.6.3 Mitigation—All the factors that apply to creating
durable concrete apply to durability against chemical attack.
Permeability of the concrete can be decreased by supplementary cementitious materials (SCMs) in the design mixture
in conjunction with good curing. Reinforcing steel protection can be achieved by several methods, including sealing
the concrete, removing any surface contamination from the
reinforcement, improving the bond between the reinforcement and the paste, using epoxy-coated reinforcement,
and increasing the minimum cover over the reinforcement.
When increasing the reinforcement, the architect/engineer
must consider that crack widths will likely increase. Chapter
3 discusses these topics in detail.
There are other options available to the design engineer
and the contractor that will increase concrete durability. Joint
fillers, joint sealers, waterstops, and surface sealers that are
resistant to and compatible with the chemicals that the structure is expected to be exposed to should be specified. Everything should be installed or applied in full compliance with
manufacturers’ recommendations. Wet curing is the recommended method to achieve the most durable concrete. Curing
agents, if used, must be compatible with the sealer or surface
coating that will be applied. Using a shrinkage-compensating
concrete in accordance with ACI 223R can improve durability,
particularly in a chemical environment. Because shrinkagecompensating concrete minimizes shrinkage cracking and the
number of contraction joints, it reduces the potential leakage
paths in an industrial environment.
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7.7—Deicing and anti-icing chemicals
The use of chemical agents has a long history in helping
maintain safe winter driving conditions. Such maintenance
generally involves two different strategies: deicing and antiicing. Deicing refers to the removal of ice after deposition on
a pavement, whereas anti-icing refers to the prior application
of chemicals to prevent the adherence of ice to the pavement. Historically, most durability problems associated with
deicers were linked to physical processes that exacerbate
scaling, rather than chemical attack. However, deicers may
be associated with significant deterioration from various
chemical attack mechanisms.
7.7.1 Occurrence—Deicers fall into two broad groups:
chloride-based and non-chloride-based. Chloride-based
solutions have seen more widespread use historically.
They include sodium chloride (NaCl), calcium chloride
(CaCl2), magnesium chloride (MgCl2), and many commercially available products comprising combinations of these
salts. Some chloride-based deicers may contain significant
concentrations of other chemicals. Other product formulations include MgCl2-based agricultural products. The use
of MgCl2 deicers has increased significantly since 2000,
particularly in the western United States. In part, this growth
resulted because MgCl2 is more effective at lower temperatures than NaCl and CaCl2.
The introduction of nonchloride deicers stemmed from
concerns of corrosion of reinforcing steel and environmental
impacts on vegetation, and because they are also more effective at lower temperatures than chloride-based deicers. The
principal non-chloride deicers include calcium-magnesium
acetate; urea; glycols consisting of ethylene, propylene, and
diethylene glycols; and alkali-acetates and alkali-formates.
Calcium-magnesium acetate deicers were introduced in the
late 1970s. Their use has become more widespread with
decreasing production costs. Urea was traditionally used
for airfield pavements but is less commonly used now due
to environmental concerns. Ethylene and propylene-based
glycols are in widespread use, primarily for deicing and antiicing of aircraft. They are also occasionally used for maintenance of airfield pavements, sometimes in combination with
urea. A new generation of alkali-acetate and alkali-formate
deicers and anti-icers emerged in the late 1980s and early
1990s to replace urea in deicing airfield pavements and to
alleviate concerns associated with the toxicity of ethylene
glycol. These deicers and anti-icers include potassium
acetate, sodium acetate, potassium formate, and sodium
formate, and are widely used.
7.7.2 Reaction mechanisms—Reaction mechanisms
linked to deicing chemicals are complex and form an active
area of ongoing research. Although initially regarded as
benign for concrete, NaCl is now understood to drive reactions that can lead to portlandite dissolution. The dissolution of portlandite may increase the porosity of the concrete
and lower the pH of the pore solution, which may destabilize the C-S-H phase. Chloride-based deicers may drive
the formation of complex calcium chloroaluminate phases
such as Freidel’s salt. Numerous studies show that CaCl2 is
aggressive to concrete (Collepardi et al. 1994). Among other
mechanisms, CaCl2 deicers drive reactions that may form
hydrated calcium oxychloride phases (Brown and Bothe
2004). The generation of hydraulic pressures from these
reactions may be disruptive to the cementitious paste. Chloride may also accelerate alkali-silica reaction (ASR) under
some conditions (Chatterji et al. 1986). MgCl2 deicers are
linked to the formation of brucite, which is not damaging,
and magnesium silicate hydrate phases (M-S-H) that form
at the expense of C-S-H. The formation of M-S-H, which
is not cementitious, can produce significant deterioration in
pavements by cracking, delamination, and, ultimately, disintegration. Disruptive oxychlorides have also been found in
mortars exposed to MgCl2 (Julio-Betancourt and Hooton
2005; Sutter et al. 2006).
Calcium-magnesium acetate may be among the most
aggressive deicers in terms of chemical attack. Some authors
(Peterson 1995; Santagata and Collepardi 2000) report that
exposure to calcium-magnesium acetate deicer solutions
significantly degrades the cement matrix, resulting in loss
of mass and compressive strength. The reaction mechanisms are similar to those of MgCl2 deicers—dissolution
and leaching of portlandite, destabilization of C-S-H and
formation of M-S-H, and precipitation of brucite and calcite.
Calcium-magnesium acetate deicers are linked to scaling,
cracking, and loss of mass and compressive strength.
Premature deterioration of some airfield concrete pavements
exposed to alkali acetate and sodium formate deicers has
caused concern. Research is ongoing to investigate potential
links between these deicers and the durability of concrete.
Rangaraju et al. (2005) suggested that alkali-acetate and
alkali-formate deicers may cause deleterious expansions due
to ASR in test specimens.
Chemical deicers may also contribute to the relatively rapid
deterioration of joints in pavements and exterior flatwork.
The damage manifests as cracking and spalling parallel to
joints that may be most severe at joint intersections and in the
wheel path. Some states in the northern United States have
observed premature deterioration at pavement joints (Taylor
et al. 2012). The relationship between chemical deicers and
accelerated joint deterioration is an area of active research
and several mechanisms are proposed to explain this deterioration. In addition to depressing the freezing point of water,
deicing and anti-icing chemicals induce fundamental changes
in the physical and chemical properties of solutions that fill
joints, such as their viscosity, surface tension, and sorption
(Spragg et al. 2011; Villani et al. 2014a). These changes
result in a higher degree of saturation and marked increase
in the frequency of cracking and microcracking events under
certain temperature cycling conditions (Villani et al. 2014b;
Farnam et al. 2014). Concrete exposed to high concentrations of chemical deicers tends to have air voids filled with
secondary deposits that include ettringite, portlandite, and
oxychloride minerals (Sutter et al. 2006; Peterson et al.
2013). Oxychloride phases are commonly observed in laboratory-based studies, but documenting their presence in field
concrete remains difficult due to their instability (Peterson et
al. 2013). Ettringite deposits may accelerate the saturation of
concrete (Stark and Bollmann 1999) and possibly diminish
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the efficacy of air void systems, making the concrete more
susceptible to damage during freezing-and-thawing cycles.
Deicing salts may also preferentially dissolve calcium-based
compounds at low temperatures in the interfacial transition
zone of coarse aggregate particles exposed in the saw cut
(Zhang and Taylor 2012). Some cracking has been detected
in concrete exposed in the laboratory to deicer solutions
at temperatures above freezing (Farnam et al. 2014). This
cracking may occur as a result of mineralogical changes in
the cement paste that result from reactions involving deicing
solutions, although the specific reaction mechanisms remain
uncertain at this time.
7.7.3 Mitigation—Mitigation of the effects of chemical
deicers includes steps described previously to minimize
the permeability and control the reactivity of the concrete.
Good curing is especially important in mitigating the effects
of deicers. Care must be taken to avoid exposure to deicers
during the first year of service. Mitigation also requires the
provision and maintenance of adequate drainage to minimize duration of exposure.
7.8—Environmental structures
Environmental structures are designed to contain liquids
and gases. Environmental structures include water treatment
plants; domestic and industrial wastewater treatment plants;
storage tanks and reservoirs; water and wastewater pump
stations; conduits, sewers, manholes, and junction chambers;
and hazardous materials containment structures defined in
ACI 350.2R. When the concentration of the contained chemicals is sufficient, they can attack the concrete. The chemicals that are contained may be bulk process chemicals or
those that exist in the liquid or gas contained in the environmental structures. The attack can be more aggressive at
higher temperatures.
7.8.1 Occurrence—Aggressive chemicals typically
occur in bulk storage containers and in pumping systems.
Secondary containment structures must keep spilled chemicals from reaching the soil. Chemical attack of the concrete
may occur at the point of injection of the chemicals into the
process system, particularly if the injection is toward the
concrete rather than into the process liquid.
Chemical agents in environmental structures range widely
from relatively benign to highly aggressive in terms of their
ability to attack concrete. ACI 350 classifies chemical agents
in three broad categories. Refer to R4.5.1.4 of ACI 350.1
for a more complete discussion. Group 1 chemicals are not
considered to be directly harmful to concrete but may be a
concern if they combine with other chemicals that can react
with concrete; Group 2 chemicals such as activated carbon
and potassium permanganate may stain concrete; and Group
3 chemicals are corrosive to concrete. ACI 350.1 differentiates Group 3 chemicals according to the rate at which they
will corrode concrete under typical exposure conditions
associated with environmental structures. Group 3a chemicals have a slow rate of corrosion, Group 3b chemicals have
a moderate rate of corrosion, and Group 3c chemicals have a
rapid rate of corrosion.
45
7.8.2 Reaction mechanisms—The reaction mechanisms
that attend environmental structures range widely, as there
are a multitude of different types of exposure conditions.
In general, the reaction mechanisms of many exposures
are similar to many of the mechanisms discussed in other
sections of this chapter. Environmental structures exposed
to chemicals with low pH are subject to leaching and corrosion of the concrete due to the dissolution of cementitious
components, and possibly aggregates, in the concrete. Some
aggressive chemicals may react with cementitious phases
to stain the concrete. Water treatment plants may expose
concretes to relatively fresh water. Some structures may
encounter exposure to solutions that have very low alkalinity.
Wastewater treatment plants and facilities may expose
concrete to both chemical and biological activity, resulting in
highly aggressive exposure conditions. Bacterial action under
anaerobic conditions may lead to the generation of hydrogen
sulfide gas. Aerobic bacteria present on wet sewage facility
walls, such as pipe linings, may convert hydrogen sulfide into
sulfuric acid. In these environments, significant deterioration
of the concrete can occur from the dissolution of cementitious phases such as portlandite and C-S-H, the deposition of
secondary products such as gypsum, and leaching of ferruginous and calcareous components in the aggregates. Under
other conditions, solutions in wastewater treatment plants
lead to the deposition of struvite, which is a magnesium
ammonium phosphate mineral (MgNH4PO4·6H2O). Struvite
deposits can impede the efficiency of treatment processes and
cause maintenance problems.
7.8.3 Mitigation—The process design of environmental
concrete structures can affect the potential for chemical
attack of the concrete. ACI 350 establishes the minimum
requirements for the design of concrete environmental structures, including specific durability and protection requirements. Chemical attack in environmental structures may be
controlled by using a process and structural design that does
not increase the corrosiveness of the liquid being processed;
use of a properly designed concrete mixture for the expected
service conditions; use of properly selected concrete materials; and use of good concreting procedures related to
handling, placing, and curing. Even when these procedures
are followed, it may still be necessary to provide a protective
barrier, especially when the chemical that may be in contact
with the concrete can cause unacceptably rapid deterioration
for the expected time the concrete must contain the chemical. Refer to ACI 350 for requirements concerning the use
of protective barriers.
CHAPTER 8—PHYSICAL SALT ATTACK
8.1—Introduction
Physical salt attack is a deterioration mechanism caused
by crystallization of salts in pores near concrete evaporative surfaces. The mechanism is called physical salt attack
because chemical reactions between concrete and crystallizing salt are not involved (Mehta and Monteiro 2006).
Deterioration ranges from very fine surface crumbling and
scaling, which is primarily cosmetic, to severe progressive
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disintegration. Deterioration occurs at evaporative surfaces
above the soil line or on flatwork where water may wick
(Fig. 8.1a and 8.1b). The most common salts linked to
physical salt attack include, in order of decreasing aggressiveness, sodium sulfate, sodium carbonate, and sodium
chloride.
Physical salt attack has also been called salt crystallization, salt hydration distress, salt damp, and salt weathering.
Reports identifying physical salt attack include Bates et al.
(1913), Wig et al. (1917), Reading (1975, 1982), Novak
and Colville (1989), Yen and Bright (1990), Haynes et
al. (1996), Hime et al. (2001), and Erlin and Jana (2003).
Physical salt attack on concrete began to receive attention
in the early 1990s, with research initiated by Folliard and
Sandberg (1994) and theoretical developments by Scherer
(1999). Prior research generally focused on chemical sulfate
attack on concrete, while physical sulfate attack was often
overlooked or misidentified.
Physical salt attack also occurs as a deterioration mechanism on exposed stone and brick masonry, as well as other
porous materials. This deterioration mechanism is an important component of the broader phenomenon known as salt
weathering, where differential expansion and contraction of
pore solutions within porous building materials plays a role
in deterioration. Literature from the geosciences, as well as
building and art conservation, provides useful information
on the phenomena associated with salt weathering (Doehne
2002; Evans 1970; Goudie and Viles 1997; Winkler 1997).
Scaling of concrete surfaces by physical salt attack
should not be confused with scaling of concrete surfaces by
freezing and thawing of concrete in the presence or absence
of deicing salts. Concrete surfaces exposed to freezing and
thawing conditions in the presence of deicing salts may deteriorate by chemical and physical mechanisms (Chapter 4)
(Mehta and Monteiro 2006; Valenza and Scherer 2007).
8.2—Occurrence
Physical salt attack occurs throughout the world. The
process requires water-soluble salts from seawater, groundwater, soil, and other sources and ambient environmental
conditions, which usually involve fluctuations in temperature and relative humidity, typically on a diurnal basis. In
natural environments, this commonly occurs in arid regions,
such as the southwestern United States; portions of southern
Europe; coastal areas of Australia; and in much of the Middle
East, particularly in the Gulf regions. Additionally, localized
microclimates can provide similar conditions. Landscaping
practices and irrigation can also lead to conditions conducive to physical salt attack.
Physical salt attack may occur even when salt concentrations in soils are low because the salts concentrate over time
at concrete evaporative surfaces. Unlike chemical sulfate
attack, there is no reported threshold concentration of salts
or chemical components that indicate the potential severity
of attack. A useful approach for assessing the potential
for attack involves understanding the potential for cyclic
wetting and drying of evaporative surfaces, and determining
the types of water-soluble salts in contacting soils and water
Fig. 8.1a—Physical salt attack on concrete residential foundation. Scaling of the concrete surface resulted from sodium
sulfate in the soil pore water. The salt accumulated behind a
paint coating on the stem wall (Haynes and Bassuoni 2011).
Fig. 8.1b—Physical salt attack on concrete garage slab
caused by sodium carbonate. Scaling occurred along an
evaporation front (Haynes and Bassuoni 2011).
(atmospheric, ground, and irrigation). This involves determining whether soils or water contain water-soluble anions
such as sulfate, carbonate, bicarbonate, and chloride, and
cations such as sodium, calcium, potassium, and magnesium.
Solutions of sodium sulfate contain sulfate ions, so the
potential for chemical sulfate attack on concrete exists
when this salt is present in the exposure environment. Other
sulfate salts, such as calcium, magnesium, potassium, iron,
and ammonia sulfate, can participate in chemical sulfate
attack on concrete, but do not appear to damage concrete by
physical salt attack.
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8.3—Background
Much of the early literature reported evidence of physical
salt attack by observations of field-exposed concrete, usually
in sulfate-bearing soils and occasionally in carbonatebearing soils. In 1912, tests were conducted by Bates et al.
(1913) to investigate the cause of field distress to concrete
exposed to alkali waters (mostly sodium sulfate). This led to
a major test program incorporating test sites in seven areas of
the United States known to have soils with high concentrations of alkali salts. Williams and Furlong (1926) observed
that crystallization of salts in concrete pores occurred along
with some other unknown chemical action. The U.S. Bureau
of Reclamation (USBR) (1963) noted that salts such as
sodium carbonate can cause surface disintegration by crystallizing in concrete pores, and that this action appears to be
purely physical. Later, Reading (1975, 1982) reported that a
concrete tailrace wall of a dam showed deterioration caused
by physical action of sodium sulfate.
One of the more significant test programs identifying physical salt attack was the Portland Cement Association’s longterm field tests on concrete exposed to sulfate-containing soils
(McMillan et al. 1949; Stark 1982, 1989b, 2002). Overall,
thousands of specimens were tested. The early work by
McMillan et al. (1949) identified and discussed the anhydrous form (thenardite [Na2SO4]) and the decahydrate form
(mirabilite [Na2SO4·10H2O]) of sodium sulfate, but these salt
phases were not associated with a deterioration mechanism.
Stark (2002) examined deteriorated specimens microscopically and identified physical salt attack as a main cause of
deterioration in the field-exposed concrete specimens. Test
results showed resistance to deterioration was improved with
decreasing w/cm for concrete mixtures with and without
supplementary cementitious materials (SCMs). Concrete
mixtures found most resistant to physical salt attack contained
portland cement with no SCMs and 0.38 w/cm.
Stark (1982, 1989b, 2002) found that SCMs added to
concrete may decrease the resistance to surface deterioration.
Other investigators also observed detrimental effects related
to SCMs (Irassar et al. 1996; Bassuoni and Nehdi 2009).
Folliard and Sandberg (1994) conducted exploratory tests
on small concrete specimens to identify mechanisms of deterioration by inducing salt crystallization due to hydration,
evaporation, and changes in temperature. The most aggressive test environment was that of submerging specimens in
sodium sulfate solution and cycling the temperature between
41 and 86°F (5 and 30°C), with damage occurring as the
temperature decreased. The salt solution became supersaturated with respect to mirabilite as the temperature decreased;
hence, mirabilite precipitation or growth was likely responsible for deterioration. Haynes et al. (2008, 2010) reported on
a 3-year study in which concrete was partially submerged in
sodium sulfate, sodium carbonate, or sodium chloride solution and found that cycling of ambient environmental conditions caused more deterioration than when ambient conditions remained steady; however, damage was still observed
even when the ambient conditions were held steady. Sodium
sulfate was the most aggressive salt, followed closely by
sodium carbonate. Sodium chloride exposure produced
47
minor deterioration in laboratory testing in comparison to
the other salts.
8.4—Mechanism
Crystallization pressure is the primary cause of physical salt
attack. Scherer (2004a) provides a summary of the equilibrium thermodynamics that govern the development of crystallization pressure. As a general rule, crystallization pressures
increase with decreasing pore size within the concrete. These
pressures impose stresses on pore walls that ultimately cause
microcracking when pressures exceed tensile strength.
Salt crystallization occurs at evaporative surfaces because
the salts concentrate with evaporation, the solution becomes
supersaturated, and the salts precipitate. Once the salt crystallizes, they grow and generate pressure (Scherer 2004b).
Cycles of dissolving and recrystallization also cause damage.
For example, sodium chloride is hygroscopic; it absorbs
water from the air at relative humidities above 75.5 percent
(temperature range of 32 to 86°F [0 to 30°C]) until the crystals
dissolve, and then recrystallizes at lower relative humidity.
Sodium sulfate and sodium carbonate dissolve and recrystallize by a different mechanism. They can experience phase
changes due to changes in ambient temperature and relative humidity. Thenardite (Na2SO4), the anhydrous form
of sodium sulfate, is stable at ambient conditions of 68°F
(20°C) and relative humidity up to 75 percent. At higher
relative humidity, however, the crystals will absorb moisture from the air and dissolve, from which the decahydrate
form (mirabilite [Na2SO4·10H2O]) crystallizes. Above a
temperature of 90.3°F (32.4°C), only thenardite is stable.
These characteristics of the salt permit phase change to occur
on diurnal cycles, with thenardite present during the hot, dry
daytimes and mirabilite during the cold, damp nighttimes. The
phase changes are accompanied by a change in crystal size.
Thenardite-to-mirabilite conversion is accompanied by a 314
percent increase in crystal size. For sodium carbonate, conversion of the low-hydrate phase (thermonatrite [Na2CO3·H2O])
to the decahydrate phase (natron [Na2CO3·10H2O]) is accompanied by a 260 percent increase.
Factors important to distress by crystallization pressures
are the degree of supersaturation of the solution, the distribution of the salt in the pores (proximity of the evaporation
front to the surface), pore size and pore size distribution,
and sorptivity and tensile strength of concrete. The degree
of supersaturation is important because the potential for
crystallization increases with supersaturation and can lead
to more damage. Thermodynamic models show that higher
crystallization pressures develop in smaller pores (Scherer
2004a). Consequently, efflorescence is usually not linked to
salt-based deterioration because salt crystallization occurs
on exposed surfaces where such growth is easily accommodated. Evaporation fronts, however, can develop below
exposed surfaces when fine pore networks decrease the rate
of transport of solution to surfaces, which can also happen
when evaporation is rapid. Where supersaturation occurs
below exposed surfaces, crystallization occurs within the
small pores rather than on free surfaces. This is known as
subfloresence rather than efflorescence. When subfloresence
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occurs, damage may result because pressures develop in the
fine pores, which can cause microcracking.
8.5—Recommendations
Research directed to better understand this mechanism of
deterioration is needed, particularly in the areas of defining the
salt concentrations that can lead to deterioration, determining
the ambient environmental conditions that cause various
degrees of deterioration, and understanding the properties of
concrete that make it susceptible to physical salt attack.
Specific recommendations cannot be made to prevent physical salt attack; however, the salts of sodium sulfate and sodium
carbonate are primarily responsible for physical salt attack on
concrete, while sodium chloride causes less deterioration. Other
common salts such as calcium sulfate and magnesium sulfate
do not participate in physical salt attack. While Stark’s longterm study on concrete exposed to sulfate soils is the main reference work on the topic, the test program inadvertently obtained
results on physical salt attack. Two trends were observed:
concrete having low w/cm showed less deterioration than
concrete having high w/cm, and concrete mixtures containing
fly ash or slag cement did not perform as well as companion
specimens containing only portland cement.
Where the risk of physical salt attack is unacceptable,
construction methods should be used that separate concrete
from contact with salt solutions, such as using capillary
breaks or protective coatings.
CHAPTER 9—CORROSION OF METALS
AND DEGRADATION OF OTHER MATERIALS
EMBEDDED IN CONCRETE
9.1—Introduction
Understanding the conditions that cause corrosion
(rusting) of reinforcing and prestressing steel is vital. The
risk of excessive corrosion in concrete structures containing
embedded steel can be minimized to promote long service
lives. The purpose of this chapter is to summarize the mechanisms of steel corrosion, the conditions under which such
corrosion occurs, the methods and techniques that can be
used to prevent or limit steel corrosion, and the preservation
of other embedded materials.
Concrete protects against corrosion of embedded steel
because of the highly alkaline environment provided by the
pore fluid of the portland cement paste. The adequacy of
the protection depends on the depth of concrete cover, the
quality of the concrete, the details of the construction, the
degree of exposure to chlorides from concrete component
materials and from the environment, and the service environment. A more comprehensive treatment of the subject can
be found in ACI 222R and Broomfield (2007).
9.2—General principles of corrosion initiation in
concrete
9.2.1 General—The process of corrosion of steel in
concrete is divided into several phases:
1) Initiation: the normal protective passive layer on the
steel breaks down
2) Corrosion growth (propagation): the (active) corrosion
process is established and corrosion progresses
3) Damage: corrosion is sufficiently severe that cracking,
spalling, or both, occur and eventually the structural element
may not perform its intended function.
9.2.2 Protection mechanism in concrete
9.2.2.1 General—As described in ACI 222R, the high
alkalinity, with a pH greater than 12.5, of concrete protects
embedded steel reinforcement in concrete from corrosion.
When oxygen is present, the high pH of the pore solution
causes an ultra-thin corrosion film to form on the steel
surface, termed a “passive film”. The composition of this
film depends upon the metallurgy of the metal and is understood to be a combination of hydroxides and oxides. This
film is in equilibrium with the environment, slows corrosion reactions, and, thus, the steel is protected against active
corrosion and is said to be “passivated”.
Depending on the penetrability of concrete cover over
the steel and the alkalinity of the concrete pore solution, the
passive film is maintained. If the passive film breaks down,
termed “depassivation,” corrosion rate accelerates and the
propagation phase begins. The film can break down locally
so that localized corrosion results. If breakdown occurs over
larger areas, more uniform general corrosion takes place.
The primary causes of film breakdown include:
a) Chemical, physical, or mechanical degradation of the
concrete cover
b) Chloride penetration to the reinforcement
c) Carbonation of the concrete to reinforcement depth
d) Change of polarization of the reinforcing steel such as
in dissimilar metal corrosion or stray current leakage.
9.2.2.2 Corrosion process—Corrosion of steel in concrete
is an electrochemical process that requires the development
of an anode where oxidation takes place and a cathode where
reduction takes place. At the anode, electrons are liberated
and ferrous ions are formed.
Fe → Fe++ + 2e–
(9.2.2.2a)
At the cathode, electrons are consumed and hydroxyl ions
are formed and liberated.
2H2O + O2 + 4e– → 4(OH)–
(9.2.2.2b)
The ferrous ions may subsequently combine with oxygen
or hydroxyl ions and produce various forms of corrosion
products or rust. The formation of rust often causes expansion that, in turn, may cause cracking and spalling of the
concrete cover. Refer to ACI 222R for a more detailed
description of the corrosion process.
9.2.2.3 Breakdown due to insufficient oxygen supply—
The passive layer requires an oxygen flux corresponding to
approximately 0.2 to 0.3 mA/m2 (1.3 × 10–4 to 1.9 × 10–4
mA/in.2). If the oxygen flux is less than this, the passive film
will gradually reduce in thickness, exposing bare steel. The
result is corrosion at an extremely low rate (corresponding
to the oxygen flux), but at an active, though very negative,
potential (Fidjestøl and Nilsen 1980). Such low oxygen
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diffusion can occur in submerged concrete, in seawater or
fresh water (Fidjestøl et al. 1985), or by exposure below the
groundwater table.
9.2.2.4 Carbonation—The reaction between CO2,
commonly from the ambient air, and cement paste can reduce
the pH to less than 9, resulting in the loss of passivity. Corrosion rate increases, and the rate depends on availability of
oxygen as well as the moisture content, because this affects
the electrical resistivity of the concrete.
9.2.2.5 Local breakdown due to chloride—Chloride above
a certain concentration known as the chloride threshold will
cause local breakdown of the passive layer, leading to corrosion. The rate of corrosion depends on the availability of
oxygen and chlorides, the anode-cathode area ratio, and the
electrical resistivity of the concrete. The chloride concentration necessary to cause breakdown of the film depends on
the pH and composition of the pore water, the quality of the
oxide film, and the characteristics of the steel and concrete
interface, and is therefore not one value for all cases. Note
that the chloride threshold is a distribution of values.
Although modern concrete has more variations in porefluid composition, chiefly due to the use of supplementary
cementitious materials (SCMs) such as fly ash, slag, and
silica fume, the limits for chlorides in new construction as
given in ACI 222R are still conservative and appropriate.
ACI 318-14 contains higher allowable limits.
Because some concrete materials contain chloride that
will not be released into the concrete, past good performance
of these materials may provide a basis for permitting higher
chloride contents. The suggested chloride contents provide
a conservative approach that should result in low risk of
corrosion; this conservative approach is necessary because
of conflicting reports on chloride thresholds, the effects of
different exposure environments, and materials combinations. The conservative approach is also recommended
because exposure conditions, such as those encountered in
bridge decks, parking structures, and marine environments,
allow the penetration of chlorides from the environment.
Concrete should be made with constituents such that the
total chloride content in the concrete is within the guidelines
given in ACI 222R.
9.3—Propagation of corrosion
9.3.1 General—Once corrosion has been initiated, a structure may still have many years of service life, especially if
the rate of corrosion is very low. The factors in 9.3.2 through
9.3.4 control the corrosion rate.
9.3.2 Anodic control—Anodic control is based on sufficiently controlling the rate of dissolution of corrosion products formed at the anode. The rate at which the dissolution of
iron takes place determines the corrosion rate.
9.3.3 Cathodic control—The rate of corrosion is
controlled by the availability of oxygen at the cathode and
the ratio between cathodic and anodic areas, and is limited
by the availability of oxygen, the size of the cathode, or both.
Normally, the supply of oxygen at the cathode far exceeds
that needed to sustain corrosion, so the rate is controlled
by other factors. Coating the reinforcement is one way
49
of limiting the oxygen supply to the cathode surface. The
coating also prevents access of aggressive media to the steel
surface. Cathodic control also occurs where the concrete
is completely water-saturated, which greatly reduces the
oxygen flux from the concrete surface to the steel.
9.3.4 Resistivity control— An electrolyte is essential
for corrosion to propagate. Resistivity of the electrolyte
can limit corrosion propagation for uncracked concrete
(Marcotte and Hansson 2003). Resistivity control requires
the electrical resistance of the concrete to be high enough
as to limit the current that can be developed from the two
half-cell reactions.
There may be a large difference in half-cell potential
between the cathodic and anodic areas of the steel. If the
electrical resistance R of the concrete between the two areas
is sufficiently large, however, most of the potential difference is spent in overcoming the voltage drop IR, caused
by the current flow I against this resistance, even at minute
corrosion current densities. This effect can often be seen in
fully carbonated concrete in a reasonably dry environment,
in concrete that contains SCMs, or where there is a great
distance between anodic and cathodic areas. This means that
corrosion can be insignificant despite large differences in
half-cell potential.
9.4—Corrosion-related properties of concreting
materials
Materials for concrete should satisfy relevant standards
for structural concrete.
9.4.1 Portland cement—The alkalinity of portland cement
paste results from the presence of hydroxides of calcium,
potassium, and sodium in the pore solution. Calcium
hydroxide (CH) is the most abundant, and constitutes 15
to 25 percent of the paste. While the pH of saturated solutions of CH is only 12.4, the pH of 13.5 to 14 often found in
concrete pore water (Justnes and Nygaard 1994) is explained
by the OH– ions associated with alkalis in the concrete.
The presence of C3A in the cement can have two benefits:
reducing both chloride ingress and in binding admixed or
intruding chlorides. This was first established by Verbeck
(1968) and has since been confirmed by other research
(Rasheeduzzafar et al. 1992). The main conclusion of this
work is that the use of very low C3A (Type V) cements in a
strong chloride environment is generally not recommended.
9.4.2 Supplementary cementitious materials—Fly ash, slag
cement, and silica fume are generally assumed to improve the
resistance of concrete to chloride-induced corrosion. While
the introduction of such materials to concrete will consume
some of the Ca(OH)2 that acts as a buffer against changes in
pH due to carbonation of concrete (Bijen and van Selst 1991;
Horiguchi et al. 1994; Branca et al. 1992), improvements in
pore distribution and permeability can counteract this depletion in CaOH2 (Torii and Kawamura 1994; Hakkinen 1992).
Also, SCMs can increase the electrical resistivity of the
uncracked concrete, thus reducing the rate of any corrosion
that has been initiated (Schiessl et al. 1994; Fidjestøl 1987,
1991; Alonso et al. 1992).
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
In the 1990s, the development of corrosion-resistant
concrete focused on using blends of portland cement and
other cementitious materials (Baweja et al. 1994; Maage and
Helland 1991; Maage et al. 1994; Berke et al. 1991; Collepardi
et al. 1994; Anqi et al. 1991; Hussain and Rasheeduzzafar
1994; Decter et al. 1989; Haque et al. 1992; Ozyildirim 1994).
Malhotra et al. (2000) and Smith et al. (2004) all reported
that concrete containing moderate to high volumes of fly ash
exhibited superior resistance to the penetration of chlorides
and improved corrosion resistance.
Slag cement has been used in marine work since the early
1900s; the experiences with respect to resistance against
chloride-induced corrosion are generally good (Wiebenga
1984; Hope et al. 1985; Pal et al. 2002). Slag cement has also
been shown to improve resistance to penetration of deicer
salts (McGrath and Hooton 1997; Bleszynski et al. 2002).
Silica fume works in several ways to reduce the risk of
corrosion. The reduced permeability of silica fume concrete
means a greatly reduced rate of chloride penetration in
marine structures and structures exposed to deicing salts.
Such concrete also has very high electrical resistivity, thereby
greatly diminishing the rate of corrosion once it is initiated
(Wolsiefer 1991; Pettersson 1995; Fidjestøl 1987, 1993;
Fidjestøl and Frearson 1994; Alonso et al. 1992; Berke et al.
1992; Gautefall and Vennesland 1985; Zhang and Gjørv 1991;
Fidjestøl and Justnes 2002; Skjølsvold et al. 2007).
9.4.3 Aggregates—Aggregates can contain chloride
salts, particularly those aggregates that have been exposed
to seawater or whose natural sites are in groundwater
containing chloride. Sedimentary rock formed in ancient
seabeds can also contain significant amounts of chlorides.
There have been reported instances (Gaynor 1985) where
quarried stone, gravel, and natural sand contained small
amounts of chloride that have resulted in concrete chloride
contents that exceed the maxima described in 9.2.2.5. Note,
however, that tightly-bound chlorides in aggregate may not
contribute to corrosion of steel. ASTM C1524 can be used
to determine the water-extractable chloride content of aggregates that could potentially contribute to corrosion initiation.
9.4.4 Mixing water—Potable mixing water can contain
small amounts of chloride. ASTM C1602/C1602M has
optional limits for chlorides in mixing water: 500 ppm for
prestressed concrete, bridge decks, or as otherwise designated; and 1000 ppm for other reinforced concrete in moist
environments.
9.4.5 Admixtures
9.4.5.1 General—Admixtures containing significant
concentrations of CaCl2 should not be used in concrete
containing embedded metal. Some water-reducing admixtures can contain chloride to improve admixture performance, but contribute only small amounts of chloride to
the concrete when they are added at recommended rates.
Normal-setting admixtures that contribute much less than
0.1 percent chloride by mass of cement are most common;
their use should be evaluated based on the application.
Chemical admixtures are described in detail in ACI 212.3R.
9.4.5.2 Accelerators—Accelerating admixtures, other
than those based on CaCl2, have been used in concrete with
varying success. Accelerators that do not contain chloride
should not automatically be assumed to be noncorrosive.
The materials most commonly used in chloride-free accelerators are calcium formate, sodium thiocyanate, calcium
nitrate, and calcium nitrite. It is generally accepted that
formates (Holm 1987) are noncorrosive in concrete, and that
calcium nitrite is also an inhibitor.
9.4.5.3 Inhibitors—ACI 222.3R provides an overview of
corrosion inhibitors for concrete systems. Four corrosioninhibiting admixtures are common commercially: 1) amine
carboxylate; 2) amine-ester organic emulsion; 3) calcium
nitrite; and 4) an organic alkenyl dicarboxylic acid salt (ACI
212.3R). Amine carboxylate admixture was developed from
vapor phase inhibitors that have a long history of use in
other industries. As an anodic and cathodic inhibitor, it can
be useful in both new construction and repair applications
(Bavarian and Reiner 2004). Amine-ester organic emulsion
is reported to protect by reducing chloride ingress and by
forming a protective film at the steel surface (Nmai et al.
1992; Bobrowski and Youn 1993). Laboratory evaluations
indicate that amine-ester organic emulsion will delay the
onset and reduce the rate of corrosion (Nmai et al. 1992;
Nmai and Krauss 1994).
Calcium nitrite has been widely used as an accelerating
admixture that will also function as a corrosion inhibitor.
Laboratory studies have demonstrated that it delays the onset
of corrosion or reduces the rate after it has been initiated
(Berke 1985; Berke and Roberts 1989). The ratio of chloride ions to nitrite ions is important. Studies (Berke 1987)
show that calcium nitrite can provide corrosion protection
even at chloride-nitrite ratios exceeding 1.5 to 1.0 by mass.
Although dosage rates vary, 2 to 6 gal./yd3 (10 to 30 L/m3)
of concrete is the common range. Berke and Rosenberg
(1989) compiled an extensive review of the use of calcium
nitrite as a corrosion inhibitor for steel, galvanized steel, and
aluminum in concrete, which was later updated by Berke et
al. (1994). If the accelerating effect from calcium nitrite is
undesirable, use of a retarder is recommended. Increased
amounts of air-entraining admixture may be necessary when
calcium nitrite is used to maintain the desired air content.
Montes et al. (2004) showed that the effect of calcium nitrite
on corrosion inhibition in cracked elements was limited.
9.5—Mitigating corrosion
9.5.1 General—In mitigating corrosion, these critical
points should be evaluated in terms of concrete service life:
a) Initiation of corrosion
b) Corrosion products becoming visible (staining)
c) Serviceability, including cracking, spalling, or both
d) The load-carrying capacity of the structure if it is seriously reduced, or the structure can no longer perform its
intended purpose
Traditional service life considerations only consider the
initiation of corrosion; however, even for very conspicuous
and visible structures, criterion b) or c) may be more appropriate in design.
The corrosion of steel in concrete is dependent on the environment in which the structure is exposed. Table 19.3.1.1
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
in ACI 318-14 specifies exposure categories for corrosion
protection of reinforcement: C0 is designated for concrete
dry or protected from moisture; C1 is for concrete exposed
to moisture but not to an external source of chlorides; and
C2 is designated for concrete exposed to moisture and an
external source of chlorides from deicing chemicals, salt,
brackish water, seawater, or spray from these sources. Corrosion can be minimized by selecting processes and materials
that delay the onset of corrosion and then minimizing the
rate. Increasingly, service life prediction models are being
used to determine combinations of concreting materials
that can help achieve design service lives from a corrosion
perspective. Information on service life models are found in
ACI 365.1R. Detailed guidance on preventive strategies can
be found in ACI 222.3R. A general discussion of some of
the factors affecting corrosion resistance are described in the
following sections.
9.5.2 Design and process
9.5.2.1 Concrete quality and cover over steel
9.5.2.1.1 Cover depth—Extensive studies (Clear 1976;
Pfeifer et al. 1987; Marusin and Pfeifer 1985) have shown
that 1 in. (25 mm) cover over bare steel bars is inadequate
for chloride protection in severe corrosion environments,
even if the concrete has a w/cm as low as 0.30. Tests have also
shown that the chloride content in the top 1/2 in. (12 mm) of
concrete can be very high compared with that at depths of 1
to 2 in. (25 to 50 mm), even in concrete of high quality such
as one having a w/cm of 0.30. As a result, cover for moderateto-severe corrosion environments should be a minimum of
1-1/2 to 2 in. (40 to 50 mm).
Concrete will absorb salts applied in deicing operations. To
postpone initiation of corrosion, cover should be maximized.
Trejo and Reinschmidt (2007) reported that increasing the
concrete cover is the most effective way to increase the time
to corrosion and the service life of a concrete structure. Too
much cover, however, can increase cracking. Weyers et al.
(2003) reviewed the influence of cover depths and recommended cover depths for bridges of 2.75 in. (70 mm). ACI
318-14 specifies concrete cover requirements for Exposure
Condition C2 and these depend on additional exposure
conditions, member type, and reinforcement type and size.
9.5.2.1.2 Concrete quality—Numerous test programs have
shown that concrete made with a low w/cm and adequate
cover over the steel performs significantly better than
concrete made with a higher w/cm. Chloride ion penetration
to a 1 in. (25 mm) depth is approximately 400 to 600 percent
greater for concrete made with w/cm of 0.40 and 0.50 than
for concrete with w/cm of 0.32 (Pfeifer et al. 1987). Similarly, the proper use of SCMs can extend the time to corrosion and reduce its rate.
ACI 318 and other specifications place strict requirements
on the mixture proportions for severe chloride environments.
The Norwegian Public Roads Administration (2009) has
specific requirements for mixture proportions, depending
on the location of the structure (w/cm less than 0.38 to 0.40
and a certain silica fume content). The specifications for the
Great Belt (Storbælt) project required a low w/cm with the
use of silica fume and fly ash to provide a 100-year service
51
Table 9.5.2.1.2—Limits to chlorides in newly
constructed concrete (as recommended by ACI
222R)
Acid-soluble
(ASTM C1152/
C1152M)
Water-soluble
(ASTM C1218/
C1218M)
Soxhlet
method*
Prestressed
concrete
0.08
0.06
0.06
Reinforced
concrete in wet
conditions
0.10
0.08
0.08
Reinforced
concrete in dry
or protected
conditions
0.20
0.15
0.15
*
Soxhlet method described in ACI 222.1R.
Note: All chloride contents expressed as percent Cl– by mass of cement.
life (Storebælt Technical Publications 1999). A similar
mixture proportioning philosophy was used for 1.3 to 2.6
million yd3 (1 to 2 million m3) of concrete for the 6 mile (10
km) connection across Øresund from Sweden to Denmark
(Henriksen et al. 2000).
Admixed chlorides can also influence the quality and longterm performance of concrete structures. Table 9.5.2.1.2
gives the limits to chlorides in newly constructed concrete
recommended by ACI 222R.
Note that in specifications in other countries, such as CSA
A23.1-14/CSA A23.2 in Canada and EN206 in Europe,
chloride limits are expressed as percent by mass of total
cementing materials, which is portland cement plus SCMs.
Note that the chloride thresholds for prestressed concrete
are lower than those for reinforced concrete. Prestressing
steels have different chemical composition than conventional reinforcing steel, and often suffer from additional
corrosion mechanisms (9.6). The effect of corrosion mechanisms and high stresses in these steels is to make extra restrictions necessary compared to black steel. Structure type and
importance, fabrication methods, and exposure conditions
also influence corrosion performance of these steels.
For post-tensioned structures, the Federal Highway
Administration (FHWA) (2012) reported that although codes
allow a maximum of 0.08 percent total chloride content
by weight of cement in fresh grout, a lower total chloride
threshold value may be required.
9.5.2.1.3 Cracks—Cracks permit much faster chloride
infiltration rate than diffusion processes, and can establish
chloride concentration cells that accelerate corrosion. ACI
318-99 concluded that the role of cracks in the corrosion of
reinforcement is controversial and, without sound scientific
data relating crack width to corrosion activity, modified the
design procedure for limiting crack widths, eliminating the
crack width wmax from the design equation and substituting
an equation based on controlling the bar spacing. Although
ACI 318 does not provide explicit limits for crack widths in
the equation, this equation does indirectly limit the maximum
crack width to between 0.016 and 0.02 in. (0.4 and 0.52 mm).
In addition, the presence of cracking effects on corrosion
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
may be more than just the ingress of deleterious materials, as
when the crack tip reaches the steel; a small anode to large
cathode ratio is created, which may greatly accelerate corrosion, as well as further cracking. In structures submerged in
seawater, for example, large, dynamic (working) cracks have
been reported to be closed by brucite and aragonite (magnesium hydroxide and calcium carbonate) within a fairly short
period (Espelid and Fidjestøl 1986; Buenfeld and Newman
1986). ACI 224R and ACI 224.1R provide good overviews
of issues associated with cracks and corrosion.
To minimize crack formation, concrete should always be
made with the lowest practical water content. Also, proper
detailing of reinforcement, sufficient minimum and structural reinforcement, and control of heat of hydration and
restraint effects are important in producing a structure in
which cracks do not degrade corrosion resistance. Use of
shrinkage-reducing admixtures (SRAs), saturated prewetted
lightweight aggregate for internal curing, and mixture modification to reduce paste volume mixture optimization are
additional strategies to reduce cracking in concrete mixtures.
9.5.2.2 Concrete resistivity—When concrete is kept
moderately dry, corrosion of steel is minimized. For
example, if concrete containing up to 2 percent CaCl2 is
allowed to dry to a maximum internal relative humidity of
50 to 60 percent, embedded steel should either not corrode
or corrode at an inconsequential rate (Tutti 1982). Maintaining an internal relative humidity below 50 percent,
however, is not always possible.
While the surface regions of exposed concrete structures will have high or low electrical conductivity values
depending on the wetting and drying conditions of the environment, the concrete interior usually requires long drying
periods to achieve low electrical conductivity. Pfeifer et
al. (1987) found that 7 to 9 in. (180 to 230 mm) thick reinforced concrete slabs with w/cm ranging from 0.30 to 0.50
have essentially equal initial AC electrical-resistance values
between the top and bottom reinforcing bar mats at 28 days.
Cementitious materials that include SCMs can give very
high electrical resistivity in concrete. Slag cement, fly ash,
and, in particular, silica fume, will give concrete resistivities far in excess of what is provided by portland-cement
concrete (Cabrera and Ghoddoussi 1994; Fidjestøl and
Frearson 1994; Gautefall and Vennesland 1985; Berke
1988). Similarly, AC resistance tests on concrete made with
silica fume at a w/cm of 0.20 show extremely high initial
electrical resistance when compared with concrete having a
w/cm of 0.30 to 0.50. The high electrical resistance of silica
fume concrete can be due to denser paste microstructure; to
changes in the pore chemistry; and at low w/cm, to self-deiccation. Field investigations after more than 20 years confirm
long-term performance of SCMs (silica fume) (Fidjestøl and
Justnes 2002; Skjølsvold et al. 2007).
The high electrical resistivity of blended binder systems is
confirmed by tests using ASTM C1202 (or AASHTO T277),
which provides a method to determine conductivity that is
then used as an indirect indication of chloride diffusivity. In
several investigations, there has been a relationship to chloride diffusion determined by more conventional diffusion
or ponding tests (Detwiler and Fapohunda 1993; Wolsiefer
1991; Misra et al. 1994; Burg and Öst 1992). As implied in
the standard, however, this relationship cannot be assumed
to be universal because it also depends on the composition
of the binder system, such as content of and type of cement
and cementitious materials used.
9.5.3 Construction aspects
9.5.3.1 Workmanship—Good workmanship is vital for
securing uniform concrete with low penetrability. For lowslump concrete, segregation and honeycombing can be
avoided by good consolidation. Meeting the requirements of
the specifications pertaining to durability are essential.
9.5.3.2 Reinforcement detailing—Two factors are important to consider in detailing of the reinforcement:
1. Adequate spacing should be provided to allow for
proper placing of the concrete cover so that honeycombing
and poor compaction are avoided and good bond between
concrete and steel are obtained.
2. Corrosion is relatively more severe for small bars than
for large bars. Corrosion of a No. 3 (10 mm) bar totaling
0.04 in. (1 mm) of corrosion means nearly 40 percent loss of
cross section, whereas for a No. 8 (25 mm) bar, it will mean
15 percent loss of cross section. Note, however, that large
bars could cause larger cracks than smaller bars because
smaller bars can give better crack distribution.
9.5.3.3 Curing—Good curing reduces permeability
because of increased hydration of the cement. At least 7 days
of uninterrupted moist curing or membrane curing is ideally
specified. Limiting early thermal stresses is also important.
Good curing reduces transport parameters as well as charge
passed in ASTM C1202 (or AASHTO T277) (Acker et al.
1986; Marusin 1989). In addition, due to insufficient curing,
the chloride penetration rate in the near-surface part of the
concrete cover can be many times greater than at depth
(Hooton et al. 2002).
9.5.3.4 Formwork—Good, tight formwork is essential. Properly supported screeding equipment and correct
supports for the reinforcement are important for attaining the
cover protection specified. The use of side form spacers for
reinforcing bars in vertical formwork is similarly important.
Controlled permeability liners for formwork may improve
the quality of the cover (Sha’at et al. 1993).
9.5.4 Design—Design can do much to reduce corrosion
attack because proper detailing can minimize accumulation
of salts and the establishment of high humidity areas where
corrosion can be sustained.
9.5.4.1 General layout of structure—Placement and
general layout of the structure are important for a favorable
environment. An increase in the height of a bridge over the
sea will reduce the chloride exposure: field inspection of
concrete bridges (Fluge and Blankvoll 1995) found that an
increase in bridge height above sea level from 26 to 92 ft
(8 to 28 m) reduces the amount of chlorides deposited on
the surface by as much as 85 percent. The chloride exposure
was also up to eight times higher in the lee side of the structure than on the windward side. Similarly, moving bridge
columns away from traffic splash will reduce the chloride
exposure of the concrete.
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
9.5.4.2 Drainage—Particular attention should be given to
design details to ensure that water will drain and not pond
on surfaces. There are a number of details that are important
(Kompen 1994), such as proper slope and extended drainage
pipes that take the water away from the concrete surface.
9.5.4.3 Exposed items—Attention should be given to
partially embedded or partially exposed items, such as
bolts, that are exposed directly to corrosive environments.
The resistance of these items to the corrosive environment
should be investigated, and the coupling of dissimilar metals
should be avoided. Concrete should be carefully placed
around embedded items so that it is well consolidated and
does not create paths that will permit corrosive solutions to
easily reach the interior of the concrete.
9.5.5 Special protective systems—Costs of repairing
corrosion-induced damage are very high. Many protective
systems have been proposed, and the reader is referred to
ACI 222R, ACI 222.3R, and ACI 515.2R for a comprehensive understanding of protective and mitigating options.
Some of these protective systems have been shown to be
effective. Several of these are listed as follows:
a) Overlays and patches of very low w/cm (0.32), latexmodified concrete overlays (Clear and Hay 1973; FHWA
1975), concrete containing silica fume, and concrete
containing high-range water-reducing admixtures
b) Epoxy-coated reinforcing steel
c) Corrosion-resistant steels (Rasheeduzzafar et al.
1992; Trejo and Pillai 2004; Clemeña and Virmani 2004;
Williamson et al. 2003)
d) Waterproof membranes (Van Til et al. 1976)
e) Surface protective-barrier systems produced from
silanes, siloxanes, epoxies, polyurethanes, and methacrylates (Van Daveer and Sheret 1975)
f) Cathodic protection
9.6—Corrosion of prestressed steel reinforcement
The mechanisms and risks associated with general surface
corrosion and pitting in prestressed steel reinforcing systems
are comparable to conventional reinforcing steel (9.1 to 9.5),
with added concerns of fracture due to hydrogen embrittlement and stress corrosion cracking. Hydrogen embrittlement
is the result of a loss of ductility in the steel reinforcement
from the local absorption of atomic hydrogen released from
corrosion cells, and contact with water, hydrogen sulfide,
and other sources at the steel surface. Stress corrosion
cracking is similarly a brittle fracture event caused by the
interaction of the tensile stress within the steel reinforcement and corrosive environment produced as described
previously. Certain elements of prestressing steel systems,
such as end anchorages and slab tendons, that are exposed
to deicing solutions or come in contact with other metal
components (for example, ducts), may have added risks due
to reduced concrete cover protection, galvanic action, or
direct exposure to aggressive environments. Unlike general
surface corrosion, these mechanisms produce sudden loss of
prestressing function and permanent damage to the afflicted
element(s), and are difficult to locate via inspection or
testing when the structure is in service. Preventive measures
53
during fabrication/construction are needed to reduce the risk
of occurrence. ACI 222.2R provides addition details on the
prestressing systems, deterioration, protection, field evaluation, and remediation techniques.
Corrosion of prestressing steels is best prevented when
protective measures start at the fabrication shop and continue
through proper installation and placement in durable sheath,
concrete, or grout materials. Prestressing steel raw material
and finished anchorage products should be protected from
exposure to corrosive elements such as rain, snow, deicing
chemicals, salt-spray, and water at the fabrication shop,
during transport to the construction, and throughout the
on-site storage process until placement. Ideally, tendons and
wire should be stored indoors in a climate-controlled warehouse, shipped after shrink-wrapping and coverage with
tarps, and stored on site in a climate-controlled environment.
These actions prevent the formation of any corrosion cells
before placement.
Protection must continue once tendons, strands, or wires
are prepared for installation, placed in the formwork, and
prepared for tensioning. Temporary protection of exposed
tendons or strands and anchorage materials at construction
joints and ends is recommended, as the prestressed reinforcement may be exposed to the atmosphere during the
period when the concrete is placed and reaches the required
strength for stressing. Sheaths and ducts must be watertight, with excess water removed. Any tendon tails or extra
strand length should be cut and capped after inspection is
complete and stressing is started. Before and after stressing,
all exposed surfaces should be inspected and cleaned, with
water removed, and anchor cavities grouted shortly thereafter to avoid exposure. Systems that employ grease canisters at end anchorages should be periodically inspected to
confirm that the grease is not leaking and the canisters are
full. Certain specialized evaluation and repair procedures
have been developed to address select prestressing systems
after being placed in-service. For a more detailed discussion
of the corrosion of prestressing strand, refer to ACI 222.2R.
9.7—Degradation of materials other than steel
9.7.1 Introduction—Nonferrous metals are occasionally
used in concrete. These metals include aluminum, lead,
copper and copper alloys, zinc, cadmium, Monel metal,
stellite (cobalt-chromium-tungsten alloys), silver, and tin.
Galvanized steel and special alloys of steel, such as stainless
steels and chrome-nickel steels, have also been used. Zinc
and cadmium are used as coatings on steel.
Corrosion of nonferrous metals or alloys can result from
various phenomena. The metal may be unstable in highly
alkaline concrete or in the presence of chloride ions. The
former occurs when the concrete is relatively fresh and may
be self-limiting. The latter can initiate corrosion, particularly when the metal is in contact with a dissimilar metal.
When dissimilar metals are in electrical contact (coupled),
a galvanic cell can occur, resulting in corrosion of the more
active metal. More detailed information on corrosion of
nonferrous metals is available (Fintel 1984; Erlin 2006).
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9.7.2 Aluminum—Corrosion of aluminum embedded in
concrete can crack the concrete. Conditions conducive to
corrosion are created if the concrete contains steel in contact
with the aluminum, chlorides are present in appreciable
concentrations, or the cement is high in alkali content (Woods
1968; Erlin 2006). When the metals are coupled, increasing
ratios of steel area, particularly in the presence of appreciable
chloride concentrations, increase corrosion of the aluminum.
Additionally, hydrogen gas evolution may occur when fresh
concrete contacts aluminum. This may increase the porosity of
the concrete and, therefore, the penetration of future corrosive
agents. Some aluminum alloys are more susceptible to this
problem than others. Corrosion inhibitors, such as calcium
nitrite, have been shown to improve the corrosion resistance
of aluminum in concrete (Berke and Rosenberg 1989).
9.7.3 Lead—Lead in damp concrete will be attacked by
the CaOH2 in the concrete, and may be destroyed in a few
years. Contact of lead with reinforcing steel can accelerate
the attack. Lead should be isolated from the concrete by
protective plastic, or other materials that are unaffected by
damp concrete. Corrosion of embedded lead is not likely to
damage the concrete (Alhassan 2005).
9.7.4 Copper and copper alloys—Copper is not normally
corroded by concrete, as is evidenced by the widespread and
successful use of copper waterstops and the embedment of
copper pipes in concrete for many years (Erlin 2006). Corrosion of copper pipes, however, has been reported where
ammonia is present. Also, there have been reports that small
amounts of ammonia, and possibly of nitrates, can cause
stress corrosion cracking of embedded copper. Galvanic
corrosion of steel will occur if the steel is connected to the
copper (Erlin 2006).
9.7.5 Zinc—Zinc reacts with alkaline materials, such as
those found in concrete. Zinc in the form of a galvanizing
coating on reinforcing steel, however, is sometimes intentionally embedded in concrete. Available data are conflicting
as to the benefit, if any, of this coating (Cook 1980; Stark
and Perenchio 1975; Hill et al. 1976; Griffin 1969; Federal
Highway Administration 1976). A chromate dip on the
galvanized bars or the use of 400 ppm of chromate in the
mixing water is recommended to prevent hydrogen evolution in the fresh concrete. Use caution when using chromium
salts because of possible skin allergies. Additionally, users
are cautioned against permitting galvanized and black steel
to come in contact with each other in a structure because the
use of dissimilar metals can cause galvanic corrosion. Corrosion inhibitors, such as calcium nitrite, have been shown to
improve the corrosion resistance of zinc in concrete (Berke
and Rosenberg 1989; Page et al. 1989).
9.7.6 Other metals—Chromium and nickel alloys generally have good resistance to corrosion in concrete, as do
silver and tin. The corrosion resistance of some of these
metals may be adversely affected by the presence of soluble
chlorides from seawater or deicing salts. Use of stainless
steel may be economically justified in some high chloride
environments where the higher initial cost is offset by reduced
cost in service over the life cycle. Examples would be marine
locations and heavily deiced bridge decks. The 300 Series
stainless steels, however, are susceptible to stress corrosion
cracking when the temperature is over 140°F (60°C) and chloride solutions are in contact with the steel material. Embedded
natural-weathering steels generally do not perform well in
concrete containing moisture and chloride. Weathering steels
adjoining concrete may discharge rust and cause staining of
concrete surfaces (McDad et al. 2000).
9.7.7 Polymers—Polymers are being used increasingly
in concrete in applications such as pipes, shields, reinforcement, waterstops, chairs, and concrete reinforcement. Many
plastics are resistant to strong alkalis and would, therefore,
be expected to perform satisfactorily. Because of the great
variety of plastics and materials compounded with them,
however, specific test data should be developed for each
intended use.
9.7.8 Wood—Wood has been widely used in or against
mortars and concrete. Such use includes the incorporation of
sawdust, wood pulp, and wood fibers in the concrete and the
embedment of timber (Erlin 2006).
The use of untreated sawdust, wood chips, or fibers
usually results in slow-setting and low-strength concrete
(Erlin 2006). The addition of hydrated lime equal to onethird to one-half the volume of the cement is usually effective in minimizing these problems. The further addition of
up to 5 percent of CaCl2 dihydrate by mass of cement has
also helped to minimize these problems (Erlin 2006). CaCl2
in such amounts can cause corrosion of embedded metals,
however, and can have adverse effects on the concrete.
Another problem with such concrete is the high volume
change, which occurs even with changes in atmospheric
humidity. This volume change may lead to cracking and
warping (Erlin 2006).
The embedment of lumber in concrete has sometimes
resulted in leaching of the wood by Ca(OH)2 with subsequent
deterioration. Soft woods, preferably with high resin content,
are reported to be most suitable for such use (Erlin 2006).
9.8—Summary
Portland-cement concrete can provide excellent corrosion
protection to embedded steel. When corrosion occurs, the
costs of repairs can be exceedingly high. The use of highquality concrete; adequate cover over the steel; appropriate
reinforcement type; and proper design, including detailing
and additional protection, are prerequisites if deterioration
due to steel corrosion is to be minimized.
ACI 222R and ACI 222.3R provide a summary of the
causes and mechanisms of corrosion of steel. They include
information on how to protect against corrosion in new
structures and procedures for identifying corrosive environments and remedial measures where corrosion is occurring.
CHAPTER 10—ABRASION
10.1—Introduction
The abrasion resistance of concrete is defined as the ability
of a surface to resist being worn away by rubbing and friction.
Abrasion of floors and pavements can result from production
operations or foot or vehicular traffic. Abrasion resistance
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is, therefore, of concern in industrial floors. Wind, water, or
waterborne particles can also abrade concrete surfaces (Price
1947). There are instances where abrasion is of little concern
structurally, yet there may be a dusting problem that can be
objectionable in some kinds of service. Abrasion of concrete
in hydraulic structures is discussed only briefly in this guide;
the subject is treated in more detail in ACI 210R and 210.1R.
10.2—Testing concrete for resistance to abrasion
Research to develop meaningful laboratory tests on
concrete abrasion has been ongoing since the early 1900s.
There are several different types of abrasion, and no single
test method has been found that is adequate for all conditions. A detailed description of abrasion/erosion test methods
can be found in Bakke (2006). Prior (1966) described four
broad areas related to abrasion:
1. Wear on floor and slab construction; Table 10.2 shows
classes of floor traffic and use, and the special considerations
required for good wear resistance (ACI 302.1R).
2. Wear on concrete road surfaces due to attrition, scraping,
and percussion from heavy trucks and automobiles.
3. Erosion of hydraulic structures, such as dams, spillways, tunnels, bridge piers, and abutments, due to the action
of abrasive materials carried by flowing liquid (attrition and
scraping).
4. Cavitation action on concrete in dams, spillways,
tunnels, and other hydraulic structures due to high flow
velocities and negative pressures.
ASTM C779/C779M covers three operational procedures for evaluating floor surfaces: Procedure A, revolving
discs; Procedure B, dressing wheels; and Procedure C, ball
bearings. ASTM C944/C944M is similar to ASTM C779/
C779M Procedure B and is used for testing smaller areas
than required for ASTM C779/C779M.
Each method has been used to develop information on
wear resistance. Liu (1994) commented that the most reproducible results are obtained by the method involving the use
of revolving discs. Reproducibility of abrasion testing is an
important factor in selecting the test method. Replication of
results is necessary to avoid misleading results from single
tests.
The concrete surface condition, loose aggregates that
are dislodged and abraded during the test procedure, and
care and selection of representative samples can all affect
test results. Samples that are fabricated in the laboratory
should be identical for comparison, and the selection of
field-testing sites should be made on the basis of providing
representative results.
To set limits for abrasion resistance of concrete, it is necessary to rely on the relative values obtained during testing to
provide a prediction of service.
Underwater abrasion presents special demands for test
procedures used to assess durability. ASTM C1138 uses agitation of steel balls in water to determine abrasion resistance.
ASTM C418 uses a sandblasting apparatus to measure the
depth or wear to simulate comparative sand-impinged wear
resistance. This test provides a means for evaluating resistance to abrasion caused by wind-blown sand.
55
The abrasion resistance of pervious concrete can be
measured using ASTM C1747/C1747M. In this test method,
the impact and abrasion resistance of the pervious concrete
is determined from the percent mass loss from three cylindrical concrete specimens after they are placed in a Los
Angeles machine and rotated for 500 revolutions.
The abrasion resistance of railroad crosstie rail seats is
measured using AREMA Test 6 (AREMA 2007). This test
method simulates the abrasion that occurs when moisture
and sand are present between the concrete crosstie rail seat
and pad when heavy loads are applied at an angle. A cyclic
load is applied on the rail at a 27.5-degree angle from vertical
with sand and a water drip placed on each side of each rail
seat. Crossties that have individual components break or a
rail deflection greater than 0.2 in. (5 mm) during loading
before 1,000,000 cycles fail the test (AREMA 2007). This
system test is thought to better simulate the kinetic friction
that causes the pad and grit under the pad to move laterally
back and forth against the crosstie rail seat, causing wear
(Kernes et al. 2011).
In summary, a number of tests previously described aid in
determining the durability of a select concrete mixture design
and sample construction to simulate abrasive actions. Many
of these tests target specific exposures or structural types (for
example, railroad ties). Application of one or more of these
tests to assess future performance of a specific structural
type to one or more forms of abrasion requires consideration
and alignment of test conditions (for example, sample size
and surface conditions) to expected service conditions.
10.3—Factors affecting abrasion resistance of
concrete
The abrasion resistance of concrete is a progressive
phenomenon. Initially, the resistance is related to compressive strength of the wearing surface. Therefore, initial judgments regarding relative floor wear can be made on the basis
of compressive strength.
As softer paste wears away, however, the particles of fine
and coarse aggregate are exposed, and abrasion and impact
will cause additional degradation that is more related to the
paste-to-aggregate bond strength and the relative hardness of
the aggregate than to the compressive strength of the concrete.
Tests and field experience have generally shown that abrasion resistance is proportional to the compressive strength of
concrete (Scripture et al. 1953; Witte and Backstrom 1951).
Because abrasion occurs at the surface, it is critical that the
surface strength be maximized. Resistance can be greatly
improved by the use of dry shakes and toppings, finishing
techniques (10.4.4), and curing. In addition, the use of
concrete mixtures having low to moderately low w/cm (less
than 0.45) is recommended to improve the strength and wear
resistance of surface paste.
Although useful as a relative indicator, reliance should
not be placed solely on the results of compressive strength
tests. Inspection should be made during installation and
finishing of floor slabs to obtain an abrasion-resistant surface
by encouraging the use of both power-trowel finishing and
adequate curing (Kettle and Sadegzadeh 1987).
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Table 10.2—Floor classifications* and considerations to improve abrasion resistance
Class
Anticipated traffic type
Use
Special considerations
Final finish
1. Exposed
Exposed surface—foot traffic
Offices, churches, multiunit
residential, decorative
Uniform finish, nonslip aggregate
in specific areas, curing
Normal steel-troweled finish,
nonslip finish where required
Colored mineral aggregate, color
pigment or exposed aggregate,
stamped or inlaid patterns, artistic
joint layout, curing, surface treatment, maintenance
Burnishing or polishing to
enhance sheen as required
2. Covered
Covered surface—foot traffic
Offices, churches, commercial,
multiunit residential, institutional with floor coverings
Flat and level slabs suitably dry for
applied coverings, curing
Light steel-troweled finish
3. Topping
Exposed or covered
surface—foot traffic
Unbonded or bonded topping
over base slab for commercial
or non-industrial buildings
where construction type or
schedule dictates
Base slab—good uniform level
surface tolerance, curing
Base slab—troweled finish
under unbonded topping;
clean, textured surface under
bonded topping
Unbonded topping—bondbreaker
on base slab, minimum thickness
3 in. (75 mm), reinforced, curing
Bonded topping—properly
sized aggregate, 3/4 in. (19 mm)
minimum thickness curing
Topping—for exposed surface,
normal steel-troweled finish;
for covered surface, light
steel-troweled finish
4. Institutional/
commercial
Exposed or covered
surface—foot and light
vehicular traffic
Institutional or commercial
Level and flat slab suitable for
applied coverings, nonslip aggregate
for specific areas, curing; coordinate joints with applied coverings
Normal steel-troweled finish
5. Industrial
Exposed surface—industrial
vehicular traffic such as pneumatic wheels and moderately
soft solid wheels
Industrial floors for manufacturing, processing, and
warehousing
Good uniform subgrade, joint
layout, joint load transfer, abrasion
resistance, curing
Hard steel-troweled finish
6. Heavy
industrial
Exposed surface—heavyduty industrial vehicular
traffic such as hard wheels and
heavy wheel loads
Industrial floors subject to
heavy traffic; can be subject to
impact loads
Good uniform subgrade, joint
layout, joint load transfer required,
abrasion resistance, curing
Special metallic or mineral
aggregate surface hardener;
repeated hard steel-troweling
7. Heavy
industrial
topping
Exposed surface—heavyduty industrial vehicular
traffic such as hard wheels and
heavy wheel loads
Bonded two-course floors
subject to heavy traffic and
impact
Base slab—good uniform
subgrade, reinforcement, joint
layout, level surface, curing
Clean, textured base slab
surface suitable for subsequent bonded topping. Special
power floats for topping are
optional, hard steel-troweled
finish
Topping—composed of wellgraded all-mineral or all-metallic
aggregate. Minimum thickness
3/4 in. (19 mm)
Mineral or metallic aggregate
surface hardener applied to highstrength plain topping to toughen,
curing
*
8.Commercial/
industrial
topping
As in Classes 4, 5, or 6
Unbonded topping—on new or
old floors where construction
sequence or schedule dictates
Bondbreaker on base slab,
minimum thickness 4 in. (100
mm), abrasion resistance, curing
As in Classes 4, 5, or 6
9.Critical
surface profile
Exposed surface—superflat
or critical surface tolerance
required; special materialshandling vehicles or robotics
requiring specific tolerances
Narrow-aisle, high-bay warehouses; television studios, ice
rinks, or gymnasiums (ACI
360R)
Varying concrete quality requirements. Special application procedures and strict attention to detail
are recommended when shake-on
hardeners are used. FF 50 to FF
125, superflat floor, curing
Strictly following techniques as
indicated in 8.9 of ACI 302.1R
Taken from Table 4.1 of ACI 302.1R-15.
With a given concrete mixture, compressive strength at
the surface is improved by:
a) Avoiding segregation
b) Eliminating bleeding
c) Properly timed finishing
d) Minimizing surface w/cm (water addition to the surface
during finishing should not be permitted)
e) Hard troweling of the surface, which should not be done
on concrete containing an air-entraining admixture or having
a total air content greater than 3 percent (refer to ACI 302.1R)
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
f) Proper curing procedures
Economical proportioning of the mixture for increased
compressive strength includes using a limit on the maximum
w/cm and proper aggregate size. When supplementary
cementitious materials (SCMs) such as silica fume, slag, or
fly ash are used in concrete, the abrasion resistance is generally related to the compressive strength developed (Keshari
2009; Naik et al. 1995, 2002; Turk and Karatas 2011; Yen
et al. 2007). The use of good curing procedures with SCMs
is essential to ensure adequate abrasion resistance (Yen et
al. 2007). Polymer concrete, polymer-impregnated concrete
(Holland and Gutschow 1987), epoxy concrete (Mirza et al.
1990), calcium aluminate cement (Scrivener et al. 1999),
and calcium sulfoaluminate cement (Markey et al. 2006)
have shown exceptional abrasion resistance. Consideration
should be given to the quality of the aggregate (Scripture
et al. 1953; Smith 1958). The service life of some concrete
slabs, such as warehouse floors that are subjected to abrasion
by steel or hard rubber-wheeled traffic, is greatly lengthened
by the use of hard, tough aggregates.
The abrasion resistance of lightweight concrete is a function of the concrete compressive strength; however, the
use of only lightweight aggregates may not be advisable
for structures with high abrasion resistance requirements
(ACI 213R). The abrasion resistance of concrete containing
recycled aggregate will depend greatly on the compressive
strength and aggregate type of the recycled concrete (Ekolu
et al. 2012; de Brito 2010).
Abrasion-resistant aggregates can be used either with the
dry-shake method (ACI 302.1R) or as part of a high-strength
topping mixture. If abrasion is the principal concern, addition of high-quality quartz, traprock, or emery aggregates
properly proportioned with cement will increase the abrasion resistance by improving the compressive strength at
the surface. The aggregates used in topping mixtures or dry
shakes should be harder than the aggregate in the concrete.
For additional abrasion resistance, a change to a blend of
metallic aggregate and cement will further increase the abrasion resistance and increase service life. Another advantage
of using metallic aggregate is improved impact resistance,
especially at joints.
The use of two-course floors using a high-strength topping
is generally limited to floors where both abrasion and impact
resistance are required. While providing excellent abrasion
resistance, a two-course floor will generally be more expensive. Additional impact resistance can be obtained by using a
topping that contains portland cement and metallic aggregate.
A key element in the production of satisfactory floor
surfaces is curing (Liu 1994; ACI 302.1R; ACI 308R).
Because the uppermost part of the concrete surface is the
region that is abraded by traffic, maximum strength and
toughness are the most important elements for ensuring
resistance to surface abrasion. This is partially accomplished
through proper finishing operations, troweling techniques,
and adequate and timely curing practices (10.4.4). The effect
of curing efficiency (absorptivity) at the top-wearing surface
has been shown to be directly related to abrasion resistance.
57
Curing has less effect on the abrasion resistance of deeper
sections of the same concrete (Senbetta and Scholer 1984).
10.4—Recommendations for obtaining abrasionresistant concrete surfaces
10.4.1 Factors affecting abrasion resistance—The
following factors directly impact concrete strength and,
therefore, abrasion resistance (ACI 302.1R):
a) A low w/cm at the surface—Steps to lower w/cm include
the use of water-reducing admixtures, mixture proportions to
reduce bleeding, timing of finishing operations that avoid the
addition of water during troweling, and vacuum dewatering.
b) Well-graded fine and coarse aggregates (meeting
ASTM C33/C33M)—The maximum size of coarse aggregate
should be chosen for optimum workability and minimum
water content.
c) Lowest slump consistent with proper placement and
consolidation as recommended in ACI 309R.
d) Air content consistent with exposure conditions—In
addition to a detrimental effect on compressive strength, air
content levels can contribute to surface blistering and delamination if finishing operations are improperly timed. Entrained
air should not be used for dry-shake toppings unless special
precautions provided by the manufacturer are followed.
10.4.2 Two-course floors—High-strength toppings in
excess of 6000 psi (40 MPa) provide increased abrasion
resistance. The nominal maximum aggregate size for topping
mixtures is 1/2 in. (12.5 mm).
10.4.3 Special concrete aggregates—Selection of hard
aggregates for improved strength performance at a given w/
cm also improves abrasion resistance. Typically, aggregates
are applied as dry shakes or in high-strength, bonded toppings.
10.4.4 Proper finishing procedures—Floating and troweling operations should be delayed until the concrete has lost
its surface sheen. It may be necessary to remove free water
from the surface to permit finishing operations to continue
before the base concrete hardens. Standing water should
never be worked into concrete surfaces because it reduces the
compressive strength of the surface paste. The delay period
will vary greatly depending on temperature, humidity, air
movement, and SCMs used. Greater detail regarding proper
finishing operations is provided in ACI 302.1R.
10.4.5 Vacuum dewatering—Vacuum dewatering is a
method for removing water from concrete immediately
after placement (New Zealand Portland Cement Association
1975). While this permits a reduction in w/cm, the quality of
the finished surface is still highly dependent on the timing of
finishing and subsequent actions by the contractor. Ensuring
that proper dewatering is accomplished at the edges of the
vacuum mats is essential. Improperly dewatered areas are
less resistant to abrasion because of a localized higher w/cm.
10.4.6 Special dry shakes and toppings—When severe
wear is anticipated, the use of special dry shakes or topping
mixtures should be used. The manufacturer recommendations should be followed. Additional guidance is provided
in ACI 302.1R.
10.4.7 Proper curing procedures—For most concrete
floors, water curing (keeping the concrete continuously wet)
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
is the most effective method of producing a hard, dense
surface (Shurpali et al. 2012). Water curing, however, may
not be always practical. Curing compounds, which reduce
moisture loss in concrete from evaporation, are used as
an alternative. Curing compounds also provide protection
against early carbonation and prevent premature or excessive loss of surface moisture. Moist curing of metallic shake
toppings is not recommended because some water sources
and rainwater have a pH of less than 7, which may result
in the oxidation of the metallic aggregate particles. For
pervious concrete, it was found that curing with a plastic
sheet provided improved abrasion resistance. Application of
soybean oil or a curing compound to the pervious concrete
surface was also found to increase abrasion resistance
(Kevern et al. 2009). Latex-modified concrete was found to
greatly improve the abrasion resistance of pervious concrete
(Wu et al. 2011).
Water curing is accomplished through the use of sprays,
ponding, or wet coverings such as damp burlap, or cotton
mats. Water-resistant paper or plastic sheets are satisfactory,
provided that the concrete is first wetted and then immediately covered, with the edges overlapped and sealed using
water-resistant tape. The use of plastic sheeting without a
damp cloth layer can result in nonuniform surface color.
Curing compounds should meet the minimum requirements of ASTM C309 or ASTM C1315. They should
be applied in a uniform coat immediately after concrete
finishing and in accordance with the manufacturer’s recommendations. Recommended coverage rates will vary
depending on the surface texture of the finished surface. A
smoother floor surface will have better moisture retention
properties compared with a textured highway slab. Smaller
peaks and valleys result in a lower evaporation rate and,
therefore, require a lower coverage rate. The compound
should be covered with scuff-resistant paper if the floor
is subjected to traffic before curing is complete. Curing
compounds should not be required for surfaces that receive
paint or floor tile unless the curing compound is compatible
with these materials.
Wet curing is recommended for concrete with a low w/cm to
supply additional water for cement hydration, where cooling of
the surface is desired, where concrete will later be bonded,
or where liquid hardeners will be applied. Curing methods
are described in detail in ACI 308R.
Heaters burning fossil fuels or other sources of carbon
dioxide (CO2), such as finishing machines, vehicles, and
welding machines, should not be used without attention to
proper ventilation. CO2 can adversely affect fresh concrete
surfaces between the time of placement and application of a curing compound through a mechanism referred
to as carbonation. The severity of the effect is dependent
on the concentration of CO2, the humidity and ambient
temperature, and the length of exposure to the air (Kauer
and Freeman 1955; Matsuzawa et al. 2010). Early carbonation will greatly reduce the abrasion resistance of concrete
surfaces. The extent of the reduction depends on the depth of
carbonation. The only effective repair is to grind the surface
to sound, hard concrete.
10.5—Studded tire and tire chain wear on concrete
Abrasive materials, such as sand, are often applied to the
pavement surface when roads are slippery. Experience from
many years use of sand in winter, however, indicates that
this causes little wear if the concrete is of high quality and
the aggregates are wear-resistant.
Tire chains and studded snow tires, however, can cause
considerable wear to concrete surfaces, even where the
concrete is of high quality. Studded snow tires cause serious
damage, even to high-quality concrete. The damage is due
to the dynamic impact of the small tungsten carbide tip
of the studs, of which there are roughly 100 in each tire.
One laboratory study showed that studded tires running
on surfaces to which sand and salt were applied caused
100 times as much wear as tires without studs (Krukar and
Cook 1973). Fortunately, the use of studded tires has been
declining for a number of years or is forbidden by law in
certain jurisdictions.
Wear caused by studded tires is usually concentrated in
the wheel tracks. Ruts from 1/4 to 1/2 in. (6 to 12 mm) deep
can form in a single winter in regions where approximately
30 percent of passenger cars are equipped with studded tires
and traffic is heavy (Smith and Schonfeld 1970). More severe
wear occurs where vehicles stop, start, or turn (Keyser 1971).
Investigations have been made, principally in Scandanavia, Canada, and the United States, to examine the properties of existing concrete as related to studded tire wear
(Smith and Schonfeld 1971; Keyser 1971; Preus 1973;
Wehner 1966; Thurmann 1969). In some cases, there was
considerable variability in the data and the conclusions of
the different investigators were not in agreement; however,
most found that a hard coarse aggregate and a high-strength
mortar matrix are beneficial in resisting abrasion.
Another investigation was aimed at developing more
wear-resistant types of concrete overlays (Preus 1973).
Polymer cement concrete and polymer fly-ash concrete
provide better resistance to wear, although at the sacrifice of
skid resistance. Steel fiber concrete overlays were also tested
and showed reduced wear. Exposed fibers can adversely
affect the tire wear.
Although the reported test results show promise, no
affordable concrete surface has yet been developed that will
provide the same service life, when studded tires are used, as
concrete surfaces exposed to plain rubber tire wear. A report
(Brunette and Lundy 1996) summarizes available data on
pavement wear and on the performance and winter accident
records for studded tire use.
10.6—Skid resistance of pavements
The skid resistance of concrete pavements depends on
the surface texture of the concrete. There are two types of
surface texture:
1) Macrotexture from surface irregularities that are built in
at the time of construction
2) Microtexture from the type and hardness of fine aggregate
The microtexture is more important at speeds of less than
approximately 50 mph (80 km/h) (Kummer and Meyer
1967; Murphy 1975; Wilk 1978). At speeds greater than 50
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
mph (80 km/h), the macrotexture becomes quite important
because it is relied on to help prevent hydroplaning.
The skid resistance of concrete pavement initially depends
on the texture built into the surface layer (Dahir 1981). In time,
rubber-tire traffic abrades the surface paste, which removes
the beneficial macrotexture and exposes the coarse- and fineaggregate particles. The rate that the surface paste is removed
and the consequences on the skid resistance of a pavement
depends on the depth and quality of the surface paste and the
rock type (toughness) of the fine and coarse aggregate.
Fine aggregates containing significant amounts of silicate
minerals in the larger particle sizes will assist in slowing
down surface wear and maintaining the microtexture necessary for satisfactory skid resistance at slow speed (Fowler
and Rached 2012; Rado 2009). Certain rock types, however,
polish under rubber-tire wear. These include very finetextured limestone, dolomite, and serpentine; the finer the
texture, the more rapid the polishing. Where both the fine
and coarse aggregate are made of these rock types, there
may be a rapid polishing of the entire pavement surface and
a serious reduction in skid resistance. Where only the coarse
aggregate is of the polishing type, the problem is delayed
until the coarse aggregate is exposed by wear. However, if
the coarse aggregate is, for example, a coarse-grained silica
or vesicular slag, the skid resistance may increase when the
aggregate is exposed (Rado 2009).
Macrotexture is important because it prevents hydroplaning. An example of constructing macrotexture in pavement surfaces is placing grooves in the concrete—either
before hardening (tining) or by sawing after the concrete
has hardened—to provide channels for the escape of water
that is otherwise trapped between the tire and the pavement.
The spaces between grooves have to be especially resistant
to surface abrasion and frost action. A high-quality concrete
that is properly finished and cured has the required durability
and abrasion resistance (Ong and Fwa 2008).
10.7—Erosion
Concrete erosion is the progressive removal of mass
from the concrete surface from chemical attack, abrasion,
or cavitation (ACI 210R). Chemical attack erosion occurs
when components of the concrete paste or aggregate are
leached or dissolved. The rate of degradation can be significantly enhanced by flowing liquid due to the increased
rate of material removal and the maintenance of a low pH
near the concrete surface. Erosion by abrasion occurs when
suspended solids in the water impact or grind the surface,
causing material loss. Cavitation occurs when the local pressure in a hydraulic system drops below the liquid vaporization pressure, causing the liquid to vaporize and recondense.
As the liquid rapidly recondenses and the bubble/void phase
collapses, very high pressures are created on the concrete
wall, causing damage and material loss (ACI 210R).
This document contains a summary of the degradation
mechanism and concrete material properties that increase
the durability, while a more complete coverage of these
subjects is contained in ACI 210R. Erosion by chemical
attack is covered in more detail in Chapter 6 of this guide.
59
10.7.1 Abrasion—Suspended solids can abrade the
concrete surface uniformly, giving a rather smooth, worn
appearance. Factors that increase abrasion include high
water velocities; large, hard, sharp particles; high total
suspended solids content; long periods of exposure; and
concrete shape. Stilling basins, outlet works, locks, and
tunnel linings are common hydraulic structures that experience abrasion/erosion damage (ACI 210R; U.S. Bureau of
Reclamation 1997). Bridge abutments and other structures
placed in rivers or other flowing bodies are also candidates
for abrasion/erosion damage.
No concrete is completely immune to abrasion-related
erosion in hydraulic structures. High-quality pastes with a
dense microstructure are critical to make abrasion-resistant
concrete. Hard, abrasion-resistant concrete aggregates are
necessary. Larger coarse aggregates also help increase the
concrete resistance to abrasion erosion (Liu et al. 2006).
High-performance concrete is more resistant to abrasion
damage because of the higher-quality paste and aggregates. Concrete containing silica fume at low w/cm has
been found to increase the resistance to abrasion/erosion
damage. Polymer concrete has been shown to have excellent
resistance to abrasion/erosion (Klieger and Greening 1969;
Scrivener et al. 1999; U.S. Bureau of Reclamation 1997).
10.7.2 Cavitation—The most efficient way to prevent cavitation damage in hydraulic structures is to prevent cavitation.
This is best accomplished by considering cavitation in the
preliminary structural and hydraulic design by changing the
structure’s geometry, reducing surface irregularities, reducing
the flow rate, or by aeration (Frizell and Mefford 1991). Joints
should be avoided or minimized when possible because they
can increase turbulence. High-pressure, low-velocity systems
are less likely to experience cavitation. Strict construction
tolerances for concrete smoothness may be necessary to
reduce surface irregularities and localized turbulence.
Cavitation is such a damaging mechanism that no material
can be made cavitation proof. The service life of a structure can be extended by the use of more resistant materials.
Cavitation-resistant concrete has similar requirements as
abrasion-resistant concrete. Concrete mixture parameters
that increase the concrete abrasion erosion resistance apply
equally to resisting cavitation damage (MacDonald 2000).
Latex-modified concrete has also been shown to increase
cavitation resistance by increasing paste-aggregate bond
(MacDonald 2000).
For cavitation repair, the cause of cavitation should be
addressed, which may include the addition of aeration or
changing flow characteristics. When these methods are
impractical, the use of higher-quality, erosion-resistant
repair materials may prolong the service life of the structure. Repairs must follow strict tolerances on smoothness.
Silica fume concrete, epoxy-bonded concrete, or polymer
concrete may be used in the repair. Stainless steel surfaces
may also be used to armor the surface, although damage will
still occur. The use of polymer concrete, an epoxy coating,
or stainless steel also has the advantage of a smooth surface
that reduces turbulence (U.S. Bureau of Reclamation 1997).
American Concrete Institute – Copyrighted © Material – www.concrete.org
60
GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
CHAPTER 11—SUMMARY
The durability of concrete is one of the characteristics that
make it the most commonly used building material in the
world. This guide describes various factors that can influence
the durability of concrete, considering the particular mechanisms of deterioration in the context of the environmental
conditions to which the concrete is to be subjected. Strategies
are presented to increase the durability of concrete through
the use of appropriate materials and mixture proportions,
and emphasizes that appropriate placement practices and
workmanship are also essential to the production of durable
concrete in a given environmental exposure. Specifically,
this guide discusses the importance of concrete’s resistance
to fluid ingress as it influences durability and provides individual chapters on distress mechanisms including freezing
and thawing, alkali-aggregate reaction (AAR), sulfate attack,
aggressive chemical attack, physical salt attack, corrosion of
metals and other embedded materials, as well as abrasion.
For each of these mechanisms of distress, recommendations
are made for preventing or minimizing damage.
CHAPTER 12—REFERENCES
ACI committee documents and documents published by
other organizations are listed first by document number, full
title, and year of publication followed by authored documents listed alphabetically.
American Concrete Institute
ACI 207.1R-05(12)—Guide to Mass Concrete
ACI 207.2R-07—Report on Thermal and Volume Change
Effects on Cracking of Mass Concrete
ACI 210R-93(08)—Erosion of Concrete in Hydraulic
Structures
ACI 210.1R-94—Compendium of Case Histories
on Repair of Erosion-Damaged Concrete in Hydraulic
Structures
ACI 212.3R-10—Report on Chemical Admixtures for
Concrete
ACI 213R-14—Guide for Structural Lightweight-Aggregate Concrete
ACI 216.1-14—Code Requirements for Determining Fire
Resistance of Concrete and Masonry Construction Assemblies
ACI 221.1R-98(08)—Report on Alkali-Aggregate Reactivity
ACI 222R-01(10)—Protection of Metals in Concrete
against Corrosion
ACI 222.1R-96—Provisional Standard Test Method
for Water-Soluble Chloride Available for Corrosion of
Embedded Steel in Mortar and Concrete Using the Soxhlet
Extractor
ACI 222.2R-14—Report on Corrosion of Prestressing
Steels
ACI 222.3R-11—Guide to Design and Construction Practices to Mitigate Corrosion of Reinforcement in Concrete
Structures
ACI 223R-10—Guide for the Use of Shrinkage-Compensating Concrete
ACI 224R-01(08)—Control of Cracking in Concrete
Structures
ACI 224.1R-07—Causes, Evaluation, and Repair of
Cracks in Concrete Structures
ACI 301-16—Specifications for Structural Concrete
ACI 302.1R-15—Guide for Concrete Floor and Slab
Construction
ACI 304R-00(09)—Guide for Measuring, Mixing, Transporting, and Placing Concrete
ACI 305R-10—Guide to Hot Weather Concreting
ACI 306R-16—Guide to Cold Weather Concreting
ACI 308R-16—Guide to External Curing of Concrete
ACI 309R-05—Guide for Consolidation of Concrete
ACI 318-99—Building Code Requirements for Structural
Concrete and Commentary
ACI 318-14—Building Code Requirements for Structural
Concrete and Commentary
ACI 350-06—Code Requirements for Environmental
Engineering Concrete Structures and Commentary
ACI 350.1-10—Specification for Tightness Testing of
Environmental Engineering Concrete Containment Structures and Commentary
ACI 350.2R-04—Concrete Structures for Containment of
Hazardous Materials
ACI 357.1R-91(97)—Report on Offshore Concrete Structures for the Arctic
ACI 360R-10—Guide to Design of Slabs-on-Ground
ACI 365.1R-00—Service-Life Prediction
ACI 515.2R-13—Guide to Selecting Protective Treatments for Concrete
ASTM International
ASTM C33/C33M-16—Standard Specification for
Concrete Aggregates
ASTM C109/C109M-16—Standard Test Method for
Compressive Strength of Hydraulic Cement Mortars (Using
2-in. or [50-mm] Cube Specimens)
ASTM C114-15—Standard Test Methods for Chemical
Analysis of Hydraulic Cement
ASTM C150/C150M-16—Standard Specification for
Portland Cement
ASTM C227-10—Standard Test Method for Potential
Alkali Reactivity of Cement-Aggregate Combinations
(Mortar Bar Method)
ASTM C260/C260M-10a(2016)—Standard Specification
for Air-Entraining Admixtures for Concrete
ASTM C265-08—Standard Test Method for WaterExtractable Sulfate in Hydrated Hydraulic Cement Mortar
ASTM C289-07—Standard Test Method for Potential
Alkali-Silica Reactivity of Aggregates (Chemical Method)
(withdrawn 2016)
ASTM C295/C295M-12—Standard Guide for Petrographic Examination of Aggregates for Concrete
ASTM C309-11—Standard Specification for Liquid
Membrane-Forming Compounds for Curing Concrete
ASTM C418-12—Standard Test Method for Abrasion
Resistance of Concrete by Sandblasting
ASTM C441/C441M-11—Standard Test Method for
Effectiveness of Pozzolans or Ground Blast-Furnace Slag
American Concrete Institute – Copyrighted © Material – www.concrete.org
GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
in Preventing Excessive Expansion of Concrete Due to the
Alkali-Silica Reaction
ASTM C452/C452M-15—Standard Test Method for
Potential Expansion of Portland-Cement Mortars Exposed
to Sulfate
ASTM C457/C457M-12—Standard Test Method for
Microscopical Determination of Parameters of the Air-Void
System in Hardened Concrete
ASTM C494/C494M-16—Standard Specification for
Chemical Admixtures for Concrete
ASTM C563-16—Standard Test Method for Approximation of Optimum SO3 in Hydraulic Cement Using Compressive Strength
ASTM C586-11—Standard Test Method for Potential
Alkali Reactivity of Carbonate Rocks for Concrete Aggregates (Rock-Cylinder Method)
ASTM C595/C595M-15—Standard Specification for
Blended Hydraulic Cements
ASTM C618-15—Standard Specification for Coal Fly Ash
and Raw or Calcined Natural Pozzolan for Use in Concrete
ASTM C642-13—Standard Test Method for Density,
Absorption, and Voids in Hardened Concrete
ASTM C666/C666M-15—Standard Test Method for
Resistance of Concrete to Rapid Freezing and Thawing
ASTM C672/C672M-12—Standard Test Method for
Scaling Resistance of Concrete Surfaces Exposed to Deicing
Chemicals
ASTM C779/C779M-12—Standard Test Method for
Abrasion Resistance of Horizontal Concrete Surfaces
ASTM C845/C845M-12—Standard Specification for
Expansive Hydraulic Cement
ASTM C856-14—Standard Practice for Petrographic
Examination of Hardened Concrete
ASTM C944/C944M-12—Standard Test Method for
Abrasion Resistance of Concrete or Mortar Surfaces by the
Rotating-Cutter Method
ASTM C989/C989M-14—Standard Specification for
Slag Cement for Use in Concrete and Mortars
ASTM C1012/C1012M-15—Standard Test Method for
Length Change of Hydraulic-Cement Mortars Exposed to a
Sulfate Solution
ASTM C1017/C1017M-13—Standard Specification for
Chemical Admixtures for Use in Producing Flowing Concrete
ASTM C1038/C1038M-14—Standard Test Method for
Expansion of Hydraulic Cement Mortar Bars Stored in Water
ASTM C1105-08—Standard Test Method for Length
Change of Concrete Due to Alkali-Carbonate Rock Reaction
ASTM C1138-97—Standard Test Method for Abrasion
Resistance of Concrete (Underwater Method)
ASTM C1152/C1152M-04(2012)—Standard Test Method
for Acid-Soluble Chloride in Mortar and Concrete
ASTM C1157/C1157M-11—Standard Performance Specification for Hydraulic Cement
ASTM C1202-12—Standard Test Method for Electrical
Indication of Concrete’s Ability to Resist Chloride Ion
Penetration
ASTM C1218/C1218M-15—Standard Test Method for
Water-Soluble Chloride in Mortar and Concrete
61
ASTM C1240-15—Standard Specification for Silica
Fume Used in Cementitious Mixtures
ASTM C1260-14—Standard Test Method for Potential
Alkali Reactivity of Aggregates (Mortar-Bar Method)
ASTM C1293-08(2015)—Standard Test Method for
Determination of Length Change of Concrete Due to AlkaliSilica Reaction
ASTM C1315-11—Standard Specification for Liquid
Membrane-Forming Compounds Having Special Properties
for Curing and Sealing Concrete
ASTM C1524-02(2010)—Standard Test Method for
Water-Extractable Chloride in Aggregate (Soxhlet Method)
ASTM C1543-10—Standard Test Method for Determining
the Penetration of Chloride Ion into Concrete by Ponding
ASTM C1556-11—Standard Test Method for Determining the Apparent Chloride Diffusion Coefficient of
Cementitious Mixtures by Bulk Diffusion
ASTM C1567-13—Standard Test Method for Determining the Potential Alkali-Silica Reactivity of Combinations of Cementitious Materials and Aggregate (Accelerated
Mortar-Bar Method)
ASTM C1580-15—Standard Test Method for WaterSoluble Sulfate in Soil
ASTM C1585-13—Standard Test Method for Measurement of Rate of Absorption of Water by Hydraulic-Cement
Concretes
ASTM C1602/C1602M-12—Standard Specification for
Mixing Water Used in the Production of Hydraulic Cement
Concrete
ASTM C1747/C1747M-13—Standard Test Method for
Determining Potential Resistance to Degradation of Pervious
Concrete by Impact and Abrasion
ASTM C1778-16—Standard Guide for Reducing the Risk
of Deleterious Alkali-Aggregate Reaction in Concrete
American Association of State and Highway Transportation
Officials (AASHTO)
PP065-11—Standard Practice for Determining the Reactivity of Concrete Aggregates and Selecting Appropriate
Measures for Preventing Deleterious Expansion in New
Concrete Construction
T259-02—Standard Method of Test for Resistance of
Concrete to Chloride Ion Penetration
T277-15—Standard Method of Test for Electrical Indication of Concrete’s Ability to Resist Chloride Ion Penetration
CSA Group
CAN/CSA A3000-13—Cementitious Materials Compendium
CSA A23.1-14/CSA A23.2-14—Concrete Materials and
Methods of Concrete Construction/Test Methods and Standard Practices for Concrete
CAN/CSA
CSA A23.2-14A-14—Potential Expansivity of Aggregates; Procedure for Length Change Due to Alkali-Aggregate Reaction in Concrete Prisms
CSA A23.2-26A-14—Determination of Potential AlkaliCarbonate Reactivity of Quarried Carbonate Rocks by
Chemical Composition
American Concrete Institute – Copyrighted © Material – www.concrete.org
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
CSA A23.2-27A-14—Standard Practice to Identify Potential for Alkali-Reactivity of Aggregates and Measures to
Avoid Deleterious Expansion in Concrete
CSA A23.2-28A-14—Standard Practice for Laboratory
Testing to Demonstrate the Effectiveness of Supplementary
Cementing Materials and Lithium-Based Admixtures to
Prevent Alkali-Silica Reaction in Concrete
German Institute of Standardization
DIN 4030:2006-06—Assessment of Water, Soil, and
Gases for Their Aggressiveness to Concrete – Part 1: Principles and Limiting Values
U.S. Army Corps of Engineers
CRD-C 662:2010—Determining the Potential AlkaliSilica Reactivity of Combinations of Cementitious Materials, Lithium Nitrate Admixture and Aggregate (Accelerated Mortar-Bar Method)
EM 1110-2-2000:1994—Standard Practice for Concrete
for Civil Works Structures
Authored documents
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Abbas, A.; Carcasses, M.; and Ollivier, J.-P., 1999, “Gas
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Acker, P.; Foucrier, C.; and Malier, Y., 1986, “Temperature-Related Mechanical Effects in Concrete Elements and
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Alhassan, S. J., 2005, “Corrosion of Lead and Lead
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Alksnis, F. F., and Alksne, V. I., 1986, “Sur le role de la
phase siliceuse dans les processus de destruction de le pierre
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Janeiro, Brazil, pp. 170-174.
Almond, D. K., and Janssen, D. J., 1991, “The Washington
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to Freezing and Thawing,” Supplemental Papers, Second
CANMET/ACI International Conference on Durability of
Concrete, Montreal, QC, Canada, pp. 265-293.
Alonso, C.; Andrade, C.; Bacle, B.; and Fidjestøl, P.,
1992, “Corrosion de Armaduras en Microhormigones de
Humo de Silica Carbonatados (Corrosion of Reinforcement
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Andersson, K.; Allard, B.; Bengtsson, M.; and Magnusson,
B., 1989, “Chemical Composition of Cement Pore Solu-
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Andrade, C., 1993, “Calculation of Chloride Diffusion
Coefficients in Concrete from Ionic Migration Measurement,” Cement and Concrete Research, V. 23, No. 3, pp.
724-742. doi: 10.1016/0008-8846(93)90023-3
Andrade, C., and Whiting, D., 1996, “A Comparison of
Chloride Ion Diffusion Coefficients Derived from Concentration Gradients and Non-Steady State Accelerated Ionic
Migration,” Materials and Structures, V. 29, No. 8, pp.
476-484. doi: 10.1007/BF02486282
Ann, K. Y.; Jung, H. S.; Kim, H. S.; Kim, S. S.; and Moon,
H. Y., 2006, “Effect of Calcium Nitrite-Based Corrosion
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Anqi, L.; Baoyu, L.; Guoping, H.; Yeibo, C.; and Guolian,
S., 1991, “Study on Corrosion Prevention in Reinforced
Concrete Containing Condensed Silica Fume and Its Application,” Durability of Concrete, Proceedings of the Second
CANMET/ACI International Conference, SP-126, V. M.
Malhotra, ed., American Concrete Institute, Farmington
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AREMA, 2007, “Ties,” Chapter 30, Manual for Railway
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Asgeirsson, H., and Gudmundsson, G., 1979, “Pozzolanic
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Attiogbe, E. K.; Nmai, C. K.; and Gay, F. T., 1992, “AirVoid System Parameters and Freeze-Thaw Durability of
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Backstrom, J. E.; Burrows, R. W.; Mielenz, R. C.; and
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Backstrom, J. E.; Burrows, R. W.; Mielenz, R. C.; and
Wolkodoff, V. E., 1958b, “Origin, Evolution, and Effects of
the Air Void System in Concrete, Part 3—Influence of WaterCement Ratio and Compaction,” ACI Journal Proceedings,
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Backstrom, J. E.; Burrows, R. W.; Wolkodoff, V. E.;
and Powers, T. C., 1954, discussion of “Void Spacing as a
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of Tests and Properties of Concrete and Concrete-Making
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American Concrete Institute – Copyrighted © Material – www.concrete.org
GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
Bakker, R., 1980, “On the Cause of Increased Resistance
of Concrete Made from Blast Furnace Cement to the AlkaliSilica Reaction and to Sulfate Corrosion,” thesis, RWTH,
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Barona de la O, F., 1951, “Alkali-Aggregate Expansion Corrected with Portland-Slag Cement,” ACI Journal
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Bassuoni, M. T., and Nehdi, M. L., 2009, “Durability of
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Bastiensen, R.; Mourn, J.; and Rosenquist, I., 1957, “Some
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Bates, P. H., and Klein, A. A., 1917, “Properties of the
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Bates, P. H.; Phillips, A. J.; and Wig, R. J., 1913, “Action
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Bavarian, B., and Reiner, L., 2004, “Improving Durability
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Beaudoin, J. J., and Brown, P. W., 1992, “The Structure
of Hardened Cement Paste,” Proceedings of the 9th International Congress on the Chemistry of Cement, V. I, New
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Bellport, B. P., 1968, “Combating Sulphate Attack on
Concrete on Bureau of Reclamation Projects,” Performance
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Bentz, D. P., 2000, “Fibers, Percolation, and Spalling of
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Bentz, D. P.; Garboczi, E. J.; and Snyder, K. A., 1999, “A
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63
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Berke, N. S., 1985, “Effects of Calcium Nitrite and Mix
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Ash, Silica Fume, Slag, and Natural Pozzolans in Concrete:
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Berke, N. S., and Hicks, M. C., 1992, “Estimating the
Life Cycle of Reinforced Concrete Decks and Marine Piles
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Berke, N. S.; Hicks, M. C.; and Hoopes, R. J., 1994,
“Condition Assessment of Field Structures with Calcium
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Berke, N. S., and Roberts, L. R., 1989, “Use of Concrete
Admixtures to Provide Long-Term Durability from Steel
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Berke, N. S., and Rosenberg, A., 1989, “Technical Review
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Berke, N. S.; Scali, M. J.; Regan, J. C.; and Shen, D. F.,
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Fume and/or Fly Ash Containing Concretes,” Durability of
Concrete, Proceedings of the Second CANMET/ACI International Conference, SP-126, V. M. Malhotra, ed., American
Concrete Institute, Farmington Hills, MI, pp. 393-422.
Bérubé, M.-A., and Duchesne, J., 1992, “Evaluation
of Testing Methods Used for Assessing the Effectiveness
of Mineral Admixtures in Suppressing Expansion Due to
Alkali-Aggregate Reaction,” Fly Ash, Silica Fume, Slag,
American Concrete Institute – Copyrighted © Material – www.concrete.org
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GUIDE TO DURABLE CONCRETE (ACI 201.2R-16)
and Natural Pozzolans in Concrete, Proceedings of the
Fourth CANMET/ACI International Conference, SP-132, V.
M. Malhotra, ed., American Concrete Institute, Farmington
Hills, MI, pp. 549-575.
Bérubé, M.-A.; Duchesne, J.; Dorion, J.; and Rivest, M.,
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Aggregates to Concrete and Application to Concrete Structures Affected by Alkali-Silica Reactivity,” Cement and
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Bérubé, M.-A., and Fournier, B., 1992a, “Accelerated
Test Methods for Alkali-Aggregate Reactivity,” Advances in
Concrete Technology, V. M. Malhotra, ed., CANMET/EMR,
Ottawa, ON, Canada, pp. 583-627.
Bérubé, M.-A., and Fournier, B., 1992b, “Effectiveness
of the Accelerated Mortar Bar Method, ASTM C9 Proposal
P214 or NBRI, for Assessing Potential AAR in Quebec
(Canada),” Proceedings of the Ninth International Conference on Alkali-Aggregate Reaction in Concrete, Concrete
Society, Slough, UK, pp. 92-101.
Bérubé, M.-A., and Fournier, B., 1993, “Testing for
Alkali-Aggregate Reactivity in Concrete,” Memoria del
Seminario Internacional sobre Tecnología del Concreto,
R. Rivera Villareal, ed., Universidad Autónoma de Nuevo
León, Nuevo León, Mexico, Oct., pp. 54-78.
Bessey, G. E., and Lea, F. M., 1953, “The Distribution of
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Bhatty, M. S. Y., and Greening, N. R., 1978, “Interaction
of Alkalis with Hydrating and Hydrated Calcium Silicates,”
Proceedings of the Fourth International Conference on the
Effects of Alkalis in Cement and Concrete, Purdue University, West Lafayette, IND, pp. 87-112.
Bickley, J. A.; Hemmings, R. T.; Hooton, R. D.; and
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