Construction and Design
of Prestressed Concrete Segmental Bridges
Walter Podolny, Jr., Ph.D., P.E.
Bridge Division
Office of Engineering
Federal Highway Administration
U.S. Department of Transportation
Jean M. Muller
Chairman of the Board
Figg and Muller Engineers, Inc.
1982
A Wiley-Interscience Publication
John Wiley &? Sons
New
York
Chichester
Brisbane
Toronto
Singapore
zyxwvu
Series Preface
J
The Wiley Series of Practical Construction Guides
provides the Ivorking constructor \vith up-to-date
information that can help to increase the job profit
margin. These g uid ebo o ks, ivhich are sc aled
mainly for practice, but include the necessary
theory and design, should aid a construction contractor in approaching \+vork problems with more
knolvledgeable confidence. The guides should be
useful also to engineers, architects, planners,
specification tvriters, project managers, superintendents. materials and equipment manufacturers
and. the source of all these callings, instructors and
their students.
Construction in the United States alone will
reach $250 billion a year in the early 1980s. In all
nations. the business of building will continue to
grow at a phenomenal rate, because the population
proliferation demands new living, lvorking, and
recreational facilities. This construction will have
to be more substantial, thus demanding a more
professional performance from the contractor. Before science and technology had seriously affected
the ideas, job plans, financing, and erection of
structures. most contractors developed their
knolv-holy by field trial-and-error. Wheels, small
and large. jvere constantly being reinvented in all
sectors, because there was no interchange of
knolvledge.
The current complexity of construction, even in more rural areas, has revealed a clear
need for more proficient, professional methods
and tools in both practice and learning.
Because construction is highly competitive, some
practical technology is necessarily proprietary. But
most practical day-to-day problems are common to
the Fvhole construction industry. These are the
subjects for the Wiley Practical Construction
Guides.
M. D. MORRIS , P.E.
zyxwvut
Preface
J
Prestressed concrete segmental bridge construction has evolved, in the natural course of events,
from the combining of the concepts of prestressing, box girder design, and the cantilever method
of bridge construction. It arose from a need to
overcome construction difficulties in spanning
deep valleys and river crossings without the use of
conventional falsework, which in some instances
may be impractical, economically prohibitive, or
detrimental to environment and ecology.
Contemporary prestressed, box girder, segmental bridges began in Western Europe in the
1950s. Ulrich Finsterw alder in 1950, for a crossing of the Lahn River in Balduinstein, Germany,
was the first to apply cast-in-place segmental construction to a bridge. In 1962 in France the first
application of precast, segmental, box girder construction was made by Jean Muller to the ChoisyLe-Roi Bridge crossing the Seine River. Since then
the concept of segmental bridge construction has
been improved and refined and has spread from
Europe throughout most of the world.
The first application of segmental bridge construction in North America was a cast-in-place
segmental bridge on the Laurentian Autoroute
near Ste. Adele, Quebec, in 1964. This was followed in 1967 by a precast segmental bridge crossing the Lievre River near Notre Dame du Laus,
Quebec. In 1973 the first U.S. precast segmental
bridge was opened to traffic in Corpus Christi,
Texas, followed a year later by the cast-in-place
segmental Pine Valley Bridge near San Diego,
California. As of this date (1981) in the United
States more than eighty segmental bridges are
completed, in construction, in design, or under
consideration.
Prestressed concrete segmental bridges may be
identified as precast or cast in place and categorized by method of construction as balanced
cantilever, span-by-span, progressive placement,
or incremental launching. This type of bridge has
extended the practical and competitive economic
span range of concrete bridges. It is adaptable to
almost any conceivable site condition.
The objective of this book is to summarize in one
volume the current state of the art of design and
construction methods for all types of segmental
bridges as a ready reference source for engineering faculties, practicing engineers, contractors, and
local, state, and federal bridge engineers.
Chapter 1 is a quick review of the historical evolution to the current state of the art. It offers the
student an appreciation of the way in which segmental construction of bridges developed, the
factors that influenced its development, and the
various techniques used in constructing segmental
bridges.
Chapters 2 and 3 present case studies of the predominant methodology of constructing segmental
bridges by balanced cantilever in both cast-in-place
and precast concrete. Conception and design of
the superstructure and piers, respectively, are discussed in Chapters 4 and 5. The other three basic methods of constructing segmental bridgesprogressive placement, span-by-span, and incremental launching-are presented in Chapters 6
and 7.
Chapters 2 through 7 deal essentially with girder
type bridges. However, segmental construction
may also be applied to bridges of other types.
Chapter 8 discusses application of the segmental
concept to arch, rigid frame, and truss bridges.
Chapter 9 deals with the cable-stayed type of
bridge and Chapter 10 with railroad bridges. The
practical aspects of fabrication, handling, and
erection of segments are discnssed
in Chapter 11.
In selected a bridge type for a particular site, one
of the more important parameters is economics.
Economics, competitive bidding, and contractual
aspects of segmental construction are discussed in
Chapter 12.
Most of the material presented in this book is not
vii
Preface
original: Although acknowledgment of all the
many.source$&. not possible, full credit is given
wherever the specific so;rce can be identified.
Every effort has been. made to eliminate errors;
the authors will appreciate notification from the
reader ‘of any that remain.
The authors are indebted to numerous publications, o rganizatio ns,
and individuals for their
assistance and permission to reproduce photo-
graphs, tables, and other data. Wherever possible,
credit is given in the text.
WALTER PODOLNY,
JEAN M. MUILEK
Burke, Virginia
Par%, Francr
Jarmar? 1982
JK.
Contents
1
Prestressed Concrete Bridges and
Segmental Construction
1.1
1.2
1.3
1.4
1.5
1.6
1.7
1.8
1.9
1.10
1.11
2
Introduction, 1
Development of Cantilever
Construction, 2
Evolution of Prestressed
Concrete, 4
Evolution of Prestressed Concrete
Bridges, 5
Long-Span Bridges with
Conventional Precast
Girders, 8
Segmental Construction, 10
Various Types of Structures, 12
Cast-in-Place and Precast
Segmental Construction, 17
Various Methods of
Construction, 18
Applications of Segmental
Construction in the United
States, 26
Applicability and Advantages of
Segmental Construction, 28
References, 30
Cast-In-Place Balanced Cantilever Girder
Bridges
2.1
2.2
2.3
2.4
2.5
2.6
2.7
Introduction, 3 1
Bendorf Bridge, Germany, 35
Saint Adele Bridge, Canada, 37
Bouguen Bridge in Brest and
Lacroix Falgarde Bridge,
France, 38
Saint Jean Bridge over the
Garonne River at Bordeaux,
France, 4 1
Siegtal and Kochertal Bridges,
Germany, 43
Pine Valley Creek Bridge,
U.S.A., 46
2.8
2.9
2.10
2.11
2.12
1
2.13
2.14
2.15
2.16
3
Precast Balanced Cantilever Girder
Bridges
3.1
3.2
3.3
3.4
31
Gennevilliers Bridge, France, 52
Grand’Mere Bridge, Canada, 55
Arnhem Bridge, Holland, 58
Napa River Bridge, U.S.A., 59
Koror-Babelthuap, U.S. Pacific
Trust Territory, 61
Vejle Fjord Bridge,
Denmark, 63
Houston Ship Channel Bridge,
U.S.A., 68
Other Notable Structures, 71
Co nclusio n, 8 1
References, 8 1
3.5
3.6
3.7
3.8
3.9
3.10
3.11
3.12
3.13
3.14
3.15
3.16
82
Introduction, 82
Choisy Le Roi Bridge and Other
Structures in Greater Paris,
France, 83
Pierre Benite Bridges near Lyons,
France, 89
Other Precast Segmental Bridges
in Paris, 91
Oleron Viaduct, France, 96
Chillon Viaduct, Switzerland, 99
Hartel Bridge, Holland, 103
Rio-Niteroi Bridge, Brazil, 106
Bear River Bridge, Canada, 108
JFK Memorial Causeway,
U.S.A., 109
Saint Andre de Cubzac Bridges,
France, 113
Saint Cloud Bridge, France, 114
Sallingsund Bridge,
Denmark, 122
B-3 South Viaducts, France, 124
Alpine Motorway Structures,
France, 129
Bridge over the Eastern Scheldt,
Holland, 134
ix
X
3.17 Captain Cook Bridge,
A ustralia, 136
3.18 Other Notable Structures, 139
References, 147
4
Design of Segmental Bridges
4.1
4.2
4.3
4.4
4.5
4.6
4.7
4.8
4.9
4.10
4.11
4.12
4.13
4.14
4.15
4.16
4.17
4.18
4.19
5
5.4
5.6
5.7
Introduction, 225
Loads Applied to the Piers, 230
Suggestions on Aesthetics of Piers
and Abutments, 232
Moment-Resisting Piers and
Their Foundations, 234
5.8
5.9
148
Introduction, 148
Live Load Requirements, 149
Span Arrangement and Related
Principle of Construction, 149
Deck Expansion, Hinges, and
Co ntinuity , 15 1
Type, Shape and Dimensions of
the Superstructure, 159
Transverse Distribution of Loads
Between Box Girders in Multibox
Girders, 164
Effect of Temperature Gradients
in Bridge Superstructures, 170
Design of Longitudinal Members
for Flexure and Tendon
Profiles, 173
Ultimate Bending Capacity of
Longitudinal Members, 190
Shear and Design of Cross
Section, 193
Joints Between Match-Cast
Segments, 199
Design of Superstructure Cross
Section, 202
Special Problems in
Superstructure Design, 203
Deflections of Cantilever Bridges
and Camber Design, 205
Fatigue in Segmental
Bridges, 2 10
Provisions for Future
Prestressing, 2 12
Design Example, 2 12
Quantities of Materials, 219
Potential Problem Areas, 220
References, 224
Foundations, Piers, and Abutments
5.1
5.2
5.3
5.5
Piers with Double Elastomeric
Bearing s, 24 1
Piers with Twin Flexible Legs, 253
Flexible Piers and Their Stability
During Construction, 263
A butments, 27 1
Effect of Differential Settlements
on Continuous Decks, 276
References, 280
6 Progressive and Span-by-Span
Construction of Segmental Bridges
6.1
6.2
6.3
6.4
6.5
6.6
Introduction, 281
Progressive Cast-in-Place
Bridges, 283
Progressive Precast Bridges, 289
Span-by-Span Cast-in-Place
Bridges, 293
Span-by-Span Precast
Bridges, 308
Design Aspects of Segmental
Progressive Construction, 3 14
References, 3 19
7 Incrementally Launched Bridges
7.1
7.2
7.3
7.4
7.5
7.6
7.7
7.8
7.9
7.10
8
225
8.3
8.4
8.5
8.6
32
Introduction, 32 1
Rio Caroni, Venezuela, 323
Val Restel Viaduct, Italy, 327
Ravensbosch Valley Bridge,
Holland, 329
Olifant’s River Bridge, South
Africa, 33 1
Various Bridges in France, 333
Wabash River Bridge, U.S.A., 335
Other Notable Bridges, 338
Design of Incrementally
Launched Bridges, 343
Demolition of a Structure by
Incremental Launching, 352
References, 352
Concrete Segmental Arches, Rigid
Frames, and Truss Bridges
8.1
8.2
2,
Introduction, 354
Segmental Precast Bridges over
the Marne River, France, 357
Caracas Viaducts, Venezuela, 363
Gladesville Bridge, Australia, 37 1
Arches Built in Cantilever, 374
Rigid Frame Bridges, 382
35
Contents
8.7
11
Truss Bridges, 392
References, 399
9.3
9.4
9.5
9.6
9.7
9.8
9.9
9.10
Introduction, 400
Lake Maracaibo Bridge,
Venezuela, 405
Wadi Kuf Bridge, Libya, 407
Chaco/ Corrientes Bridge,
Argentina, 408
Mainbrticke, Germany, 410
Tie1 Bridge, Netherlands, 412
Pasco-Kennewick Bridge,
U.S.A., 418
Brotonne Bridge, France, 419
Danube Canal Bridge,
A ustria, 427
Notable Examples of
Concepts, 430
References, 439
10 Segmental Railway Bridges
10.1
10.2
10.3
10.4
10.5
10.6
10.7
10.8
10.9
10.10
10.11
Introduction to Particular
Aspects of Railway Bridges and
Field of Application, 44 1
La Voulte Bridge over the
Rhone River, France, 442
Morand Bridge in Lyons,
France, 442
Cergy Pontoise Bridge near
Paris, France, 444
Marne La Vallee and Torcy
Bridges for the New Express
Line near Paris, France, 444
Clichy Bridge near Paris,
France, 449
Olifant’s Bridge, South
Africa, 452
Incrementally Launched
Railway Bridges for the
High-Speed Line, Paris to
Lyons, France, 453
Segmental Railway Bridges in
Japan, 457
Special Design Aspects of
Segmental Railway Bridges, 458
Proposed Concepts for Future
Segmental Railway Bridges, 464
Technology and Construction of
Segmental Bridges
11.1
11.2
9 Concrete Segmental Cable-Stayed Bridges 400
9.1
9.2
xi
11.3
1.1.4
11.5
11.6
11.7
11.8
441
12
Scope and Introduction, 465
Concrete and Formwork for
Segmental Construction, 466
Post-tensioning Materials and
Operations, 470
Segment Fabrication for
Cast-In-Place Cantilever
Construction, 475
Characteristics of Precast
Segments and Match-Cast
Epoxy Joints, 485
Manufacture of Precast
Segments, 493
Handling and Temporary
Assembly of Precast
Segments, 507
Placing Precast Segments, 509
References, 5 17
Economics and Contractual Aspects of
Segmental Construction
12.1
12.2
12.3
13.4
13.5
13.6
Index
Index
Index
Index
518
Bidding Procedures, 5 18
Examples of Some Interesting
Biddings and Costs, 523
Increase in Efficiency in
Concrete Bridges, 528
References, 535
13 Future Trends and Develofnnents
13.1
13.2
13.3
465
536
Introduction, 536
Materials, 536
Segmental Application to
Bridge Decks, 542
Segmental Bridge Piers and
Substructures, 543
Application to Existing or New
Eridge Types, 544
Summary, 548
References, 549
of Bridges
of Personal Names
of Firms and Organizations
of Subjects
551
555
557
559
Construction and Design
of Prestressed Concrete
Segmental Bridges
1
Prestressed Concrete Bridges
and Segmental Construction
1.1
INTRODUCI’ION
1 . 2 DEVELOPMENT OF CANTILEVER CONSTRUCITON
1 . 3 EVOLUTION OF PRESTRESSED
CONCRETE
1 . 4 EVOLUTION OF PRESTRESSED CONCRETE BRIDGES
1.5 LONGSPAN BRIDGES WITH CONVENTIONAL PRECAST GIRDERS
1.6 SEGMENTAL CONSTRUCTION
1 . 7 VARIOUS TYPES OF STRUCl-URFS
1.7.1
1.8.1
1.8.2
1.8.3
1.9
Girder Bridges
1.7.2 Trusses
1.7.3 Frames with Slant Legs
1.7.4 Concrete Arch Bridges
1.7.5 Concrete CabkStayed Bridges
1.8 CAST-IN-PLACE AND PRECAST SEGMENTAL CONSTRUCTION
1 . l Zntroduction,
The conception, development, and worldwide acceptance of,segmental construction in the field of
prestressed concrete bridges represents one of the
most interesting and important achievements in
civil engineering during the past thirty years. Recognized today in all countries and particularly in
the United States as a safe, practical, and economic
construction method, the segmental concept probably owes its rapid growth and acceptance to its
founding, from the beginning, on sound construction principles such as cantilever construction.
Using this method, a bridge structure is made up
of concrete elements usually called segments
(either precast or cast in place in their final position
in the structure) assembled by post-tensioning. If
the bridge is cast in place, Figure 1.1, travelers are
used to allow the various segments to be constructed in successive increments and progressively
1.10
1.11
Characteristics of Cast-in-Place Segments
Characteristics of Precast Segfnents
Choice between Cast-in-Place and Precast
Construction
VARIOUS METHODS OF CONSTRUCTION
1.9.1 Cast-in-Place Balanced Cantilever
1.9.2 Precast Balanced Cantilever
1.9.3 Span-by-Span Construction
1.9.4 Progressive Placement Construction
1 . 9 . 5 Incremental Launching or Push-Out Construction
APPLICATIONS OF SEGMENTAL CONSTRUCTION
IN THE UNITED STATES
APPLICABILITY AND ADVANTAGES OF SEGMENTAL CONSTRUCI’ION
REFERENCES
prestressed together. If the bridge is precast, segments are manufactured in a special casting yard
or factory, transported to their final position, and
placed in the structure by various types of launch-
FIGURE 1.1 Cast-in place form traveler.
1
Prestressed
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Concrete Stidges ad Segmental Constrt4ction
FIGURE 1.2. Oleron Viaduct, segmental construction in progress. One typical
precast segment placed in the Oleron Viaduct.
ing equipment, Figure 1.2, while prestressing
achieves the assembly and provides the structural
strength.
Most early segmental bridges were built as cantilevers, where construction proceeds in a symmetrical fashion from the bridge piers in successive increments to complete each span and finally the
entire superstructure, Figure 1.3. Later, other construction methods appeared in conjunction with
______
Llzcr---/.#-------. ,% --------------/-------l-r
-77
-------------.------3-r
FIGURE 1.3. Cantilever construction applied to prestressed concrete bridges.
the segmental concept to further its field of application.
1.2
Development of Cantilever Construction
The idea of cantilever construction is ancient in the
Orient. Shogun’s Bridge located in the city of
Nikko, Japan, is the earliest recorded cantilever
bridge and dates back to the fourth century. The
Wandipore Bridge, Figure 1.4, was built in the
seventeenth century in Bhutan, between India and
Tibet. It is constructed from great timbers that
are corbeled out toward each other from massive abutments and the narrowed interval finally
capped with a light beam.’
FIGURE 1.4. Wandipore Bridge.
Develofwnent
of Cantilever Construction
3
That half an arc should stand upon the
ground
Without support while building, or a rest;
This caus’d the theorist’s rage and sceptic’s
jest.
Prefabrication techniques were successfully
combined with cantilever construction in many
bridges near the end of the nineteenth century, as
exemplified by such notable structures as the Firth
of Forth Bridge, Figure 1.6, and later the Quebec
Bridge, Figure 1.7, over the Saint Lawrence River.
These structures bear witness to the engineering
genius of an earlier’ generation. Built more recently, the Greater New Orleans Bridge over the
Mississippi River, Figure 1.8, represents modern
contemporary long-span steel cantilever construction.
Because the properties and behavior of prestressed concrete are related more closely to those
of structural steel than those of conventional reinforced concrete, the application of this material to
cantilever construction w as a logical step in the
continuing development of bridge engineering.
FIGURE 1.7. Quebec Bridge.
FIGURE 1.8. Greater New Orleans Bridge.
Prestressed Concrete Bridges and Segmental Construction
4
This application has evolved over many years by
the successive development of many concepts and
innovations. In order to see how the present state
of the art has been reached, let us briefly trace the
development of prestressed concrete and in particular its application to bridge construction.
1.3
Evolution
of
Prestressed
Concrete
The invention of reinforced concrete stirred the
imagination of engineers in many countries. They
envisioned that a tremendous advantage could be
achieved, if the steel could be tensioned to put the
structure in a permanent state of compression
greater than any tensile stresses generated by the
applied loads. The present state of the art of prestressed concrete has evolved from the effort and
experience of many engineers and scientists over
the past ninety years. However, the concept of prestressing is centuries old. Swiss investigators have
show n that as early as 2700 B . C . the ancient Egyptians prestressed their seagoing vessels longitudinally. This has been determined from pictorial
representations found in Fifth Dynasty tombs.
The basic principle of prestressing was used in
the craft of cooperage when the cooper wound
ropes or metal bands around wooden staves to
form barrels.3 When the bands were tightened,
they were under tensile prestress, which created
compression between the staves and enabled them
to resist hoop tension produced by internal liquid
pressure. In other words, the bands and staves
were both prestressed before they were subjected
to any service loads. The wooden cartwheel with its
shrunk-on iron rim is another example of prestressed construction.
The first attempt to introduce internal stresses in
reinforced concrete members by tensioning the
steel reinforcement was made about 1886 when
P. H. Jackson, an engineer in San Francisco, obtained
a United States patent for tightening steel rods in
concrete members serving as floor slabs. In l&S,
C. E. W. DGhring of Berlin secured a patent for the
manufacture of slabs, battens, and small beams for
structural engineering purposes by embedding
tensioned wire in concrete in order to reduce
cracking. This was the first attempt to provide precast concrete units with a tensioned reinforcement.
Several structures were constructed using these
concepts; however, only mild steel reinforcement
was available at the time. These structures at first
behaved according to predictions, but because so
little prestress force could be induced in the mild
steel, they lost their properties because of the creep
and shrinkage of the concrete. In order to recover
some of the losses, the possibility of retightening
the reinforcing rods after some shrinkage and
creep o f the c o nc rete had taken p lac e w as
suggested in 1908 by C. R. Steiner of the United
States. Steiner proposed that the bond of embedded steel bars be destroyed by lightly tensioning the bars while the concrete was still young and
then tensioning them to a higher stress when the
concrete had hardened. Steiner was also the first to
suggest the use of curved tendons.
In 1925, R. E. Dill of Nebraska took a further
step toward freeing concrete beams of any tensile
stresses by tensioning high-tensile steel wires after
the concrete had hardened. Bonding was to be
prevented by suitably coating the wires. He
explicitly mentioned the advantage of using steel
with a high elastic limit and high strength as compared to ordinary reinforcing bars.
In 1928, E. Freyssinet of France, who is credited
with the modern development of prestressed concrete, started using high-strength steel wires for
prestressing. Although Freyssinet also tried the
method of pretensioning, where the steel was
bonded to the concrete without end anchorages,
the first practical application of this method was
made by E. Hoyer about 1938. Wide application of
the prestressing technique was not possible until
reliable and economical methods of tensioning and
end anchorage were devised. From approximately
1939 on, E. Freyssinet, Magnel, and others developed different methods and procedures. Prestress began to gain some importance about 1945,
while alternative prestressing methods were being
devised by engineers in various countries.
During the past thirty years, prestressed concrete in the United States has grown from a
brand-new idea into an accepted method of concrete construction. This growth, a result of a new
application of existing materials and theories, is in
itself phenomenal. In Europe the shortage of materials and the enforced economies in construction
gave prestressed concrete a substantial start. Development in the United States, however, was
slower to get underway. Designers and contractors
hesitated mainly because of their lack of experience and a reluctance to abandon more familiar
methods of construction. Contractors, therefore,
bid the first prestressed concrete work conservatively. Moreover, the equipment available for prestressing and related techniques was essentially
new and makeshift. However, experience was
gained rapidly, the quality of the work improved,
Evolution
of
Prestressed
Concrete
Bridges
5
FIGURE 1.9. Freyssinet’s Esblv Bridge on the Marne
River.
and prestressed concrete became more and more
competitive with other materials.
1.4
Evolution of Prestressed Concrete Bridges
Although France took the lead in the development
of prestressed concrete, many European countries
such as Belgium, England, Germany, Switzerland,
and Holland quickly showed interest. As early as
1948, Freyssinet used prestressed concrete for the
construction of five bridges over the Marne River
near Paris, w ith 240 ft (74 m) spans of an exceptionally light appearance, Figure 1.9. A survey
made in Germany showed that between 1949 and
1953, out of 500 bridges built, 350 were prestressed.
FI GUR E 1 . 1 0 W a l n u t L a n e B r i d g e , Phil,~dcll~hia
(courtesy of the Portland Cement Association).
Prestressing in the United States followed a different course. Instead of linear prestressing, circular prestressing as applied to storage tanks took
the lead. Linear prestressing as applied to beams
did not start until 1949. The first structure of this
type was a bridge in Madison County, Tennessee,
followed in 1950 by the well-known 160 ft (48.80
m) span Walnut Lane Bridge in Philadelphia, Figure 1.10. By the middle of 1951 it was estimated
that 175 bridges and 50 buildings had been constructed in Europe and no more than 10 structures
in the United States. In 1952 the Portland Cement
Association conducted a survey in this country
showing 100 or more structures completed or
FIGURE 1.11. AASHTO-PC1 I-girder cross sections.
6
Prestressed
Concrete
Bridges
under construction. In 1953 it was estimated that
there were 75 bridges in Pennsylvania alone.
After the Walnut Lane Bridge, which was cast in
place and post-tensioned, precast pretensioned
bridge girders evolved, taking advantage of the inherent economies and quality control achievable
with shop-fabricated members. With few exceptions, during the 1950s and early 196Os, most multispan precast prestressed bridges built in the
United States were designed as a series of simple
spans. T h e y w e r e d e s i g n e d w i t h s t a n d a r d
AASHTO-PCI* girders of various cross sections,
Figure 1.11, for spans of approximately 100 ft
(30.5 m), but more commonly for spans of 40 to 80
ft (12 to 24 m). The advantages of a continuous
cast-in-place structure were abandoned in favor of
t h e s i m p l e r c o n s t r u c t i o n o f f e r e d b y plantproduced standardized units.
At this time, precast pretensioned members
found an outstanding application in the Lake
Pontchartrain crossing north of New Orleans,
Louisiana. The crossing consisted of more than
2200 identical 56 ft (17 m) spans, Figures 1.12
through 1.14. Each span was made of a single 200
ton monolith with pretensioned longitudinal gird*American Association of State Highway and Transportation
Officials (previously known as AASHO, American Association
of State Highway Officials) and Prestressed Concrete Institute.
and
Segmental
Construction
FIGURE 1.12. Lake Pontchartrain Bridge, U.S.A.
ers and a reinforced concrete deck cast integrally,
resting in turn on a precast cap and two prestressed spun piles. The speed of erection was incredible, often more than eight complete spans
placed in a single day.
In the middle 1960s a growing concern was
shown about the safety of highways. The
AASHTO Traffic Safety Committee called in a
1967 report 4 for the “ . . . adoption and use of twospan bridges for overpasses crossing divided highways . . . to eliminate the bridge piers normally
placed adjacent to the shoulders,” Figure 1.15. Interstate highways today require overpasses with
two, three, and four spans of up to 180 ft (54.9 m)
or longer. In the case of river or stream crossings,
FIGURE 1.13. Lake Pontchartrain Bridge, U.S.A.
33'4
18'4'
I
I
zyxwvu
I
(b)
FIGURE 1.14.
v erse sectio n.
Lake Pontchartrain Bridge, U.S.A. (a) Longitudinal section. (b) Trans7
-
bestressed Concrete Bridges and Segmental Construction
8
STANMRD 4-SPAN INTERSTATE CROSSING
I
tg
177’
250’
FIGURE 1.15. Standard four-span interstate crossing
(courtesv
of the Portland Cement Association).
longer spans in the range of 300 ft (91.5 m) or
longer may be required, and there is a very distinct
trend toward longer-span bridges. It soon became
apparent that the conventional precast pretensioned AASHTO-PC1 girders were limited by their
transportable length and weight. Transportation
over the highways limits the precast girder to a
length of 100 to 120 ft (30.5 to 36.6), depending
upon local regulations.
I .5
Long-Span Bridges with
Precast Girders
Conventional
As a result of longer span requirements a study was
conducted by the Prestressed Concrete Institute
(PCI) in cooperation with the Portland Cement Association (PCA).S This study proposed that simple
spans up to 140 ft (42.7 m) and continuous spans
up to 160 ft (48.8 m) be constructed of standard
precast girders up to 80 ft (24 m) in length joined
by splicing. To obtain longer spans the use of inclined or haunched
piers was proposed.
The follow ing discussion and illustrations are
based on the grade-separation studies conducted
by PC1 and PCA. Actual structures will be illus-
trated, where possible, to emphasize the particular
design concepts.
The design study illustrated in Figure 1.16 uses
cast-in-place or precast end-span sections and a
two-span unit with AASHTO I girders.6 Narrow
median piers are maintained in this design, but the
abutments are extended into the spans by as much
as 40 ft (12 m) using a precast or cast-in-place
frame in lieu of a closed or gravity abutment.
When site conditions warrant, an attractive type of
bridge can be built with extended abutments.
A similar span-reducing concept is developed in
Figure 1.17, using either reinforced or prestressed
concrete for cantilever abutments. An aesthetic
abutment design in reinforced concrete was developed for a grade-separation structure on the
Trans-Canada Highway near Drummondville in
the Province of Quebec, Figure 1.18. This provided a 324 ft (9.9 m) span reduction that led to the
use of type IV Standard AASHTO I girders to
span 974 ft (29.7 m) to a simple, narrow median
pier.
A cast-in-place reinforced concrete frame with
outward-sloping legs provides a stable, center supporting structure that reduces span length by 29 ft
(8.8 m), Figure 1.19. This enables either standard
box sections or I sections 84 ft (25.6 m) long to be
used in the tw o main spans. This layout w as used
for the Hobbema Bridge in Alberta, B.C., Canada,
shown in Figure 1.20. This bridge was built with
precast channel girder sections, but could be built
with AASHTO I girders or box sections. The median frame w ith inclined legs w as cast in place.
The schematic and photograph in Figures 1.21
and 1.22 show the Ardrossan Overpass in Alberta.
It is similar to the Hobbema Bridge except that the
spans are longer and, with the exception of a
cast-in-place footing, the median frame is made up
of precast units post-tensioned together, Figure
1.21. The finished bridge, Figure 1.23, has a
zyxwvutsr
Carl-in-place Froma
SECTION
A -A
FIGURE 1.16. Extended abutments (courtesy of the Prestressed Concrete Institute,
from ref. 6).
Long-Span
APPROX .
Bridges with
36’
Conventional
Precast
9
Girders
A P P R O X . * I ’-
ELEV AT I ON
r;
I S ’-0 ”
I t s’-0 ”
T Y P E lx
A A S H O OlROt
OI ROER
R
SECTION
FIGURE 1.1’7. Cantilevered
Institute, from ref. 6).
\\,\_ \ \\\
\\\\
\
,, \ \ \\
\\
\
\
abutments
\
FIGURE 1.18. Drummondville Bridge (courtesy of
the Portland Cement Association).
(courtesy
of
the
Prestressed
Concrete
pleasing appearance. The standard units w ere
channel-shaped stringers 64 in. w ide and 41 in.
deep (1.6 m by 1.04 m). The use of precast units
allowed erection of the entire superstructure, ineluding the median frame, in only three weeks.
The bridge was opened to traffic just eleven weeks
after construction began in the early summer of
1966.
By use o f tem p o rary bents, Fig ure 1.24,
standard units 60 ft (18.3 m) long can be placed
over the median pier and connected to main span
units with cast-in-place reinforced concrete splices
located near the point of dead-load contraflexure.
ELEV AT I ON
S E C T I O N S A -A
FIGURE 1.19. Median frame cast in place (courtesy of the Prestressed
Concrete Institute, from ref. 6).
10
Prestressed
Concrete
Bridges
and
Segmental
Construction
from the side pier over the main pier to the hingesupport for the suspended span.
The type of construction that uses long,
standard, precast, prestressed units never quite
achieved the recognition it deserved. As spans increased, designers turned toward post-tensioned
cast-in-place box girder construction. The California Division of Highways, for example, has been
quite successful with cast-in-place, multicell, posttensioned box girder construction for multispan
structures with spans of 300 ft (91.5 m) and even
longer. However, this type of construction has its
own limitations. The extensive formwork u s e d
during casting often has undesirable effects on the
environment or the ecology.
FIGURE 1.20. Hobbema Bridge, completed structure
(courtesy of the Portland Cement Association).
1.6 Segmental Construction
This design is slightly more expensive than previ-
ous ones but it provides the most open type twospan structure.
The structural arrangement of the Sebastian
Inlet Bridge in Florida consists of a three-span unit
over the main channel, Figure 1.25. The end span
of this three-span unit is 100 ft (30.5 m) long and
cantilevers 30 ft (9 m) beyond the piers to support
a 120 ft (36.6 m) precast prestressed drop-in span,
Figure 1.26. The end-span section was built in two
segments with a cast-in-place splice with the help of
a falsework bent. The Napa River Bridge at Vallejo, California (not to be confused with the Napa
River Bridge described in Section 2.1 l), used a
precast concrete cantilever-suspended span concept similar to the Sebastian Inlet Bridge, at about
the same time. The only difference was that the
cantilever girder was a single girder extending
ELE
V A
Segmental construction has been defined’ as a
method of construction in which primary loadsupporting members are composed of individual
members called segments post-tensioned together.
The concepts developed in the PCI-PCA studies
and described in the preceding section come under
this definition, and we might call them “longitudinal” segmental construction because the individual
elements are long with respect to their width.
In Europe, meanwhile, segmental construction
proceeded in a slightly different manner in conjunction with box girder design. Segments were
cast in place in relatively short lengths but in fullrpadway width and depth. Today segmental construction is usually understood to be the type developed in Europe. However, as will be shown
later, the segments need not be of full-roadway
T
I O N
81p-40
AASHO-PCI
BOX SECTION
3’-6”
6’-6*
zyxwvutsrqponmlkjihgfedcbaZYXW
b’-6”
& X -I ON S A - A
FIGURE
1.21.
Median frame precast (courtesy of the Prestressed Cot xrete
Institute, from ref. 6).
Segmental Construction
FIGURE 1.22. Ardrossan Overpass precast median
frame (courtesy of the Portland Cement Association).
width and can become rather long in the longitudinal direction of the bridge, depending on the
construction system utilized.
Eugene Freyssinet, in 1945 to 1948, w as the first
to use precast segmental construction for prestressed concrete bridges. A bridge at Luzancy
over the Marne River about 30 miles east of Paris,
Figure 1.27, was followed by a group of five precast
bridges over that river. Shortly thereafter, Ulrich
Finsterwalder applied cast-in-place segmental prestressed construction in a balanced cantilever
fashion to a bridge crossing the Lahn River at Balduinstein, Germany. This system of cantilever
segmental construction rapidly gained wide acceptance in Germany, after construction of a
bridge crossing the Rhine at Worms in 1952, as
shown in Figure 1.28,s w ith three spans of 330,
371, and 340 ft (100, 113, and 104 m). More than
300 such structures, w ith spans in excess of 250 ft
(76 m), were constructed between 1950 and 1965
SECflON
FIGURE 1.23. Completed
Ardrossan
o\crpass
(courtesy of the Portland Cement Association).
in Europe.s Since then the concept has spread
throughout the world.’
Precast segmental construction also was evolving
during this period. In 1952 a single-span county
bridge near Sheldon, New York, was designed by
the Freyssinet Company. Although this bridge was
constructed of longitudinal rather than the European transverse segments, it represents the first
practical application of match casting. The bridge
girders were divided into three longitudinal segments that were cast end-to-end. The center segment was cast first and then the end segments were
cast directly against it. Keys were cast at the joints
so that the three precast elements could be joined
at the site in the same position they hid in the precasting yard. Upon shipment to the job site the
three elements of a girder were post-tensioned together with cold joints. l”,ll
The first major application of match-cast, precast segmental construction was not consummated
A-A
FIGURE 1.24. Field spike for continuity (courtesy of the Prestressed Concrete Institute, from ref. 6).
12
Prestressed
Concrete
Bridges
and
1.
FIGURE 1.25. Sebastian Inlet Bridge (courtesy of the
Po rtland Cement A ssociation).
until 1962. This structure, designed by Jean Muller
and built by Entreprises Campenon Bernard, was
the Choisy-le-Roi Bridge over the Seine River
south of Paris, Figure 1.29. This concept has been
refined and has spread from France to all parts of
the world.
The technology of cast-in-place or precast segmental bridges has advanced rapidly in the last
decade. During its initial phase the balanced
cantilever method of construction was used. Currently, other techniques such as span-by-span, incremental launching, or progressive placement
also are available. Any of these construction
methods may call on either cast-in-place or precast
segments or a combination of both. Consequently,
a variety of design concepts and construction
methods are now available to economically produce segmental bridges for almost any site condition.
Segmental bridges may be classified broadly by
four criteria:
Segmental
Construction
The ultimate use of the bridge-that is, highw ay o r railway structure or combination
thereof. Although many problems are common to these two categories, the considerable
increase of live loading in a railway bridge
poses special problems that call for specific solutions.
2. The ty p e o f structure in term s o f statical
scheme and shape of the main bending members. Many segmental bridges are box girder
bridges, but other types such as arches or
cable-stayed bridges show a wide variety in
shape of the supporting members.
3.
The use of cast-in-place or precast segments or
a combination thereof.
4. The method of construction.
The sections that follow will deal briefly with the
last three classifications.
1.7
Various Types of Structures
From the point of view of their statical scheme,
there are essentially five categories of structures:
(1) girders, (2) trusses, (3) rigid frames, (4) arch
frames, and (5) cable-stayed bridges.
1.7.1 GIRDER BRIDGES
Box girders in the majority of cases are the most
efficient and economical design for a bridge. When
constructed in balanced cantilever, box girder
decks were initially made integral with the piers
w hile a special expansion joint w as provided at the
center of each span (or every other span) to allow
Conventional
\zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
Section A-A
FIGURE 1.26. Sebastian Inlet Bridge (courtesy of the Prestressed Concrete Institute, from ref. 6).
13
Various Types of Structures
FIGURE 1.29. Choisy-le-Roi
FIGURE 1.27.
I,uzanc~
Bridge over the Marne River.
FIGURE 1.28.
b’ornx Bridge (courtesy of Dyckerhoff
& LVidmann).
CF
6lb’
E N D PIEI
1 7 6 ’-0 ’
1
for v o lu m e changes and to control differential
deflections between individual cantilever arms. It is
now recognized that continuity of the deck is desirable, and most structures are now continuous over
several spans, bearings being provided between
deck and piers for expansion.
Today, the longest box girder bridge structure
that has been built in place in cantilever is the
Koror Babelthuap crossing in the Pacific Trust territories with a center span of 790 ft (241 m), Figure
1.30.r2 A box girder bridge has been proposed for
zyxwvutsrqpo
M AI N
f-
Bridge.
PIER
I_
L
\
FIGURE 1.30.
/J
12%
._
176’-0“
I
I
12/-O”
Koror-Babelthuap Bridge, elevation and cross section (ref. 12).
14
Prestressed
Longitudmal
Concrete Bridges and Segmental Construction
section
r
1
G-r-r
Typical sections at span
and over main piers
center
IF-4
FIGURE
FIGURE 1.31. The Great Belt Project.
the Great Belt Project in Denmark with a 1070 ft
(326 m) clear main span, Figure 1.31. The box
girder design has been applied with equal success to the construction of difficult and spectacular
structures such as the Saint Cloud Bridge over
the Seine River near Paris, Figure 1.32, or to the
construction of elevated structures in very congested urban areas such as the B-3 Viaducts near
Paris, Figure 1.33.
1.33.
R-3 Viaciuc t\. FI ‘111~ e.
The cantilever method has potential applications
between the optimum span lengths of typical box
girders for the low ranges and of stayed bridges for
the high ranges.
1.7.3 FRAMES WITH SLANT LEGS
When the configuration of the site allows, the use
of inclined legs reduces the effective span length.
1.7.2 TRUSSES
When span length increases, the typical box girder
becomes heavy and difficult to build. For the purpose of reducing dead weight while simplifying
casting of very deep web sections, a truss with open
webs is a very satisfactory type that can be conveniently built in cantilever, Figure 1.34. The technological limitations lie in the complication of connections b e t w e e n p r e s t r e s s e d d i a g o n a l s a n d
chords. An outstanding example is the Rip Bridge
in Brisbane, Australia, Figure 1.35.
FIGURE
1.32. Saint Cloud Bridge, France.
FIGURE 1.34. Long-span concrete trusses.
FIGURE
1.35.
Rip Bridge, BI ishne, Xu\tl
nli,l
Vario us Ty pes
of St ruct ures
15
FIGURE 1.36. Long-span frame.
Provisional back stays or a temporary pier are
needed to permit construction in cantilever, Figure
1.36. This requirement may sometimes present
difficulty. An interesting example of such a scheme
is the Bonhomme Bridge over the Blavet River in
France, Figure 1.37.
The scheme is a transition between the box
girder with vertical piers and the true arch, where
the load is carried by the arch ribs along the pressure line with minimum bending while the deck is
supported by spandrel columns.
FIGURE 1.37. Bonhomme
Bridge.
1.7.4 CONCRETE ARCH BRIDGES
Concrete arches are an economical way to transfer
loads to the ground where foundation conditions
are adequate to resist horizontal loads. Eugene
Freyssinet prepared a design for a 1000 meter
(3280 ft) clear span 40 years ago. Because of construction difficulties, however, the maximum span
built to date (1979) has been no more than 1000 ft
(300 m). Construction on falsework is made
difficult and risky by the effect of strong winds
during construction.
The first outstanding concrete arch was built at
Plougastel by Freyssinet in 1928 with three 600 ft
(183 m) spans, Figure 1.38. Real progress was
achieved only when free cantilever and provisional
stay methods were applied to arch construction,
Figure 1.39. The world record is presently the Kirk
Bridge in Yugoslavia, built in cantilever and com-
FIGURE 1.38. Plougastel Bridge, France.
Prestressed Concrete Bridges and Segmental Construction
16
I
FIGURE 1.39. Concrete arches.
,,,)
.
.
.
\
,.
.
.
,,~
\
pleted in 1979 w ith a clear span of 1280 ft (390 m),
Figure 1.40.
,_
i
1.7.5 CONCRETE CABLE-STAYED BRIDGES’”
FIGURE 1.40. Kirk Bridges, Yugoslavia.
When a span is beyond the reach of a conventional
girder bridge, a logical step is to suspend the deck
by a system of pylons and stays. Applied to steel
structures for the last twenty years, this approach
gained immediate acceptance in the field of concrete bridges when construction became possible
FIGURE 1.41. Long-span concrete cable-stayed bridges.
m
dr
Cast-in-Place and Precast Segmental
Construction
17
the structure’s deformability, particularly during
construction. Deflections of a typical cast-in-place
cantilever are often two or three times those of the
same cantilever made of precast segments.
The local effects of concentrated forces behind
the anchors of prestress tendons in a young concrete (two or four days old) are always a potential
source of concern and difficulties.
I.82 CHARACTERISTICS OF PRECAST SEGM ENTS
FIGURE 1.42. Krotonne Bridge, France.
and economical in balanced cantilever with a large
number of stays uniformly distributed along the
deck, Figure 1.41, The longest span of this type is
the Brotonne Bridge in France with a 1050 ft (320
m) clear main span over the Seine River, Figure
1.42. Single pylons and one line of stays are located
along the centerline of the bridge.
1.8
Cast-in-Place and Precast Segmental
Construction
1.8.1 CHARACTERISTICS OF CAST- IN- PLACE
SEGMENTS
In cast-in-place construction, segments are cast one
after another in their final location in the structure.
Special equipment is used for this purpose, such as
travelers (for cantilever construction) or formwork
units moved along a supporting gantry (for spanby-span construction). Each segment is reinforced
with conventional untensioned steel and sometimes by transverse or vertical prestressing or both,
while the assembly of segments is achieved by longitudinal post-tensioning.
Because the segments are cast end-to-end, it is
not difficult to place longitudinal reinforcing steel
across the joints between segments if the design
calls for continuous reinforcement. Joints may be
treated as required for safe transfer of all bending
and shear stresses and for water tightness in aggressive climates. Connection between individual
lengths of longitudinal post-tensioning ducts may
be made easily at each joint and for each tendon.
The method’s essential limitation is that the
strength of the concrete is always on the critical
path of construction and it also influences greatly
In precast segmental construction, segments are
manufactured in a plant or near the job site, then
transported to their final position for assembly.
Initially, joints between segments were of conventional type: either concrete poured wet joints or
dry mortar packed joints. Modern segmental construction calls for the match-casting technique, as
used for the Choisy-le-Roi Bridge and further developed and refined, whereby the segments are
precast against each other, preferably in the same
relative order they will have in the final structure.
No adjustment is therefore necessary between
segments before assembly. The joints are either
left dry (in areas where climate permits) or made of
a very thin film of epoxy resin or mineral complex,
which does not alter the match-casting properties.
There is no need for any waiting period for joint
cure, and final assembly of segments by prestressing may proceed as fast as practicable.
Because the joints are of negligible thickness,
there is usually no mechanical connection between
the individual lengths of tendon ducts at the joint.
Usually no attempt is made to obtain continuity
of the longitudinal conventional steel through the
joints, although several methods are available and
have been applied successfully (as in the Pasco
Kennewick cable-stayed bridge, for example).
Segments may be precast long enough in advance
o f their assem b ly in the structure to reach
sufficient strength and maturity and to minimize
both the deflections during construction and the
effects of concrete shrinkage and creep in the final
structure.
If erection of precast segments is to proceed
smoothly, a high degree of geometry control is required during match casting to ensure accuracy.
1.8.3
CHOICE BETW EEN CAST-IN-PLACE AND
PRECAST CONSTRUCTION
Both cast-in-place methods and precast methods
have been successfully used and produce substan-
18
Prestressed Concrete Bridges and Segmental Construction
tially the same final structure. The choice depends
on local conditions, including size of the project,
time allowed for construction, restrictions on access and environment, and the equipment available
to the successful contractor. Some items of interest
are listed below:
1. Speed
of Construction Basically, cast-in-place
cantilever construction proceeds at the rate of one
pair of segments 10 to 20 ft (3 to 6 m) long ever)
four to seven days. On the average, one pair of
travelers permits the completion of 150 ft (46 m) of
b rid g e d ec k p er m o nth, exc lu d ing the transf er
from pier to pier and fabrication of the pier table.
On the other hand, precast segmental construction
allows a considerably faster erection schedule.
a. For the Oleron Viaduct, the average speed of
completion of the deck was 750 ft (228 m) per month
for more than a year.
b . Fo r b o th the B- 3 V iad u c ts in Paris and the
Long Key Bridge in Florida, a typical 100 to 150 ft
(30 to 45 m) span was erected in two working days,
representing a construction of 1300 ft (400 m) offinished bridge per month,
c. Saint Cloud Bridge near Paris, despite the exc ep tio nal d iffic ulty o f its g eo m etry and d esig n
scheme, was constructed in exactly one year, its
total area amounting to 250,000 sq ft (23,600 sq
m).
It is evident, then, that cast-in-place cantilever construction is basically a slow process, while precast
segmental with matching joints is among the fastest.
2. Investment in Special Equipment Here the
situation is usually reversed. Cast-in-place requires
usually a lower investment, which makes it competitive on short structures with long spans [for
exam p le, a ty p ic al three- sp an stru c tu re w ith a
center span in excess of approximately 350 ft (100
Ml.
In long, repetitive structures precast segmental
may be more economical than cast-in-place. For
the Chillon Viaducts with twin structures 7000 ft
(2 134 m) long in a difficult environment, a detailed
c o m p arativ e estim ate sho w ed the c ast-in-p lac e
method to be 10% more expensive than the precast.
3. Size and Weight of Segments Precast segmental is limited by the capacity of transportation
and placing equipment. Segments exceeding 250
tons are seldom economical. Cast-in-place construction does not have the same limitation, al-
though the weight and cost of the travelers are directly proportional to the weight of the heaviest
segment.
4. Environment Restrictions Both precast and
cast-in-place segmental permit all work to be performed from the top. Precast, however, adjusts
more easily to restrictions such as allowing work to
proceed over traffic or allowing access of workmen
and materials to the various piers.
1.9
Various
Methods
of
Construction
Probably the most significant classification of segmental bridges is by method of construction .41though construction methods may be as varied as
the ingenuity of the designers and contractors,
they fall into four basic categories: (1) balanced
cantilever, (2) span-by-span construction, (3) prog ressiv e p lac em ent c o nstru c tio n, and (4) incremental launching or push-out construction.
1.9.1 CAST-I.\‘-PL4CE
BAL,-I,VCED C.4.iTILEC’ER
The balanced or free cantilever construction concept
w as o rig inally d ev elo p ed to elim inate
falsework. Temporary shoring not only is expensive but can be a hazard in the case of sudden
floods, as confirmed by many failures. Over navigable waterways or traveled highways or railways,
falsework is either not allowed or severely restricted.’ Cantilever construction, whether cast in
place or precast, eliminates such difficulties: construction may proceed from the permanent piers,
and the structure is self-supporting at all stages.
The basic principle of the method was outlined in
Section 1.1 (Figure 1.3).
In cast-in-p lace co nstructio n the formw ork is
supported from a movable form carrier, Figure
1.1. Details of the form travelers are shown in Figure 1.43. The form traveler moves forward on rails
attached to the deck of the completed structure
and is anchored to the deck at the rear. With the
form traveler in place, a new segment is formed,
cast, and stressed to the previously constructed
segment. In some instances a covering may be provided on the form carrier so that work may proceed during inclement weather, Figure 1.44.
The o p eratio n seq uenc e in c ast-in-p lac e b alanced cantilever construction is as follows:
1.
Setting up and adjusting carrier.
2.
Setting up and aligning forms.
Various Methods of Construction
CENTERJACK
FORM TRAVELLER
8,i-J?i,!
-Lu
,.
! I
i-HUN I AL
WORKING PLATFORM
ADDITIONAL
REAR GANG-BOARD
FIGURE
3.
4.
~ONTAL LOWER
WORKING PLATFORM
1.43. Form traveler (courtesy of Dyckerhoff & Widmann).
Placing reinforcement and tendon ducts.
Concreting.
5.
Inserting prestress tendons in the segment and
stressing.
6. Removing the formwork.
7.
\BOTTOM
FRAME WORK
Moving the form carrier to the next position
and starting a new cycle.
Initially, the normal construction time for a
segment was one week per formwork unit. Advances in precast segmental construction have been
applied recently to the cast-in-place method in
order to reduce the cycle of operations and increase the efficiency of the travelers. With today’s
technology it does not seem possible to reduce the
FIGURE 1.44. Bendorf Bridge form traveler (courtesy of Dyckerhoff & Widmann).
construction time for a full cycle below two working days, and this only for a very simple structure
with constant cross section and a moderate amount
of reinforcing and prestress. For a structure with
variable depth and longer spans, say above 250 ft
(75 m), the typical cycle is more realistically three to
four working days.
Where a long viaduct type structure is to be constructed of cast-in-place segments, an auxiliary
steel girder may be used to support the formwork,
Figure 1.45, as on the Siegtal Bridge. This equip-
FIGURE
1.45. Siegtal Bridge, use of an auxiliary truss
in cast-in-place construction.
20
Prestressed
Concrete Bridges and Segmental Construction
ment may also be used to stabilize the free-standing
pier by the anchoring of the auxiliary steel girder
to the completed portion of the structure. Normally, in construction using the form traveler previously described, a portion of the end spans (near
the abutments) must be cast on falsework. If the
auxiliary steel girder is used, this operation may be
eliminated. As soon as a double typical cantilever is
completed, the auxiliary steel girder is advanced to
the next pier. Obviously, the economic justification
for use of an auxiliary steel girder is a function of
the number of spans and the span length.
I-9.2. PRECAST BALANCED CANTILEVER
For the first precast segmental bridges in Paris
(Choisy-le-Roi, Courbevoie, and so on, 1961 to
1965) a floating crane was used to transfer the precast segments from the casting yard to the barges
that transported them to the project site and was
used again to place the segments in the structure.
The concept of self-operating launching gantries
was developed shortly thereafter for the construction of the Oleron Viaduct (1964 to 1966). Further
refined and extended in its potential, this concept
has been used in many large structures.
The erection options available can be adapted to
almost all construction sites.
1. Crane Placing Truck or crawler cranes are
used on land where feasible; floating cranes may
be used for a bridge over navigable water, Figure
1.46. Where site conditions allow, a portal crane
may be used on the full length of the deck, preferably with a casting yard aligned with the deck near
one abutment to minimize the number of handling
operations, Figure 1.47.
2. Beam and W inch Method If access by land or
water is available under the bridge deck, or at least
around all permanent piers, segments may be
lifted into place by hoists secured atop the previously placed segments, Figure 1.48. At first this
method did not permit the installation of precast
pier segments upon the bridge piers, but it has
been improved to solve this problem, as will be explained later.
There are essentially
two families of launching gantries, the details of
which will be discussed in a later chapter. Here we
briefly outline their use.
3.
Launching
Gantries
In the first family developed for the Oleron Viaduct, Figures 1.49 and 1.50, the launching gantry
is slightly more than the typical span length, and
the gantry’s rear support reaction is applied near
the far end of the last completed cantilever. All
segments are brought onto the finished deck and
placed by the launching gantry in balanced cantilever; after completion of a cantilever, after
placing the precast segment over the new pier, the
launching gantry launches itself to the next span to
start a new cycle of operations.
In the second family, developed for the Deventer Bridge in Holland and for the Rio Niteroi
Bridge in Brazil, the launching gantry has a length
approximately twice the typical span, and the reaction of the legs is always applied above the permanent concrete piers, Figures 1.51 and 1.52.
Placing segments w ith a launching gantry is now
in most cases the most elegant and efficient
method, allowing the least disturbance to the environment.
1.9.3 SPAN-BY-SPAN CONSTRUCTION
The balanced cantilever construction method was
developed primarily for long spans, so that construction activity for the superstructure could be
accomplished at deck level without the use of extensive falsework. A similar need in the case of
long viaduct structures with relatively shorter
spans has been filled by the development of a
span-by-span methodology using a form traveler.
The follow ing discussion explains this methodol13.14.15.16
FIGURE 1.46. Segment erection by barge-mounted
crane, Capt. Cook Bridge, Australia (courtesy of G. Beloff, Main Roads Department, Brisbane, Australia).
%Y*zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLK
In long viaduct structures a segmental span-byspan construction may be particularly advantageous. The superstructure is executed in one direc-
Various Methods of Construction
COUPE
21
TRANSVERSALE
FIGURE 1.47. Mirabeau Bridge at Tours, France.
tion, span by span, by means of a form traveler,
Figure 1.53, with construction joints or hinges located at the point of contraflexure. The form carrier in effect provides a type of factory operation
transplanted to the job site. It has many of the ad:
. .
the field. The form traveler may be supported on
the piers, or from the edge of the previously completed construction, at the joint location, and at the
forward pier. In some instances, as in the approaches of Rheinbrticke, Dusseldorf-Flehe, the
movable formwork may be supported from the
ground, Figure 1.54. The form traveler consists of
a steel superstructure, which is moved from the
completed portion of the structure to the next span
to be cast. Fo r an abo v e-d eck carrier, larg e
formwork elements are suspended from steel rods
during concreting. After concreting and post-tensioning, the forms are released and rolled forward
by means of the structural steel outriggers on both
sides of the form traveler’s superstructure. For a
below-deck carrier, a similar procedure is followed.
Many long bridges of this type have been built in
Germany, France, and other countries. Typical
c o nstru c tio n tim e fo r a 100 ft (30 m ) sp an
superstructure is five to eight working days, depending upon the complexity of the structure.
Deck configuration for this type of construction is
usually a monolithic slab and girder (T beam or
double T), box girder, or a mushroom cross sec-
Prestressed
22
J-I
5 2 .0 0
m
170f t
Concrete Bridges and Segmental Construction
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
180 f t
54 .OOm _
10 6 . 0 0
Ill 3 5 0
ff
(6)
4
80.00
m
260 f t
c
FIGURE 1.51. Second family of launching gantries,
Rio Niteroi Bridge.
J
10 6 . 0 0
Cc)
FIGURE
1.49.
zyxwvutsrq
Il l 3 5 0
ft
I
First family of launching gantries (Ole-
ron Viaduct).
tion. This method has been used recently in the
United States on the Denny Creek project in the
state of Washington.
In its initial form, as described above, the spanby-span method is a cast-in-place technique. The
same principle has been applied in conjunction
with precast segmental construction for two very
large structures in the Florida Keys: Long Key
Bridge and Seven Mile Bridge, with spans of 118 ft
(36 m) and 135 ft (40 m), respectively. Segments
are assembled on a steel truss to make a complete
FIGURE 1.50. Placing precast segments on the Oleron Viaduct.
span. Prestressing tendons then assure the assembly of the various segments in one span while
achieving full continuity with the preceding span,
Figures 1.55 and 1.56. The floating crane used to
place the segments over the truss also moves the
truss from span to span. The contractor for the
Seven Mile Bridge modified the erection scheme
from that used for Long Key Bridge by suspending
a span of segments from an overhead falsework
truss. This is the first application of a method that
seems to have a great potential for trestle structures in terms of speed of construction and economy.
1.9.4
PROGRESSIVE
PLACEMENT
CONSTRUCTION
Progressive placement is similar to the span-byspan method in that construction starts at one end
of the structure and proceeds continuously to the
FIGURE 1.52. Rio Niteroi launching girder.
i of Construction
23
FIGURE 1.56.
Placing segments on assembly truss for
Long Key Bridge.
FIGURE 1.55. Span-by-span assembly of precast segments.
other end. It derives its origin, however, from the
cantilever concept. In progressive placement
the precast segments are placed from one end of
the structure to the other in successive cantilevers
on the same side of the various piers rather than
by balanced cantilevers on each side of a pier. At
present, this method appears practicable and
economical in spans ranging from 100 to 300 ft
(30 to 90 m).
Because of the length of cantilever (one span) in
relation to construction depth, a movable temporary stay arrangement must be used to limit the
cantilever stresses during construction to a reasonable level. The erection procedure is illustrated in
Figure 1.57. Segments are transported over the
completed portion of the deck to the tip of the
cantilever span under construction, where they are
positioned by a swivel crane that proceeds from
one segment to the next. Approximately one-third
of the span from the pier may be erected by the
free cantilever method, the segments being held in
position by exterior temporary ties and final prestressing tendons. For the remaining two-thirds of
the span, each segment is held in position by temporary external ties and by two stays passing
through a tower located over the preceding piers.
All stays are continuous through the tower and anchored in the previously completed deck structure.
The stays are anchored to the top flange of the box
girder segments so that the tension in the stays can
be adjusted by light jacks.
Used for the first time in France on several
structures, Figure 1.58, progressive placement is
being applied in the United States for the construction of the Linn Cove Viaduct in North
Carolina. In this bridge the precast pier construction proceeds also from the deck to solve a difhcult
problem of environmental restrictions.
24
Prestressed Co ncrete Bridges and Segmental Co nstructio n
FIGURE 1.57. Progressive placement erection procedure.
The progressive placement method may also be
applied to cast-in-place construction.
1.9.5. INCREM ENTAL LAUNCHING OR PUSH-OUT
C O N STR U C TI O N
This concept was first implemented on the Rio Caroni Bridge in Venezuela, built in 1962 and 1963
by its originators, Willi Baur and D r. Fritz
Leonhardt of the consulting firm of Leonhardt
and Andra (Stuttgart, Germany).”
Segments of the bridge superstructure are cast
in place in lengths of 30 to 100 ft ( 10 to 30 m) in
stationary forms located behind the abutment(s),
Figure 1.59. Each unit is cast directly against the
previous unit. After sufficient concrete strength is
reached, the new unit is post-tensioned to the previous one. The assembly of units is pushed forward
in a stepwise manner to permit casting of the succeeding segments, Figure 1.60. Normally a work
cycle of one week is required to cast and launch a
segment, regardless of its length. Operations are
Various Methods of Construction
FIGURE 1.60. Incremental
launching
(courtesy of Prof. Fritz Leonhardt).
25
sequence
superstructure under its own weight at all stages of
launching and in all sections. Four methods for this
purpose are used in conjunction with one another.
1.
A first-stage prestress is applied concentrically
to the entire cross section and in successive increments over the entire length of the
superstructure.
2. To reduce the large negative bending moments in the front (particularly just before the
superstructure reaches a new pier) a fabricated
structural steel launching nose is attached to
the lead segment, Figure 1.62.
3. Long spans may be subdivided by means of
temporary piers to keep bending moments to a
reasonable magnitude. This construction
technique has been applied to spans up to 200
ft (60 m) without the use of temporary
falsework bents. Spans up to 330 ft (100 m)
have been built using temporary supporting
bents. The girders must have a constant depth,
which is usually one-twelfth to one-sixteenth of
the longest span.
4. Another method has been used successfully in
France to control bending moments in the
26
Prestressed
FIGURE 1 . 6 1 . Incremental launching
(courtesy of Prof. Fritz Leonhardt).
Concrete
Bridges
and
Segmental
Construction
o n J GUI ve
deck in the forward part of the superstructure.
A system using a tower and provisional stays is
attached to the front part of the superstructure. The tension of the stays and the corresponding reaction of the tower on the deck are
controlled
automatically and continuously
during all launching operations to optimize the
stress distribution in the deck, Figure 1.63.
After launching is complete, and the opposite
abutment has been reached, additional prestressing is added to accommodate moments in the final
structure, while the original uniform prestress
must resist the varying moments that occur as the
superstructure is pushed over the piers to its final
position.
Today, the longest incrementally launched clear
span is over the River Danube near Worth, Germany, with a maximum span length of 550 ft (168
m). Two temporary piers were used in the river for
launching. The longest bridge of this type is the
Olifant’s River railway viaduct in South Africa with
23 spans of 147 ft (45 m) and a total length of 3400
‘FIGURE 1.62. Steel launching nose (courtesy of Prof.
Fritz Leonhardt).
ft (1035 m). The incremental launching technique
was used successfully for the first time in the
United States for the construction of the Wabash
River Bridge at Covington, Indiana.
1 .I 0
Applications of Segmental Construction
in the United States
The state of the art of designing and constructing
prestressed concrete segmental bridges has advanced greatly in recent years. A wide variety of
structural concepts and prestressing methods are
used, and at least a thousand segmental bridges
have been built throughout the world. We may
conclude that segmental prestressed concrete construction is a viable method for building highway
bridges. There are currently no known major
problems that should inhibit utilization of segmental prestressed concrete bridges in the United
States. They have been successfully consummated
in other countries and are increasingly being employed in the United States.
27
Applications of Segmental Construction in the United States
fbJ
FIGURE 1.64. ‘Three Sisters Bridge.
Cd)
FIGURE 1.63. Incremental launching with provisional tower and stays.
One of the earliest projects for which segmental
construction was considered was the proposed Interstate I-266 Potomac River Crossing in Washington, D.C., Figure 1.64, otherwise known as the
Three Sisters Bridge. This structure contemplated
a 750 ft (229 m) center span w ith side spans of 440
ft (134 m) on reverse five-degree curves, built with
cast-in-place segmental construction. Because of
environmental objections, this p ro ject nev er
reached fruition.
The JFK M em o rial Causeway (Intracoastal
Waterway), Corpus Christi, Texas, Figure 1.65,
represents the first precast, prestressed, segmental,
balanced cantilever construction completed in the
United States. It was opened to traffic in 1973. Designed by the Bridge Division of the Texas Highway Department, it has a center span of 200 ft (61
m) with end spans of 100 ft (30.5 m).
The first cast-in-place, segmental, balanced cantilever, prestressed concrete bridge constructed in
the United States is the Pine Valley Bridge in
California, on Interstate I-8 about 40 miles (64 km)
east of San Diego. Designed by the California Department of Transportation, .the dual structure,
Figure 1.66, has a total length of 1716 ft (53.6 m)
FIGURE 1.65. JFK hlcnwr ial Causewav.
Christi, Texas.
Co rp u s
FIGURE 1.66. Pine Valley Bridge (courtesy
CALTRANS).
of
Pt-estressed
Concrete Bridges and Segmental Construction
of
Houston
with spans of 270, 340, 450, 380, and 276 ft (82.3,
103.6, 137.2, 115.8, and 84.1 m).
As indicated previously, numerous segmental
bridge projects have been constructed or are contemplated in the United States. Many of them will
be discussed in detail in the following chapters.
Among the most significant are the Houston Ship
Channel Bridge with a clear span of 750 ft (228 m),
which will be the longest concrete span in the
Americas, Figure 1.67, and the Seven Mile Bridge,
which will be the longest segmental bridge in
North America, Figure 1.68.
1 .I 1
FIGURE
1.67.
Rendering
FIGURE 1.68. Rendering of’ Seven Mile Bridge.
Ship
Channel
Bridge.
Applicability and Advantages of Segmental
Construction
Segmental construction has extended the practical
range of span lengths for concrete bridges. Practical considerations of handling and shipping limit
the prestressed I-girder type of bridge construction to spans of about 120 to 150 ft (37 to 46 m).
Beyond this range, post-tensioned cast-in-place
box girders on falsework are the only viable concrete alternative. At many sites, however, falsework
is not practical or even feasible, as when crossing
deep ravines or large navigable waterways.
Falsework construction also has a serious impact
upon environment and ecology.
Prestressed concrete segmental construction has
been developed to solve these problems while extending the practical span of concrete bridges to
about 800 ft (250 m) or even 1000 ft (300 m). With
cable-stayed structures the span range can be extended to 1300 ft (400 m) and perhaps longer with
the materials available today.13 Table 1.1 summarizes the range of application of various forms
of construction by span lengths.
Although the design and construction of verylong-span concrete segmental structures pose an
important challenge, segmental techniques may
Applicability
TABLE 1.1
and
Advantages
Segmental
Construction
29
Range of Application of Bridge Type by Span Lengthsa
Brid ge Types
Sp a n
o- 150
loo- 300
loo- 300
250- 600
200- 1000
800-1500
of
ft
ft
ft
ft
ft
ft
I-type pretensioned girder
Cast-in-place post-tensioned box girder
Precast balanced cantilever segmental, constant depth
Precast balanced cantilever segmental, variable depth
Cast-in-place cantilever segmental
Cable-stay with balanced cantilever segmental
“1 fi = 0.3048 tn.
find even more important applications in moderate
span lengths and less spectacular structures. Especially in difficult urban areas or ecology-sensitive
sites, segmental structures have proven to be a valuable asset.
Today most sites for new bridges can be adapted
for segmental concrete construction. The principal
advantages of segmental construction may be
summarized as follows:
1. Segmental construction is an efficient and
economical method for a large range of span
lengths and types of structure. Structures with
sharp curves and variable superelevation may be
easily accommodated.
2. Concrete segmental construction often provides for the lowest investment cost. Savings of 10
to 20% over conventional methods have been
realized by competitive bidding on alternate designs or by realistic cost comparisons.
3. Segmental construction permits a reduction
of construction time. This is particularly true for
precast methods, where segments may be manufactured while substructure work proceeds and
be assembled rapidly thereafter. Further cost savings ensue from the lessening of the influence of
inflation on total construction costs.
4. Segmental construction protects the environment. Segmental viaduct-type bridges can
minimize the impact of highway construction
through environmentally sensitive areas. Whereas
conventional cut-and-fill type highway construction can scar the environment and impede wildlife
migration, an elevated viaduct-type structure requires only a relatively narrow path along the
alignment to provide access for pier construction.
Once the piers have been constructed, all construction activity proceeds from above. Thus, the
impact on the environment is minimized.
5. Interference with existing traffic during
construction is significantly reduced, and expensive detours can be eliminated. Figure 1.69 indi-
cates how precast segments may be handled while
traffic is maintained with a minimum disturbance.
6. Segmental construction contributes toward
aesthetically pleasing structures in many different
sites. A long approach viaduct (Brotonne, Figure
1.70), a curved bridge over a river (Saint Cloud,
Figure 1.7 l), or an impressive viaduct over a deep
valley (Pine Valley, Figure 1.66) are some examples
where nature accepts human endeavor in spite of
its imperfections.
7. Materials and labor are usually available locally for segmental construction. The overall labor
requirement is less than for conventional construction methods. For the precast option a major
part of the work force on site is replaced by plant
labo r.
8. As a consequence, quality control is easier to
perform and high-quality work may be expected.
9. Segmental bridges when properly designed
and when constructed by competent contractors
under proper supervision will prove to be practically free of maintenance for many years. Only
bearings and expansion joints (usually very few for
continuous decks) need to be controlled at regular
intervals.
FIGURE 1.69. Saint Cloud Bridge, segments placed
over traffic.
Prestressed
30
Concrete
Bridges
FIGURE 1.70. Brotonne Bridge approach.
10. During construction, the technique shows
an exceptionally high record of safety.
Precast segmental construction today is competitive in a wide range of applications with other
materials and construction methods, while it adds a
further refinement to the recognized advantages
of prestressed concrete.
FIGURE 1.71. Saint
bridge over a river.
C lo ud
Bridge, France, curved
References
1 . H. G. Tyrrell, History of Bridge Engineeting,
Henry G.
Tyrrell, Chicago, 1911.
2. Elizabeth B. Mock, The Architecture of Bridges, The
Museum of Modern Art, New York, 1949.
3. T. Y. Lin, Design of Prestressed Concrete Structures,
John Wiley & Sons, Inc., New York, 1958.
4. Anon., “Highway Design and Operational Practices
Related to Highway Safety,” Report of the Special
AASHO Traffic Safety Committee, February 1967.
5 . Anon., Prestressed Concrete for Long Span Bridges, Prestressed Concrete Institute, Chicago, 1968.
and
Segmental
Construction
6. Anon., “Long Spans with Standard Bridge Girders,”
PC1 Bridge Bulletin, March-April 1967, Prestressed
Concrete Institute, Chicago.
7. “Recommended Practice for Segmental Construction in Prestressed Concrete,” Report by PC1 Committee on Segmental Construction, Journal of the
Prestressed Concrete Instztute, Vol. 20, No. 2, MarchApril 1975.
8. Ulrich Finsterwalder, “Prestressed Concrete Bridge
Construction,” Journal oj the Amerzcan Concrete Instztute, Vol. 62, No. 9, September 1965.
9. F. Leonhardt, “Long Span Prestressed Concrete
Bridges in Europe,” Journal of the Pre.,tressed Concrete
Institute, Vol. 10, No. 1, February 1965.
10. Jean Muller, “Long-Span Precast Prestressed Concrete Bridges Built in Cantilever,” Fzrst International
Symposium, Concrete Bridge Design, AC1 P u b l i c a t i o n
SP-23, Paper 23-40, American Concrete Institute,
Detroit, 1969.
11. Jean Muller, “Ten Years of Experience in Precast
Segmental Construction,” Journal of the Prestressed
Concrete Instatute, Vol. 20, No. 1, January-February
1975.
12. Man-Chung Tang, “Koror-Babelthuap Bridge-A
World Record Span,” Preprint Paper 3441, ASCE
Convention, Chicago, October 16-20, 1978.
13. C. A. Ballinger, W. Podolny, Jr., and M. J. Abrahams, “A Report on the Design and Construction
of Segmental Prestressed Concrete Bridges in Western Europe- 1977,” International Road Federation, Washington, D.C., June 1978. (Also available
from Federal Highway Administration, Offices of
Research and Development, Washington, D.C., Report No. FHWA-RD-78-44.)
14. Ulrich Finsterwalder, “New Developments in Prestressing Methods and Concrete Bridge Construction,” Dywzdag-Berzchte,
4-1967, September 1967,
Dyckerhoff & Widmann KG, Munich, Germany.
15. Ulrich Finsterwalder, “Free-Cantilever Construction
of Prestressed Concrete Bridges and MushroomShaped Bridges,” First International Symposaum,
Concrete Bridge Deszgn, AC1 Publication SP-23, Paper SP
23-26, American Concrete Institute, Detroit, 1969.
16. C. A. Ballinger and W. Podolny, Jr., “Segmental
Construction in Western Europe-Impressions of
an IRF Study Team,” Proceedings, Conference conducted by Transportation Research Board, National
Academy of Sciences, Washington, D.C., TRR 665,
Vol. 2, September 1978.
17. Willi Baur, “Bridge Erection by Launching is Fast,
Safe, and Efficient,” Czvzl Engineerzng-AXE, Vol.
47, No. 3, March 1977.
18. Walter Podolny, Jr., and J. B. Scalzi, “Construction
and Design of Cable-Stayed Bridges,” John Wiley &
Sons, Inc., New York, 1976.
zy
2
Cast-in-Place Balanced Cantilever Girder Bridges
2.5
2.6
2.7
2.8
2.9
2.10
2.11
2.12
2.18
INTRODUCTION
BENDORF BRIDGE, GERMANY
SAINT ADELE BRIDGE, CANADA
BOUGUEN BRIDGE IN BREST AND LACROIK FALGARDE BRIDGE, FRANCE
SAINT JEAN BRIDGE OVER THE GARONNE RIVER
AT BORDEAUX, FRANCE
SIEGTAL AND KOCHERTAL BRIDGES, GERMANY
PINE VALLEY CREEK BRIDGE, U.S.A.
GENNEVILLIERS BRIDGE, FRANCE
GRAND’MFRE BRIDGE, CANADA
ARNHEM BRIDGE, HOLLAND
NAPA RIVER BRIDGE, U.S.A.
KOROR-BABELTHUAP, U.S. PACIFIC TRUST
TERRITORY
VEJLE FJORD BRIDGE, DENMARK
2.1
Introduction
Developed initially for steel structures, cantilever
construction was used for reinforced concrete
bridges as early as fifty years ago. In 1928, Freyssinet used the cantilever concept to construct the
springings of the arch rib in the Plougastel Bridge,
Figure 2.1. The reactions and overturning moments applied by the falsework to the lower part of
the arch ribs were balanced by steel ties connecting
the two short cantilevers. A provisional prestress
was thus applied by the ties to the arch ribs with the
aid of ja c ks and deviation saddles.
The first application of balanced cantilever construction in a form closely resembling its present
one is due to a Brazilian engineer, E. Baumgart,
who designed and built the Herval Bridge over the
Rio Peixe in Brazil in 1930. The 220 ft (68 m)
center span was constructed by the cantilever
method in reinforced concrete with steel rods extended at the various stages of construction by
threaded couplers. Several other structures fol-
2.14 HOUSTON SHIP CHANNEL BRIDGE, U.S.A.
2.15 OTHER NOTABLE STRUCXURFS
2.15.1
Medway Bridge, U.K.
2.15.2
Rio Tocantins Bridge, Brazil
‘2.153
Pueute Del Azufre, Spain
2.15.4
Schubeuamdie Bridge, Canada
2.15.5
Inci- Bridge, Guatemala
2.15.6
!3etubal Bridge, Argentina
2.15.7
Kipapa Stream Bridge, U.S.A.
2.15.8
Parrots Ferry Bridge, U.S.A.
2.15.9
Magnan Via’duct, France
2.15.10 Puteaux Bridge, Frame
2.15.11 Tricastiu Bridge, France
2.15.12 Eschachtal Bridge, Germauy
2.16
CONCLUSION
R EF ER EN CES
lowed in various countries, particularly in France.
Albert Caquot, a leading engineer of his time, built
several reinforced concrete bridges in cantilever.
Show n in Figures 2.2 through 2.4 is Bezons
Bridge over the River Seine near Paris, with a clear
center span of 310 ft (95 m), being constructed
in successive cantilever segments with auxiliary
trusses. This bridge design w as prepared in 1942.
The method was not widely used at that time,
because the excessive amount of reinforcing steel
Jack,
/ Ties
f
Overturning
moment due
to centering
FIGURE 2.1. Cantilever construction of arch springings for Plougastel Bridge, France.
31
FIGURE 2.2. Bezons Bridge over the Seine River, France, typical longitudinal and
transverse sections.
33
Introduction
,w--. ---.-._ -_-..--._
z
I
.!
I-
- _._.._- _______ _
:
:
zyxwv
h’
..*gr- _ _ ._- -.__. --I .- ._-__ ____L_
--/
::
:
.Izyxwvutsrqponml
il
FIGURE 2.3. Bezo ns Bridge, co nstructio n pro cedure.
required to balance the cantilever moments produced the tendency toward cracking inherent in
an overreinforced slab subject to permanent tensile stresses.
The introduction of prestressing in concrete
structures dramatically changed the situation.
Used successfully in 1950 and 195 1 by Finsterwalder with the German firm of Dyckerhoff & Widmann for the construction of the two bridges of
Balduinstein and Neckarrews, balanced cantilever
construction of prestressed concrete bridges experienced a continuous popularity in Germany
FIGURE 2.4. Bezons Bridges under construction.
34
Cast-in-Place Balanced Cantilever Girder Bridges
FIGURE 2.5. La Voulte Bridge, France.
and surrounding countries. Nicolas Esquillan designed and built a large bridge by the cantilever
method over the Rhine River in France, La Voulte
Bridge (J952), where an overhead truss was used
during construction, Figure 2.5.
Between 1950 and 1965 more than 300 such
bridges were constructed in Europe alone. Initially
all &uctures
were prestressed by high-strength
bars, and hinges were provided at the center of the
various spans. Later other prestressing methods
with parallel wire or strand tendons were also used.
More important, a significant improvement in
structural behavior and long-term performance
was made possible by the achievement of deck
continuity between the various cantilever arms.
The first cantilever bridges with continuous decks
were designed and built in France in 1962: the
Lacroix Falgarde Bridge and Bouguen Bridge,
Figures 2.6 and 2.22. Subsequently, the advantages
of continuity were recognized and accepted in
many countries.
From 1968 to 1970 cantilever construction was
considered for the Three Sisters Bridge in Washington, D.C., Figure 1.64. This project never
reached the construction stage. The first cast-inplace balanced cantilever segmental bridge built in
the United States is the Pine Valley Creek Bridge
in California (1972 to 1974), Figure 2.7. To date,
all segmental bridges constructed in the United
States have been either precast or cast-in-place
cantilever construction, with the following exceptions:
Wabash River Bridge, incrementally launched
(Chapter 7)
Denny Creek and Florida Keys Bridges, span-byspan construction (Chapter 6)
FIGURE 2.6. Bouguen Bridge in Brest, France. First continuous rigid-frame structure
built in balanced cantilever.
35
Bendorf Bridge, Germany
FIGURE 2.8. Bendorf Bridge (courtesy of Dvckerhoff
& Widmann).
FIGURE 2.7. Pine Valley Creek Bridge.
Linn Cove Viaduct, progressive placement construction (Chapter 6)
The balanced cantilever method of construction
has already been briefly described. In this chapter
we shall see how this method has been implemented on various structures before we go on
to consider specific design and technological aspects.
(west) are the river spans consisting of a symmetrical seven-span continuous girder with an overall
length of 1721 ft (524.7 mj. In part two (east) are
the nine-span continuous approach girders with
the spans ranging from 134.5 ft (41 m) to 308 ft (94
mj and having an overall length of 1657 ft (505 mj,
Figures 2.9 and 2.10.
The continuous, seven-span, main river structure consists of twin, independent, single-cell box
girders. Total width of the bridge cross section is
101 ft (30.86 mj. Each single-cell box has a top
flange width of 43.3 ft (13.2 mj, a bottom flange
width of 23.6 ft (7.2 mj, and webs with a constant
thickness of 1.2 ft (0.37 m). Girder depth is 34.28 ft
(10.45 m) at the pier and 14.44 ft (4.4 mj at
midspan
representing, with respect to the main
span, a depth-to-span ratio of l/ 20 and l/ 47, respectively. Girder depth of the end of this sevenspan unit reduced to 10.8 ft (3.3 mj. The main
navigation span has a hinge at midspan that is deHinge
2.2 Bendorf Bridge, Germany
Longitudinal
An early and outstanding example of the cast-inplace balanced cantilever bridge is the Bendorf
autobahn bridge over the Rhine River about 5
miles (8 km) north of Koblenz, West Germany.
Built in 1964, this structure, Figure 2.8, has a total
length of 3378 ft (1029.7 mj w ith a navigation span
of 682 ft (208 mj. The design competition allowed
the competing firms to choose the material,
configuration, and design of the structure. Navigation requirements on the Rhine River dictated a
328 ft (100 m) wide channel during construction
and a final channel w idth of 672 ft (205 mj. The
w inning design w as a dual structure of cast-inplace concrete segmental box girder construction,
consummated in two distinct portions. In part one
Cross sectton
river pier
at
section
Cross section
at pier G
FIGURE: 2.9. Bendorf Bridge, Part one (West), longitudinal section, plan, and cross secnons at the river
pier and pier G, from ref. 1 (courtesy of Beton- und
Stahlbeto nbauj.
36
Cast-in-Place Balanced Cantilever Girder Bridges
-~~ ss,o -L- SP.0 --L-- so0 -$A--zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
--5O$Om-
Longitudinal
-
-
~
section
Plan
FIGURE 2.10. Bendorf Bridge, Part Two (East), longitudinal section and
ref. 1 (courtesy of Beton- und Stahlbetonbau).
plan, from
signed to transmit shear and torsion forces only,
thus allow ing the su p erstru c tu re to b e c ast
monolithically with the main piers.1,2 After construction of the piers, the superstructure over the
navigable portion of the Rhine was completed
within one year. The repetition of the procedure in
240 segments executed one after the other offered
numerous occasions to mechanize and improve the
erection method.3,4
The deck slab has a longitudinally varying thickness from 11 in. (279.4 mm) at midspan to 16.5 in.
(419 mm) at the piers. The bottom flange varies in
thickness from 6 in. (152 mm) at midspan to 7.87 ft
(2.4 m ) at the p iers. To reduce dead-weight
bending-moment stresses in the bottom flange
concrete, compression reinforcement was used
extensiv ely in reg io ns aw ay fro m the p iers.
Thicknesses of the various elements of the cross
section are controlled partly by stress requirements
and partly by clearance requirements of the tendons and anchorages.
The stru c tu re is three-dimensionally prestressed: longitudinal prestressing uniformly distributed across the cross section; transverse
prestressing in the top flange; and inclined prestressing in the webs. A total of 560 Dywidag bars
la-in. (32 mm) in diameter resists the negative bending moment produced by a half-span, Figure 2.11.
The maximum concrete compressive stress in the
bottom flange at the pier is 1800 psi (12.4 MPa). As
a result of the three-dimensional prestress the tensile stresses in the concrete were negligible. The
longitudinal prestressing is incrementally decreased from the pier to the hinge at midspan and
to the adjacent piers; thus, shear stresses in the
webs on both sides of the main piers are almost
constant. Therefore, the web thickness remains
constant and the diagonal prestressing remains
very nearly constant.
Construction began on March 1, 1962. After
completion of the foundations and piers, balanced
cantilever operations began from the west river
pier in July 1963 and were completed at the end of
that year. Segments were 12 ft (3.65 m) in length in
the river span and 11.4 ft (3.48 m) in the remaining
spans. Segments were cast on a weekly cycle. As the
segments became shallower, the construction cvcle
was advanced to two segments per week. During
winter months, to protect operations from inclement weather, the form traveler was provided with
an enclosure, Figure 2.12.
FIGURE 2.11. Bendorf Bridge, cross section showing
tendons in the deck, ref. 2, (courtesy of the American
Concrete Institute).
FIGURE 2.12. Bendorf‘ Bridge, protective covering
for form traveler (courtesy of Ulrich Finsterwalder).
Saint Adele Bridge, Canada
FIGURE 2.13.
Ste. Adele Bridge, elevation, from ref. 5 (courtesy of
In the construction of the approach spans, the
five spans from the east abutment were built in a
routine manner with the assistance of falsework
bents. The four spans over water were constructed
by a progressive placement cantilever method (see
Chapter 6), which employed a temporary cablestay arrangement to reduce the cantilever stresses.
Eng$mritzg ~V~7o.~-R~cord).
This structure, built in 1964 (the same year as the
Bendorf Bridge), represents the first segmental
bridge, in the contemporary sense, constructed in
North America. It crosses the River of the Mules
near Ste. Adele, Quebec, and is part of the
Laurentian Autoroute. It is a single-cell box girder
continuous three-span dual structure with a center
span of 265 ft (80.8 m) and end spans of 132 ft 6 in.
(40.4 m), Figure 2.13. At one end is a prestressed
concrete 55 f-t (16.8 m) simple span. The bridge has
a 100 ft (30.5 m) vertical clearance over the river in
the canyon below.
The variable-depth girder is 16 ft 3 4 in. (4.96 m)
deep at the piers and 6 ft (1.83 m) deep at midspan
and its extremities, Figure 2.14. Each dual structure consists of a single-cell rectangular box 23 ft (7
m) wide with the top flange cantilevering on each
side 9 ft (2.75 m) for a total width of 41 ft (12.5 m),
Figure 2.15, providing three traffic lanes in each
direction. Thickness of bottom flange, webs, and
top flange are respectively 1 ft l# in. (0.35 m), 1 ft 6
in. (0.46 m), and 1 ft (0.3 m).5
A total of 70 prestressing tendons were required
in each girder. Each tendon of the SEEE system
consists of seven strands of seven 0.142 in. (3.6
mm) wires. The seven strands are splayed out
through a steel ring in the anchorage and held in a
circular pattern by steel wedges between each of
the strands. The number of tendons anchored off
at each segment end varies with the distance from
the pier, increasing from an initial six tendons to
eight tendons at the eighth segment, then decreasing to two tendons at the eleventh segment at
midspan. There are an additional 44 positivemoment tendons in the center span located in the
bottom flange.5
FIG U RE 2.14. Stc. A dele HI idge, v i e w 01 variabledepth box girder (courtesy of the Portland Cement ASsociation).
FIGURE 2.15. Ste. Adele Bridge, view of end of box
girder segment (courtesy of the Portland Cement Association) .
2.3
Saint Adele Bridge, Canada
38
Cast-in-Place Balanced Cantilever Girder Bridges
counterweighted with 70 tons (63.5 mt) of concrete
block, which was gradually diminished as construction proceeded and the depth of the segments
decreased. The first pair of segments (at the pier),
each with a length of 21 ft 23 in. (6.47 m), were cast
on a temporary scaffolding braced to the pier,
Figure 2.18, which remained fixed in position
throughout the erection process.5
Construction of four segments per week, one at
each end of a cantilever from two adjacent piers,
was attained by the following five-day construction
cycles:
FIGURE 2.16. Ste. Adele Bridge, dual structure
under construction by the balanced cantilever method,
from ref. 5 (courtesy of Engineering New s- Reco rd).
Forty-seven segments are required for each
structure, eleven cantilevered each side of each
pier, a closure segment at midspan of the center
span, and a segment cast in place on each abutment. Segments cast by the form traveler were 10
ft 78 in. (3.24 m) in length.5 Four traveling forms
were used on the project: one pair on each side of
the pier for each of the dual structures, Figures
2.16 and 2.17.
The forms were supported by a pair of 42 ft
(12.8 m) long, 36 in. (914.4 mm) deep structural
steel beams spaced 15 ft (5.57 m) on centers, that
cantilevered beyond the completed portion of the
structure. Initially the cantilevered beams were
Traveling forms moved, bottom flange
First day:
formed, reinforced, and cast. In the parallel span
there was a one-day lag such that crews could shift
back and forth between adjacent structures.
Second day :
Concrete placed for webs and top
flange, cure begun.
Third day :
Fourth day : T e n d o n s p l a c e d a n d p r e s t r e s s i n g
jacks positioned while concrete was curing.
Fifth day : Prestressing
accomplished. Forms
stripped; preparations made to repeat cycle.
The cycle began on Monday. Since there was a
lag of one day on the parallel structure, a six-day
work week was required. Upon completion of the
eleventh segment in each cantilever the contractor
installed temporary falsework to support the
abutment end and then cast the closure segment at
midspan. Counterweights were installed at the
abutment end to balance the weight of the closure
forms and segment weight. After installation and
stressing of the continuity tendons, abutment segments were cast and expansion joints installed.5
2.4
FIGURE 2.17. Ste. Adele Bridge, view of form travelers cantilevering from completed portion of the structure, from ref. 5 (courtesy of Engineering News- Record).
Reinforcement placed for webs and
top flange.
Bouguen Bridge in Brest and Lacroix Falgarde
Bridge, France
The Bouguen Bridge in Brittany, West Province in
France, is the first rigid-frame continuous structure built in balanced cantilever (1962 to 1963).
The finished bridge is shown in Figure 2.6, while
dimensions are given in Figure 2.19. It carries a
three-lane highway over a valley 145 ft (44 m)
deep-Le Vallon du Moulin H Poudre-and provides a link between the heart of Brest city and Le
Bouguen, a new urban development.
The total length of bridge is 684 ft (208 m). The
main structure is a three-span rigid frame with
Bouguen Bridge in Brest
and Lacroix
Falgarde
39
Bridge, France
FIGURE 2.18. Ste. Adele Bridge, schematic of construction sequence, from
ref. 5 (courtesy of Engineering New s- Reco rd).
box girder is 10 ft (3 m); web thickness also is constant throughout the deck and is equal to 9$ in.
(0.24 m).
Piers consist of two square box columns 10 ft by
10 ft (3 x 3 m) with wall thickness of 9$ in. (0.24 m)
located under each deck girder. Two walls 84 in.
(0:22 m) thick with a slight recess used for architectural purposes connect the two columns.
Both piers are of conventional reinforced concrete
construction, slip-formed at a speed reaching 14 ft
(4.25 m) per day in one continuous operation.
piers elastically built-in on rock foundations with
span lengths of 147,268, and 147 ft (45,82, and 45
m). At one end the deck rests on an existing
masonry wall properly strengthened; at the other
end a shorter rigid frame with a clear deck span of
87 ft (26.5 m) provides the approach to the main
bridge.
The deck consists of two box girders with vertical
webs of variable height, varying from 15 ft 1 in.
(4.6 m) at the support to 6.5 ft (2 m) at midspan
and the far ends of the side spans. Width of each
Midspan
section
Pier
section
Plan section at pier
(b)
FIGURE 2.19. Bouguen Bridge, France, general dimensions. (a) Longitudinal section.
(6) Cross sections.
40
Cast-in-Place Balanced Cantilever Girder Bridges
FIGURE 2.20. Bouguen Bridge, construction of east
cantilever.
The superstructure box girders are connected to
the pier shaft by transverse diaphragms made integral with both elements to insure a rigid connection between deck and main piers. Construction of
the deck proceeded in balanced cantilever with 10
ft (3 m) long segments cast in place in form travelers with a one-week cycle, Figures 2.20 and 2.21.
High-early-strength concrete was used and no
steam curing was required. Concrete was allowed
to harden for 60 hours before application of prestress. The following cube strengths were obtained
throughout the project:
60 hours (time of prestress)
7 days
28 days
90 days
3700 psi (25.5 MPa)
5500 psi (37.9.MPa)
7000 psi (48.3 MPa)
8200 psi (56.5 MPa)
Only one pair of form travelers was used for the
entire project, but each traveler could accommodate the construction of both girders at the same
time.
.\’
FIGURE
2.21.
Bouguen Bridge, view of’ the traveler.
During construction of the deck, much attention
was given to the control of vertical deflections.
Adequate camber was given to the travelers to fully
compensate for short- and long-term concrete
deflections. The cumulative deflection at midspan
of the first cantilever arm was 14 in. (40 mm) at
time of completion. Concrete creep caused this
deflection to reach 3 in. (75 mm) at the time the
second cantilever arm reached the midspan section. Proper adjustment of the travelers allowed
both cantilever arms to meet within t in. (3 mm) at
the time continuity was achieved. Flat jacks were
provided over the outer supports to allow for any
further desired adjustment.
The structure is prestressed longitudinallv by
tendons of eight 12 mm strands:
76 tendons over the top of the pier segment,
32 tendons at the bottom of the crown section,
20 tendons in the side spans,
and transversely by tendons of seven 12 mm
strands.
The Lacroix Falgarde Bridge over Ariege
River in France, built in 1961 and 1962, is similar
to the Bouguen Bridge and represents the first
continuous deck built in balanced cantilever (see
the photograph of the finished bridge, Figure
2.22). It consists of three continuous spans 100,
200, and 100 ft (30.5, 61, and 30.5 m). The single
box girder has a depth varying between 4 ft 5 in.
and 10 ft 6 in. (1.35 to 3.2 m). Dimensions are
given in Figure 2.23. The superstructure rests on
both piers and abutments through laminated
bearing pads.
The deck was cantilevered and the construction
started simultaneously from the two piers with
four travelers working symmetrically. During con-
FIGURE 2.22. Lacroix-Falgardc Bridge, view of’ the
structure during construction.
Saint Jean Bridge Over the Gardonne River at Bordeaux, France
FIGURE 2.23.
Lacroix-Falgarde Bridge, elevation and cross section.
struction, the deck was temporarily fixed to the
piers by vertical prestress. The structure is prestressed longitudinally by tendons of twelve 8 mm
strands and transversely by tendons of twelve 7
mm strands.
2.5
Saint Jean Bridge over the Garonne River at
Bordeaux, France
Completed in April 1965, the Saint Jean Bridge in
Bordeaux is a remarkable application of the new
concepts developed at that time in cast-in-place
cantilever construction. The main structure has an
overall length of 1560 ft (475 m) and is continuous
with expansion joints only over the abutments. The
deck is f’ree to expand on neoprene bearings located on all river piers, Figure 2.24. A very
efficient method of pier and foundation construction was also developed, which will be described in
more detail in Chapter 5.
The bridge was built in the heart of the city of
Bordeaux over the Garonne River between a 175year-old multiple-arch stone structure and a 120year-old railway bridge designed by Eiffel, the engineer who designed the Eiffel Tower.
The main structure includes six continuous
spans. The central spans are 253 ft (77 m) long and
allow a navigation clearance of 38 ft (11.60 m)
above the lowest water level, while the end spans
are only 222 ft (67.80 m) long. Short spans at both
ends, 50 ft (15.40 m) long, provide end restraint of
the side spans over the abutments. The overall
width of the bridge is 88 ft (26.80 m), consisting of
six traffic lanes, two walkways, and two cycle lanes.
Superstructure dimensions are shown in Figure
2.25.
41
The deck consists of three box girders. The constant depth of 10.8 ft (3.30 m) has been increased
to 13 ft (3.90 m) over a length of 50 ft (15 m) on
each side of the piers to improve the bending
capacity of the pier section and reduce the amount
of cantilever prestress. No diaphragms were used
except over the supports. The results of a detailed
analysis performed to determine the transverse
behavior of the deck confirmed this choice (see
detailed description in Chapter 4).
Longitudinal prestressing consists of tendons
with twelve 8 mm and twelve t in. strands. Transverse prestressing consists of tendons with twelve 8
mm strands at 2.5 ft (0.75 m) intervals. Vertical
prestressing is also provided in the webs near the
supports.
As indicated in Figure 2.26, three separate pier
columns support the three deck girders. They are
capped with large prestressed transverse diaphragms. The piers are founded in a gravel bed located at a depth of 45 ft (14 m) below the river level
by means of a reinforced concrete circular caisson
FIGURE 2.24. Saint Jean at Bordeaux, view of the
completed structure.
COUPE LONGITUDINALE
CULEE R D
CVLEE NE
- _
5
FIGURE 2.25. Saint Jean ar Bordeaux. (a) Longitudinal and (6) cross sections.
FIGURE 2.26.
42
Saint Jean Bridge at Bordeaux, typical section at river piers.
Siegtal and Kochertal Bridges, Germany
FIGURE 2.27. Saint Jean Bridge at Bordeaux, work
progress on piers and deck.
18.5 ft (5.60 m) in diameter and 10 ft (3 m) high,
floated and sunk to the river bed and then opendredged to the gravel bed. Precast circular matchcast segments prestressed vertically make up the
permanent walls of caissons, while additional segments are used temporarily as cofferdams and
support for the deck during cantilever construction. A lower tremie seal allows dewatering and
placing of plain concrete fill inside the caisson. The
reinforced concrete footing and pier shaft are
finally cast in one day.
The superstructure box girders were cast in
place in 10 ft (3.05 m) long segments using twelve
form travelers, allowing simultaneous work on the
three parallel cantilevers at two different piers.
The 20 ft (6.1 m) long pier segment was cast on the
temporary supports provided by the pier caissons,
allowing the form travelers to be installed and cantilever construction to proceed. Six working days
were necessary for a complete cycle of operations
on each traveler. Work progress is shown in Figures 2.27 and 2.28. Total construction time for the
entire 130,000 sq ft (12,000 m*) was approximately
FIGURE 2.28. Saint Jean Bridge at Bordeaux, cantilever construction on typical pier.
43
one year, as shown on the actual program of work
summarized in graphic form in Figure 2.29. To
meet the very strict construction deadline of the
contract, it was necessary to bring to the project site
another set of three travelers to cast the last cantilever on the left bank and achieve continuity with
the southern river pier cantilever. Altogether,
meeting the two-year construction schedule was
recognized as an engineering achievement.
Exactly one hundred years earlier, Gustave Eiffel had built the neighboring railway bridge in
exactly two years-food for thought and a somewhat humbling reflection for the present generation.
2.6
Siegtal and Kochertal Bridges, Germany
The Siegtal Bridge near the town of Sieger, north
of Frankfort, Germany, represents the first industrial application of cast-in-place cantilever construction with an auxiliary overhead truss. This
method was initially developed by Hans Wittfoht
and the firm of Polensky-und-Zollner and subsequently used for several large structures in Germany and other countries. One of the most recent
and remarkable examples of this technique is the
Kochertal Bridge between Ntiremberg and Heilbron, Germany. Both structures will be briefly described in this section, while a similar application in
Denmark is covered in another section of this
chapter.
Siegtal Bridge is a twelve-span structure 3450 ft
(1050 m) long resting on piers up to 330 ft (100 m)
high, with maximum span lengths of 344 ft (105
m), Figure 2.30. Two separate box girders carry
the three traffic lanes in each direction for a total
width of 100 ft (30.5 m), Figure 2.31. Structural
height of the constant-depth box girder is 19 ft (5.8
m), corresponding to a span-to-depth ratio of 18.
The deck is continuous throughout its entire
length, with fixed bearings provided at the three
highest center piers and roller bearings of highgrade steel for all other piers and end abutments.
Piers have slip-formed reinforced concrete hollow
box shafts with a constant transverse width of 68
ft (20.7 m) and a variable width in elevation with
a slope of 40 to 1 on both faces.
The superstructure w as cast in place in balanced
cantilever from all piers in 33 ft (10 m) long segments with an auxiliary overhead truss supporting
the two symmetrical travelers, and a cycle of one
week was obtained without difficulty for the construction of two symmetrical 33 ft (10 m) long seg-
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FIGURE 2.29.
Saint Jean Bridge at Bordeaux, actual program of work.
Elevation t
Cross section 1
‘Cross section 2
Horizontal section
FIGURE 2.30. Siegtal Bridge, general dimensions.
45
Siegtal and Kochertal Bridges, Germany
II 59
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FIGURE 2.31. Siegtal Bridge, typical cro ss sectio n.
merits. The auxiliary truss w as also first used to cast
the pier segment above each pier, Figure 2.32, before cantilever construction could proceed, Figure
2.33. Because the pier shafts are flexible and have
limited bending capacity, it was inadvisable to subject them to unsymmetrical loading conditions
during deck construction. Thus, the overhead
truss also served the purpose of stabilizing the cantilever arms before continuity was achieved with
the previous cantilever.
The auxiliary steel truss is made of high-strength
steel (50 ksi yield strength). Prestressing is applied
to the upper chord, which is subjected to high tensile stresses in order to reduce the weight of the
equipment. The overall length of the truss is 440 ft
(135 m) long to accommodate the maximum span
length of 344 ft (105 m). The total weight of the
truss and of the two suspended travelers, allowing
casting of two 33 ft (10 m) long segments, was 660 t
(600 mt). Deck concrete was pumped to the various
segments through pipes carried from the finished
deck bv the auxiliary truss, Figures 2.34 and 2.35.
Work commenced on the superstructure in
March 1966. The first box girder was completed in
FIGURE 2.32. Sicgtal Bridge, pier segment ca5ting.
April 1968. The truss and travelers were immediately transferred to the second box girder,
which was completed in September 1969. Thus,
the average rate of casting was as follows:
First bridge: 3450 ft (1050 m) in 25 months, or 140
ft (42 m) per month
Secohd bridge: 3450 ft (1050 m) in 17 months, or
200 ft (62 m) per month
Both bridges: 6900 ft (2100 m) in 42 months, or
160 ft (50 m) per month
An outstanding contemporary example of the
same technique is the Kochertal Bridge in Germany, shown in final progress in Figure 2.36. General dimensions of the project are given in Figure
2.37. Total length is 3700 ft (1128 m) w ith typical
spans of 453 ft (138 m) supported on piers up to
600 ft (183 m) in height. The single box girder
superstructure with precast outriggers carries six
traffic lanes for a total width of 101 ft (30.76 m).
Box piers were cast in climbing forms with 14.2 ft
(4.33 m) high lifts. The top section is constant for
all piers with outside dimensions of 16.4 by 28.2 ft
FIGURE 2.33. Siegtal Bridge, canClever construction.
II
46
Cast-in-Place Balanced Cantilever Girder Bridges
FIGURE 2.34.
:
4
105m
105m
Siegtal Bridge, elevation of overhead truss and travelers.
(5 by 8.6 m). The four faces are sloped to inc re a se
the dimensions at foundation level to a maximum
of 31.2 by 49.2 ft (9.5 by 15 m) for the highest
piers. Wall thickness varies progressively from top
to bottom, to follow the load stresses, from 20 in.
(0.5 m) to 36 in. (0.9 m).
The constant-depth superstructure is cast in two
stages, Figure 2.38: (1) the single center box with a
width of 43 ft (13.1 m) and a depth of 23 ft (7 m),
and (2) the two outside cantilevers resting on a series of precast struts. To meet the very tight construction schedule of 22 months it was necessary to
use two sets of casting equipment, working simultaneously from both abutments toward the center.
Each apparatus w as made of an overhead truss
equipped with a launching nose to move from pier
to pier and two suspended travelers working in
balanced cantilevers, casting segments on a onewe e k cycle, Figure 2.39.
FIGURE 2.35. Siegtal Bridge, typical section of truss
and travelers.
F I G U R E 2 . 3 6 . Kochertal Bridgr,
project.
2.7 Pine Valley Creek Bridge, U.S.A.
The first prestressed concrete cast-in-place segmental bridge built in the United States was the
Pine Valley Creek Bridge on Interstate I-8 between
San Diego and El Centro, California, Figures 1.66
and 2.7, opened to traffic late in 1974. This structure is located approximately 40 miles (64 km) east
of San Diego and 3 miles (4.8 km) west of the
gt~nenl vim o f
FIGURE 2.37. Kochertal Bridge, elevation, plan and
cross section.
community of Pine Valley and within the Cleveland National Forest. Interstate I-8 crosses over a
semiarid region that is highly erodible when the
ground cover is disturbed; consequently stringent
c o ntro ls w ere im p o sed o n ac c ess ro ad s and
ground-cover disturbances. Structure type was
influenced by the following factors: site restrictions, economics, ecological considerations, and
Forest Service limitations. After comparing various
possible schemes such as steel arch, deck truss, or
steel box girder, the California Department of
Transportation selected a concrete box girder
bridge medicated on the use of cantilever seg-
(b)
FIGURE 2.38. Kochertal Bridge, typical cross sections. (a) First stage casting. zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
(6) Final stage.
Cast-in-Place Balanced Cantilever Girder Bridges
48
FIGURE 2.39. Kochertal Bridge, cantilever construc-
tion.
mental construction, particularly well suited to the
site because the depth and steep slopes of the valley
made the use of falsework impractical. Also, the
DATUM
LLLV.
cantilever method minimized scarring of the
natural environment, which was a major consideration for a project located in a National Forest.
The bridge has an average length of 17 16 ft (523
m) and consists of twin two-lane single-cell,
trapezoidal box girders each 42 ft (12.8 m) out-toout. The deck is 450 ft (137 m) above the creek
bed. The superstructure consists of five spans of
prestressed box girders 19 ft (5.8 m) deep. The
center span is 450 ft (137 m) in length, flanked by
side spans of 340 ft (103.6 m) and 380 ft (115.8 m),
with end spans averaging 270 ft (82.3 m) and 276 ft
(84.1 m). The bridge was constructed with four
cantilevers. Pier 2 has cantilevers 115 ft (35.1 m) in
length, piers 3 and 4 have 225 ft (69.6 m) cantilevers, and pier 5 has 155 ft (47.2 m) cantilevers,6*7*8
Figure 2.40. Provisions were made in the
design to permit the portions of spans 1 and 5 adjacent to the abutments to be constructed segmentally or on falsework at the contractors’ option. The
later option was exercised by the contractor.g*10zyxwvutsrqpon
w oo;
ELEVATION
FIGURE
2.40.
Pine Valley Creek Bridge, elevation and typical section, from ref. 8.
49
Pine Valley Creek Bridge, USA.
Hinges were provided in spans 2 and 4 at the
end of the main cantilevers. In the preliminary design, consideration was given to the concept of a
continuous structure for abutment to abutment
without any intermediate joints. Continuity has
manv advantages insofar as this particular structure’is concerned. However, it has the significant
disadvantage of large displacements under seismic
loading conditions. Because of the extreme difference in height and stiffness between piers, it was
determined that all the horizontal load was being
transmitted to the shorter piers, which were not
capable of accepting it.s
The pier foundations posed some interesting
construction
problems. The top 20 ft (6 m) of the
rock material at the structure site was badly
fissured, with some fissuring as deep as 40 ft (12
m). Narrow footings only 1 ft (0.3 m) wider than
the pier shafts, tied down with rock anchors, were
preferred to the conventional spread footings to
minimize the amount of excavation.
Although the piers are spectacular because of
their size, they are not unique in concept. The two
main piers, 3 and 4, are approximately 370 ft (113
m) in height and are made up of two vertical cellular sections interconnected with horizontal ties. In
a transverse direction the piers have a constant.
width to facilitate slip-form construction, while in
the longitudinal direction the section varies
parabolically, with a minimum width of 16 ft (4.9
m) approximately one-third down from the top. At
this point they flare out to 23 ft by 24 ft (7 by 7.3 m)
at the soffit. The pier wall thickness is a constant 2
ft (0.6 rn).‘jps
Earthquake considerations produce the critical
design load for the piers. The 1940 El Centro earthquake was used as the forcing function in the design analysis. Design criteria required that the
c,ompleted structural frame withstand this force
level without exceeding stress levels of 75% of
yield. The pier struts are an important element in
the seismic design of the piers. They provide ductility to the piers by providing energy-absorbing
joints and an increased stability against buckling
for the principal shaft elements. Because of the size
of the struts in relation to the pier legs, the majority of the rotation in the strut-leg joint occurs in the
strut. Thus, a very high percentage of transverse
confining reinforcement was required in the strut
to insure the ductility at this location.“j9
Although preliminary design anticipated the
slip-forming technique for construction of the
piers, the contractor finally elected to use a selfclimbing form system. Steel forms permitted 22 ft
(6.7 m) high lifts, and they were given a teflon
coating to facilitate stripping while producing a
high-quality finished concrete surface.
Construction of the pier caps was especially
challenging. The pier caps, Figure 2.41, consist of
two arms 60 ft (18.3 m) in height, which project
outward at an approximate angle of 60” from each
stem of the pier shafts. These arms are constructed
in four lifts in such a manner that the forms for
each lift are tied into the previous lift. Upon completion of the pier cap arms they are tied together
and the top strut is formed, reinforcement placed,
and cast. The pier cap is prestressed transversely in
order to overcome side thrust from-the superstructure.
The superstructure consists of two parallel
trapezoidal box girders 42 ft ( 12.8 m) wide and 19 ft
(5.8 m) deep with a 38 ft (11.6 m) space between
the boxes, such that an additional box girder may
be constructed for future widening, Figures 2.40
and 2.42. The boxes, in addition to being posttensioned longitudinally, have transverse prestressing in the deck slab, together with sufficient mild
steel reinforcement to resist nominal construction
loads, allowing the transverse prestressing operations to be removed from the critical path. The
zyxwvutsrqponmlk
-
I
-I9
t
SC?
-
w
110’
zeo’
zw
6d
iFFooTING
ELEVATKIN
OF PER
i
SIOE MW cf PIER
zyxwvutsrqp
E PIER
1
xSECT- x
IPlERSHlFTl
FIGURE 2.41. Pine Valley Creek Bridge, elevation,
side view, and cross section of pier, from ref. 7 (courtesy
of the Portland Cement Association).
50
Cast-in-Place Balanced Cantilever Girder Bridges
FIGURE 2.42. Pine Valley Creek Bridge, typical box
girder cross section, from ref. 7 (courtesy of the Portland
Cement Association).
sloping webs and large deck overhangs were used
to minimize the slab spans and the number of
girder webs and to accentuate a longitudinal shadow line, thus reducing the apparent depth. The web
thickness of 16 in. (406 mm) was selected to permit
side-by-side placement of the largest tendon then
being used in bridge construction and to keep the
shear reinforcement to a reasonable size and
spacing, Figure 2.42. The bottom slab at midspan
is 10 in. (254 mm) thick and flares out to 6.5 ft (1.98
m) at the pier. 6*7,g Construction of the superstructure proceeded in a balanced cantilever fashion,
Figures 2.7 and 2.43.
As shown in Figure 2.44, the erection scheme
proposed by the contractor allowed all superstructure work to be performed in a continuous
sequence, essentially from the top. Four form
travelers were used for the cantilever construction
of this project, one at each end of each cantilever
arm. Basically, one traveler consisted of an overhead steel truss used to support the formwork for
the typical 16.5 ft (5 m) long segments. The truss is
anchored, at the rear, to the previously cast segment, while the front end is equipped with hydraulic jacks used for grade adjustment. Highdensity plywood was used for all formed surfaces.
A total of 172 cast-in-place segments were required
for the entire structure. Falsework was required
close to abutments 1 and 6 to complete the side
spans beyond the balanced cantilever arms.
Formwork used in that portion of the structure
could be reused above each intermediate pier cap
to construct the 35 ft (10.7 m) long pier segment
before the actual cantilever construction proceeded.
The cross section of the superstructure allowed
PIE
FIGURE 2.43. Pine Valley Creek Bridge, auxiliary
bridge, from ref. 7 (courtesy of the Portland Cement Asso ciatio n).
an auxiliary truss to be located between the two
concrete box girders, Figure 2.43. This auxiliary
bridge consisted of a structural steel truss 10 ft
(3.05 m) square in cross section and 320 ft (97.5 m)
in length. In a stationary position it was supported
at the leading end on the pier cap strut and at the
rear end of a steel saddle between the two concrete
boxes. It was designed such that the front end
could be cantilevered out 225 ft (68.6 m), which is
one-half the main span. Electric winches allowed
longitudinal launching between the concrete box
girders. When pier 5 was completed, the auxiliary
bridge was erected in span 5-6, utilizing temporary support towers near abutment 6. Subsequent
30 ft (9.1 m) lengths of auxiliary truss were attached
at the abutment and incrementally launched toward pier 5, until its front end was supported on the
pier cap. The pier table was then constructed and
cantilever construction commenced until the
structural hinge in span 4-5 was reached. Upon
completion of the closure joint in span 5-6 the auxiliary truss was launched forward until the front
end reached pier 4. The form travelers were dism antled fro m the tip o f the cantilev er and
reerected on the pier table at pier 4, and cantilever
FIGURE 2.44. (Opposite) Pine Valley Creek Bridge,
erection scheme proposed by the contractor, from ref.
10.
0
\__
Stage 1
Cantilever
Construction
from pier 5
construction on
conventional
($l
+L
\
stage 2
from pier 4
\
Stage 3
from pier 3
Stage 5
cOmpletion
52
Cast-in-Place Balanced Cantilever Girder Bridges
construction was started again. This cycle was repeated until closure was achieved in span l-2.
The use of the auxiliary truss had the following
advantageslO:
1. Men and materials for the superstructure
could reach the location of construction from
abutment 6 over the auxiliary bridge and the
already completed portion of the superstructure without interfering with the valley below.
2. The construction equipment (tower cranes and
hoists) at the piers was required only for the
actual construction of the piers and could be
relocated from pier to pier without waiting for
completion of the superstructure.
3. Except for construction of abutment 1 and
pier 2, site installation for the entire project
was located at one location, near abutment 6.
Concrete was supplied from a batching plant located approximately 2 miles (3.2 km) from the site.
Ready-mix trucks delivered the concrete at abutment 6. The concrete was then pumped through 6
in. (152 mm) pipes down the slope to the foot of
piers 5 and 4. The concrete for the superstructure
was pumped through a pipeline installed in the
auxiliary truss right into the forms. A second
pump with a similar installation was located at
abutment 1 to supply concrete for abutment 1 and
pier 2.1°
A 5000 psi (35 MPa) concrete was specified for
the superstructure, presenting no unusual problems. However, to maintain a short cycle for the
construction of the individual segments it was necessary to have sufficient strength for prestressing
30 hours after concrete placement. This was
difficult to achieve, since the specifications did not
allow type III cement and certain additives. A solution Gas to prestress the individual tendons necessary to support the following segment to 50 percent of their final force. The form carrier could
then be advanced and the remainder of the prestressing force applied after the concrete reached
sufficient strength and before casting the next
segment.r”
Prestressing was achieved using lf in. (32 mm)
diameter Dywidag bars. Longitudinal tendons
were provided in 33 ft (10 m) lengths and coupled
as the work progressed. Temporary corrosion
protection of the bars was obtained by blowing
“VPI” powder into the ducts and coating each bar
with vinyl wash or “Rust-Van 310.“*
2.8 Genneuilliers Bridge, France
The Gennevilliers Bridge, Figures 2.45 and 2.46, is
a five-span structure with a total length of 2090 ft
(636 m). At its southern end it is supported on a
common pier with the approach viaduct from the
port of Gennevilliers. It crosses successively an entrance channel to the port, a peninsula situated
between the channel, and the Seine River itself,
Figure 2.47. It is part of the Al5 Motorway, which
traverses from the Paris Beltway through Gennevilliers, Argenteuil, the valley d’Oise, and on to
the city of Cergy-Pontoise. The present structure
provides a four-lane divided highway with provision for a future twin structure.
The superstructure is a variable-depth two-cell
box girder with spans of 345, 564, 243, 564 and
371 ft (105, 172, 74, 172 and 113 m). Depth varies
from 29.5 ft (9 m) at intermediate piers to 11.5 ft
(3.5 m) at midspan of the 564 ft (172 m) spans and
its extremities, with a depth of 23 ft (7 m) at
midspan of the short center span, Figure 2.46.
Depth-to-span ratios of the 564 ft (172 m) spans at
midspan and at the piers are respectively l/49 and
l/19. The curved portion of the structure has a
radius, in plan, of 2130 ft (650 m). The longitudinal grade is a constant 1.5 percent within the zone
of curvature. Because the short center span is subjected to negative bending moment over its entire
length, the structure behaves much as a continuous
three-span beam.
In cross section, Figure 2.48, the two-cell box
girder has a bottom flange varying in width from
42.2 ft (12.86 m) at midspan to 30.5 ft (9.3 m) at
the pier, for the 564 ft (172 m) span. Thickness of
the bottom flange varies from 47 in. (1.2 m) at the
pier to 8 in. (20 cm) at midspan. The top flange has
FIGURE 2.45. Gennevilliers Bridge, view of curved
five-span structure.
Gennevilliers Bridge, France
FIGURE 2.46.
Gennevilliers Bridge, plan and elevation, from ref. 11.
an overall width of 60.6 ft (18.48 m) with a 6 ft
(1.83 m) overhang on one side and 6.2 ft (1.88 m)
on the other. Thickness of the top flange is a constant 8 in. (20 cm). The center web has a constant
thickness of 16 in. (400 mm). Exterior webs, which
are inclined 18” to the vertical, vary in thickness
FIGURE 2.47.
from 16 in. (400 mm) at the pier to 12 in. (300 mm)
at midspan. Diaphragms, Figure 2.49, are located
at the supports. The superstructure is prestressed
in three directions, with strand tendons being
utilized longitudinally and transversely and bar
tendons utilized for the webs. Interior anchorage
Gennevilliers Bridge, aerial view of the completed bridge.
Cast-in-Place Balanced Cantilever Girder Bridges
54
At
Support
At Mid Span
la55
I1
1
611
I
366
356
,
364
,m@
6Za3
1
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366
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366
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FIGURE 2.48. Gennevilliers Bridge, cross section, from ref. 11.
FIGURE 2.49. Gennevilliers Bridge, interior view
showing diaphragm.
blocks for the longitudinal prestressing are located
at top slab level.
The superstructure is fully continuous over its
total length of 2090 ft (636 m) between the northern abutment and the southern transition pier with
the approach viaduct. The deck rests upon the
four main piers supported by large elastomeric
pads. The superstructure was cast in place using
the balanced cantilever method according to the
step-by-step scheme shown in Figure 2.50. Segments over the piers (pier segments) were constructed first on formwork, in a traditional manner, except for their unusual length [26 ft (7.9m)I
and weight [850 t (770 mt)].
Four travelers were used for casting the typical
11 ft (3.35 m) long segments varying in weight
from 242 t (220 mt) near the piers to 110 t (100 mt)
at midspan. l1 The travelers were specially designed
to achieve maximum rigidity and prevent the usual
tendency to crack a newly cast segment under the
deflections of the supporting trusses of conventional travelers. The framework used for this purpose was made of self-supporting forming panels
assembled into a monolith weighing 120 t (110 mt)
and prestressed to the preceding part of the
superstructure to make the unit substantially
deflection free, Figure 2.5 1. Stability, especially
under wind loads or in the event of an accidental
failure of the travelers during the construction period, was maintained by a pair of cables on each
FIGURE
2.50.
Gennevilliers Bridge, erection sequence, from ref. 11.
side of the pier connecting the superstructure to
pier base.
2.9 Grand’Mere Bridge, Canada
This three-lane cast-in-place segmental bridge is
located on Quebec Autoroute 55 and crosses the
St. Maurice River approximately 3 miles (4.8 km)
north of Grand’Mere, Quebec, Figure 2.52. Water
depth at the bridge site is over 110 ft (35.5 m), with
an additional 150 ft (45.75 m) depth of sand, silt,
FIGURE 2.51. Gennevilliers Bridge, superstructure
under
construction.
and debris above bedrock. The river flow at the
bridge site is 3.6 ft/ sec (1.1 m/ set).
During the preliminary design stage in 1973 and
1974, several structural solutions were considered.
The u se o f sho rt sp ans o f p rec ast c o nc rete
AASHTO sections or structural steel girders requiring a number of piers was immediately abandoned because of river depth and current velocity
at the site. Site conditions required the development of an economical long clear span with as few
piers as possible in the river. Options available
FIGURE 2.52. GrandMere
Bridge, general view
showing parabolic soffit of center span, (courtesy of the
Portland Cement Association).
56
Cast-in-Place Balanced Cantilever Girder Bridges
were structural steel, post-tensioned precast segmental, and several options of cast-in-place prestressed concrete, varying in span, cross section,
and pier requirements. The design finally selected
for preparing the bid documents was a concrete
cantilever single-cell box with a center span of 540
ft (165 m), a 245 ft (75 m) western land span, and a
150 ft (46 m) eastern land span for a total length of
935 ft (285 m). The design actually used for construction, Figure 2.53, for the same total length,
has a main span increased to 59.5 ft (181 m) and
two equal 170 ft (52 m) long side spans. The corresponding slight increase in cost of the superstructure was far more than offset by eliminating the
need to build a caisson in 48 ft (15 m) of water 98 ft
(30 m) above bedrock for the west pier. This redesign, developed in cooperation with the contractor,
allowed an overall saving of approximately 20% of
the project cost.
The two identical 170 ft (5 1.9 m) long land spans
cantilever from the main piers and act as counterweights for the main span. From a depth of 32 ft
(9.8 m) at the main piers they taper to a depth of 28
ft (8.5 m) at a point 130 ft (39.6 m) from the main
\
t
\
F I G U R E 2 . 5 3 . Granti’Mcrc Bx-icige. ccntt’~~ $p;tn
parabolic arch soffit (courtesy of’ the Portland Cement
Association).
piers, where they are supported by a secondary
pair of 4 ft by 4 ft (1.2 by 1.2 m) bearing capped
piers. The 40 ft (12.2 m) wedge-shaped shore ends
of the land spans taper from the secondary piers to
grade at the top of the abutment. The abutments,
which are just 16 in. (406 mm) thick, are designed
to support the approach slab only, Figure 2.54.
z
1*
170’
595’
i
ELEVATION
TYPICAL
DETAIL
\
OF
170’
SECTION
ABUTMENT
FIGURE 2.54. Grand’Mere Bridge, general arrangement. (a) Elevation.
(6) Typical section. (c) Detail of abutment.
Grand’mere Bridge, Canada
Modular, confined rubber expansion joints are
provided in the roadway above the abutments. The
wedge portions of the land spans are solid conCrete, helping counterbalance the weight of the
main span under service conditions as well as during the construction stage. The land spans have a
web thickness of 2 ft (0.6 m), a 3 ft (0.9 m) thick
bottom slab, and a 15 in. (38 1 mm) thick top flange.
A 2 ft (0.6 m) thick diaphragm is located 78 ft (23.8
57
m) outboard of the secondary piers to form a
chamber between the solid wedge end and the
diaphragms. This chamber was incrementally filled
with gravel in three stages to counterbalance the
main span as it was progressively constructed. The
bottom soffit of the west land span was supported
on temporary steel scaffolding. However, because
of the terrain slope, the east land-span bottom
soffit was plywood-formed on a bed of sand spread
ElevationzyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFE
fb)
FIGURE 2.55. Arnhem Bridge. (a) Plan. (b) Elevation.
58
Cast-in-Place Balanced Cantilever Girder Bridges
over the rock. Upon completion of concreting and
curing, the sand was hosed out from under the
formwork, allowing it to be stripped.‘”
2.10 Arnhem Bridge, Holland
The Arnhem Bridge, Figure 2.55, is a cast-inplace, lightweight concrete, segmental bridge crossing the Rhine River with a center span of 448 ft
(136 m), a south end span of 234 ft (71 m), and a
north end span of 238 ft (72 m) connecting to approach ramps. It is a dual structure composed of
two-cell box girders, Figure 2.56a. The western
structure has two 30 ft (9.1 m) roadways for automobile traffic. The eastern structure has a 23 ft
(7 m) roadway reserved for bus traffic, a 17 ft (5.3
m) roadway for bicycles and motorcycles, and a 7 ft
(2.1 m) pedestrian walkway. Ramp structures are
of prestressed flat slab construction, Figure 2.566.
The main three-span river crossing with an
overall width of 122.7 ft (37.4 m) consists of two
E,
0.13
j
2.62
1
double-cell box girders that vary in depth from 6.5
ft (2.0 m) at midspan to 17 ft (5.3 m) at the piers.
The western rectangular box girder has a width of
49 ft (14.8 m) with 10 ft (3 m) top flange cantilevers
for an overall width of 68.4 ft (20.84 m). The eastern rectangular box girder has a width of 35.4 ft
(10.8 m) with top flange cantilevers of 8.6 ft (2.62
m) for a total width of 52.6 ft (16.04 m), Figure
2.56a.
Construction of the main spans is by the conventional cast-in-place segmental balanced cantilever
method with form travelers. The form travelers
are owned by the Dutch Government and leased to
the contractors. Strand tendons were used for
post-tensioning, and the lightweight concrete had a
weight of about 110 lb/ ft3 (1780 kg/ m3), Figure
2.57.
Temporary supports at the pier were used for
unbalanced loading during construction, Figure
2.58. Precast exposed aggregate facia units were
used for the entire length of the structure and its
approaches, Figures 2.59 and 2.60.
zyxwvu
VAR
fb)
FIGURE 2.56. Arnhem Bridge, typical cross sections of main bridge and flat-slab
ramp. (a) M ain structure. (6) Prestressed flat-slab ramp.
Napa River Bridge, U.S.A.
FIGURE 2.57. Arnhem Bridge, center-span cantilevers.
FIGURE 2.58. Arnhem Bridge, temporary pier supports for unbalanced moments.
FIGURE 2.59. Arnhem
flat-slab ramp structure.
59
HI-idgc, Ge\\- of‘ prwtwssed
FIGURE 2.60. Arnhem Bridge, precast exposed
aggregate facia units.
2.11 Napa River Bridge, U.S.A.
The Napa River Bridge, Figure 2.61, is located on
Highway 29 just south of the city of Napa, California, and provides a four-lane, 66 ft (20 m) wide
roadway over the Napa River to bypass an existing
two-lane lift span and several miles of city streets.
The 68 ft (20.7 m) wide, 2230 ft (679.7 m) long
bridge consists of 13 spans varying in length from
120 to 250 ft (36.58 to 76.2 m) and a two-cell
trapezoidal box girder varying from 7 ft 9 in. (2.36
m) to 12 ft (3.66 m) in depth, Figure 2.62. Three
hinged joints were provided at midspan in spans 2,
6, and 10. These joints involved elaborate connections incorporating elastomeric bearing pads and
hard-rubber bumper pads to withstand severe
movement and shock during an earthquake, Figure 2.63. All other joints between the cantilevers
were normal cast-in-place closure joints.13
The
superstructure is fixed to the piers, primarily for
seismic resistance.
The Structures Division of the California Department of Transportation (CALTRANS) developed plans and specifications for three alterna-
FIGURE 2.61. Napa River Bridge, aerial view.
tive types of construction, Figure 2.62, as follows:
A. A conventional continuous cast-in-place prestressed box girder bridge of lightweight concrete.
B. A continuous structural-steel trapezoidal box
girder composite with a lightweight concrete
deck.
C. A cantilever prestressed segmental concrete
box girder bridge allowing either cast-in-place
Cast-in-Place Balanced Cantilever Girder Bridges
60
108 +20 PRVC
Elev 63 18
2820’ vc
k/C q -0 24863 % Sto
‘SO&
PROFILE GRADE
Pier 2
3
4
5
6
7
i
9
IO
II
I2
13
ELEVATION
Contllever Segmental P / S
CmveAmolzyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
t ip P/ S
Lightweight
Cone B o x Gtrder
Welded Steel Box Girder
Llghtwelght
Cone B o x Girder
ALTERNATIVE
B
ALTERNATIVE C
ALTERNATIVE
A
FIGURE 2.62.
Napa River Bridge, profile grade, elevation, and alternate sections.
or precast segments. Erection was allowed on
falsework or by the free cantilever method.
Anchor
Bolt
L Elastomeric Pad
FIGURE 2.63. Napa River Bridge, mid-span hinge
joint with seismic bumbers.
Because of poor foundations and a readily available aggregate supply, all alternatives utilized
lightweight concrete in the superstructure. Alternative C utilized transverse prestressing in the deck
to reduce the number of webs to three, as compared to seven webs required in alternative A. Of
seven bids received and opened on November 6,
1974, six were for alternative C and the seventh
and highest was for alternative B. No bids were
submitted for alternative A.
Design of the superstructure required lightweight concrete with a compressive strength of
4500 psi (3 16 kg/cm2) at 28 days and 3500 psi (246
kg/cm2) prior to prestressing. The three-web winning alternative required a minimum of formed
surfaces and forced the majority of longitudinal
prestressing into the flanges, resulting in
maximum prestress eccentricity, and therefore an
economical solution.
Contract plans showed the minimum prestress
force required at each section and permitted the
use of either 270 ksi (1862 MPa) strand or 150 ksi
(1034 MPa) bar tendons. Prestressing force diagrams were provided for both materials. The contractor had the option of balancing segment length
against prestress force to achieve the most economical structure. In addition, the plans provided
the contractor with the option of a combination of
diagonal prestressing and conventional reinforce-
Koror-Babelthuap, U.S. Pacajic
ment in the webs for shear reinforcement or the
utilization of conventional stirrup reinforcement
only. The design was based upon a 40,000 psi (276
MPa) prestress loss for the 270 ksi (1862 MPa)
strand and 28,000 psi (193 MPa) loss for the 150
ksi (1034 MPa) bars. Because the loss of prestress is
a function of the type of lightweight aggregate
used, the contractor was required to submit test
values for approval concerning the materials to be
used and relevant calculations.t4
The contractor elected to use the cantilever
cast-in-place alternative supported on falsework
until each segment was stressed, Figure 2.64.
Falsework bents with ten 70 ft (21.3 m) long, 36 in.
(914 mm) deep, wide-flange girders support each
balanced cantilever. The falsework was then
moved to the next pier, leaving the cantilever
free-standing, Figure 2.65. The entire formwork,
steel girders, and timber forms were lowered by
winches from the cantilever girder after all negative post-tensioning was completed. Positive posttensioning followed midspan closure pours.13
The 250 ft (76.2 m) long navigation span was
constructed with a complicated segment sequence
because of a U.S. Coast Guard requirement that a
70 ft (21.3 m) wide by 70 ft (2 1.3 m) high navigation channel be maintained. Approximately 60 ft
(18.3 m) of span 4, over the navigable channel, was
constructed in three segments on suspended
falsework by the conventional cast-in-place segmental method.13
All transverse and longitudinal post-tensioning tendons consist of t in. (12.7 mm) diameter
strands. Longitudinal tendons are twelve t in. (12.7
mm) diameter strand, with anchorages located in
the top and bottom flanges such that all stressing
was done from inside the box girder. Loops are
used for economy and efficiency, as shown in Figure 2.66. The longest span over the navigation
channel is prestressed by 50 (twelve 4 in. strand)
tendons. Transverse prestress in the top flange allowed a 10 ft (3 m) cantilever on each side of the
two-cell box girder. Transverse tendons consist of
four -f in. diameter strands encased in flat ducts
2.25 by 0.75 in. (57 by 19 mm) with proper splay at
both ends to accommodate a flat bearing at the
edge of the deck slab.
2 .12
FIGURE 2.64. Napa Kiver Bridge, free-standing cantilever and supporting bents for falsework
FIGURE 2.65. Napa
River Bridge, falsework bents
(courtesy of Phil Hale, CALTRANS).
61
Trust Territory
Koror-Babelthuap, U.S. Pacijk
Trust Territory
This structure currently represents (1979) the
longest concrete cantilever girder span in the
world. It connects the islands of Koror and Babelthuap, which are part of the Palau Island chain of
the Caroline Islands located in the United States
Trust Territory some 1500 miles (2414 km) east of
the Philippines, Figure 2.67.
FIGURE 2.66. Napa
tendo ns.
River Bridge, longitudinal loop
62
Cast-in-Place Balanced Cantilever Girder Bridges
FIGURE 2.67. Koror-Babelthuap Bridge, location
map, from ref. 15.
In elevation this structure has a center span of
790 ft (241 m) with side spans of 176 ft (53.6 m)
that cantilever another 61 ft (18.6 m) to the abutments, Figure 1.30. Depth of this single-cell box
girder superstructure varies parabolically from 46
ft (14 m) at the pier to 12 ft (3.66 m) at midspan of
the main span, Figure 2.68. The side span decreases linearly from the main pier to 33 ft 8 in.
(10.26 m) at the end piers and then to 9 ft (2.74 m)
at the abutments. The structure has a symmetrical
vertical curve of 800 ft (243.8 m) radius from
abutment to abutment with the approach roadways
at a 6% grade.15
Superstructure cross section, Figure 1.30, is a
single-cell box 24 ft (7.3 m) in width with the top
flange cantilevering 3 ft 9$ in. (1.16 m) for a total
top flange width of 31 ft 7 in. (9.63 m), providing
two traffic lanes and a pedestrian path. The webs
have a constant thickness of 14 in. (0.36 m). Bottom flange thickness varies from 7 in. (0.18 m) at
midspan of the center span to 46 in. (1.17 m) at the
FIGURE 2.68. Koror-Babelthuap Bridge, parabolic
soffit of main span (courtesy of Dr. Man-Chung Tang,
DRC Consultants, Inc.).
main pier and then to 21 in. (0.53 m) at an intermediate diaphragm located in the end span. This
diaphragm and the one at the end pier form a ballast compartment. Another ballast compartment is
located between the end-pier diaphragm and the
abutment. The bottom flange under the ballast
compartments is 3 ft (0.9 m) thick in order to support the additional load of ballast material. Top
flange thickness varies from 11 in. (0.28 m) at
midspan of the main span to 17 in. (0.43 m) at the
main pier and has a constant thickness of 17 in.
(0.43 m) in the end spans.15
The superstructure is monolithic with the main
piers, with a permanent hinge at midspan to accommodate concrete shrinkage, creep, and thermal movements. The hinge can only transfer vertical and lateral shear forces between the two
cantilevers and has no moment-transfer capacity.15
The superstructure was constructed in segments
w ith the end spans on falsew ork and the main span
in the conventional segmental cantilever manner,
using form travelers. After f-oundations
were completed, a 46 ft (14 m) deep by 37 ft (11.3 m) pier
segment was constructed, Figure 2.69, in three operations: first the bottom flange, then the webs and
diaphragm, and finally the top flange. Upon completion of the pier segment, form travelers were
installed and segmental construction begun. Two
form travelers were used to simultaneously ad-
FIGURE 2.69. E;oror-Baheltlluap Brid g e, p ier seg ment (courtesy of Dyckerhoff & Widmann).
Vejle Fjord Bridge, Denmark
FIGURE 2.70. ~oror-K;tt,clthrl;lp
Bridge. main-span
cantilevers advancing (courtesy of Dyckerhoff PC Widmann).
Vance the main-span cantilevers, Figure 2.70. Segments for this project were 15 ft (4.57 m) in
length. l5
On this project, each segment took slightly more
than one week to construct. A typical cycle was as
follows : I5
1.
2.
3.
4.
5.
6.
7.
When the concrete strength in the last segment
cast reached 2500 psi (17.2 MPa), a specified
number of tendons, ranging from six to 12,
were stressed to 50 percent of their final force,
thus enabling the form traveler to advance in
preparation for the following segment.
Advancing the form traveler also brought forward the outside forms of the box. The forms
were cleaned while rough adjustments of elevation were made.
Reinforcement and prestressing tendons were
placed in the bottom flange and webs. The inside forms were advanced and top flange reinforcement and tendons placed.
After the previous segment concrete had
reached a strength of 3500 psi (24.1 MPa), the
remaining tendons were stressed. The previous segment had to be fully prestressed before
concrete for the subsequent segment could be
placed.
Fine adjustment of the forms for camber and
any required correction was made.
New segment concrete was placed and cured.
When the new segment reached a concrete
strength of 2500 psi (17.2 MPa), the cycle was
repeated.
The structure was prestressed longitudinally,
transversely, and vertically. Three hundred and
two longitudinal tendons were required at the pier
segment. As the cantilever progressed, 12 to 16
tendons were anchored off at each segment, with
eight longitudinal tendons remaining for the last
segment in a cantilever at midspan. As the structure has a hinge at midspan, there were no continuity tendons in the bottom flange. Transverse
tendons in the top flange were spaced at 22 in.
(0.56 m) centers. Vertical tendons were used in the
webs to accommodate shear. Spacing for the vertical web tendons was 30 in. (0.76 m) in the center
span and 15 in. (0.38 m) in the end spans. All tendons were la in. (32 mm) diameter barsI
Side spans were constructed on falsework resting
on compacted fill. The sequence of segmental construction in the side spans was coordinated with
that in the main span, so that the unbalanced moment at the main pier was maintained within prescribed limits.
2.13
Vejle Fjord Bridge, Denmark
This structure crosses the Vejle Fjord about 0.6
mile (1 km) east of the Vejle Harbor. It is part of
the East Jutland Motorway, which will provide a
bypass around the city of Vejle, Denmark. A total
length of 5611 ft (17 10 m) makes it the second
longest bridge in Denmark.
Bid documents indicated two alternative designs,
one in steel and one in concrete. The steel alternative called for a superstructure composed of a
central box girder with cantilevered outriggers
supporting an orthotropic deck and fjord spans of
413 ft (126 m). The second alternative required a
prestressed concrete superstructure with a central
box girder to be constructed by the balanced cantilever method utilizing either precast or cast-inplace segments, with fjord spans of 361 ft (110 m).
The successful alternative was the cast-in-place
segmental prestressed concrete box girder.
The bridge, in plan, is straight without any horizontal curvature. It does have a constant grade of
0.5% falling toward the north. Navigation requirements were a minimum 131 ft (40 m) vertical
and 246 ft (75 m) horizontal clearance. Water
depth in the fjord is generally 8 to 11.5 ft (2.5 to
3.5 m) except at the navigation channel, where the
depth increases to 23 ft (7 m). Under the fjord bed
are layers of very soft foundation materials, varying in depth from 26 to 39 ft (8 to 12 m). Therefore, the piers in the fjord are founded on 8 in. (0.2
m) square driven reinforced concrete piles varying
in length from 100 to 130 ft (30 to 40 m), Figure
2.71. Piers on the south bank are founded on
64
Cast-in-Place Balanced Cantilever Girder Bridges
m
CAST I FlXCD rcmuwan
USTw4nEw.Y
_---_------_..,--as1
zyxwvutsrqpo
R E D PDII
I L ErsS
0 150 CM
MICACEOUS
,‘,“‘,‘.‘,‘,‘,‘,
FIG U RE 2.71. Vejle Fjord Bridge, fjord piers
FIGURE 2.72.
founded on driven reinforced concrete piles.
on bored piles.
bored reinforced concrete piles, 59 in. (1500 mm)
in diameter, Figure 2.72. On the north bank one
pier is founded on driven reinforced concrete piles
and one is supported directly on a spread footing.
The cross section of the bridge, Figure 2.73,
which carries four traffic lanes with a median barrier, is a variable-depth single box with a vertical
web and prestressed transverse ribs. Total width
between edge guard rails is 87 ft (26.6 m). Box
girder width is 39.4 ft (12 m), with a depth varying from 19.7 ft (6 m) at the pier to 9.8 ft (3 m)
at midspan. Each segment is cast with a length of
11.3 ft (3.44 m). Transverse top flange ribs are
spaced at 22.6 ft (6.88 m) centers-that is, every
other segment joint.
The total bridge length is divided into four separate sections by three expansion joints located at
the center of spans 4-5, 8-9, and 12-13. Longitudinal prestress is achieved by Dywidag (twelve
Vejle
Bridge, land .piers founded
I Fjord
_
0.6 in. diameter strand) tendons, as are the transverse prestress in the top slab and the continuity
prestress in the bottom slab.
A 492 ft (150 m) long steel launching girder and
two special form travelers were used for casting in
place the full width of the 11.3 ft (3.4 m) long segments in balanced cantilever. Insulating forms
followed the form travelers in order to prevent the
formation of fissures due to adverse temperature
gradients. In addition, the steel girder stabilized
the concrete structure during construction and was
used for the transportation of materials, equipment, and working crew. The total weight of the
girder including the two travelers was approximately 660 t (600 mt). A typical longitudinal section of a cantilever is shown in Figure 2.74, along
with the structure erection procedure.
Work on the bridge started in the summer of
1975 and was scheduled for completion in 1980.
C R O S S S E C T I O N 1:200
2660
l
50
300
50
750
5010030
AT MID SPAN
FIGURE
2.73.
- - I - -
OVER
PIER
Vejle Fjord Bridge, elevation, plan, and cross section.
BOX-TYPE GIRDER
z
LONGITUDINAL
POSITION
SECTION
OF
PRESTRESSING
TENDONS
I
I
SUPERSTRUCTURE, PRINCIPLE OF EXECUTION
FIGURE
2.74.
AUXILIARY EQUIPMENT ETC.
CONSTRUCTION
Vejle Fjord Bridge, longitudinal section and erection sequence.
PRINCIPLES
67
Vejle Fjo rd Bridge, D enmark
::
\\\ \\
\
\\\\
\\
FIGURE 2.75. Vejle Fjord Bridge, launching girder.
,:<
\
\
\\
>~‘
.:,
FIGURE 2.77. Vejle Fjord Bridge, pier segment with
diaphragm.
FIGURE 2.78. Vejle Fjord Bridge, construction \iew,
spring 1978 (courtesy of H. A. Lindberg).
\
;->, :
\
I_\\
1
FIGURE 2.76. Vejle Fjord Bridge, transverse ribs.
Construction progress in the spring of 1978 is illustrated in Figures 2.75 through 2.78. Figure 2.79
is an aerial view showing the structure nearing
completion. To keep within the construction
schedule, it was finally necessary to use two complete sets of launching girders and twin travelers
working simultaneously from both ends of the
bridge.
FIGURE 2.79. Vqjle Fjord Bridge. aerial view from
the northwest.
Cast-in-Place Balanced Cantilever Girder Bridges
68
2.14
Houston Ship Channel Bridge, U.S.A.
This bridge, a rendering of which is shown in Figure 1.67, includes a main structure over the Ship
Channel in Houston, Texas, and tw o approach
viaducts. The main structure is a three-span continuous box girder, cast in place in balanced cantilever. Span lengths are 375, 750, and 375 ft (114,
229, and 114 m). The navigation channel is 700 ft
(213 m) wide at elevation 95 ft (29 m) and 500 f-t
(752 m) wide at elevation 175 ft (53.4 m), Figure
2.80.
The three-web box girder carries four traffic
lanes separated by a 2 ft 3 in. (0.7 m) central barrier and has two 3 ft 9 in. (1.14 m) parapets. The
box girder is fixed to the top of the main piers to
make the structure a three-span rigid frame. Support for the box girder is provided by elastomeric
bearings on top of the transition piers, where it is
separated from the approach viaducts by expansion joints.
shrinkage, superimposed dead loads, and live
loads). They are, therefore, heavilv reinforced;
their dimensions are:
Total height (from top of footing to bottom of pier
segments): 160 ft 10 in. (49 m)
Length (parallel to centerline of highway): 20 ft
constant (6.1 m)
Width: variable from 38 ft at the bottom to 27 ft 7
in. at the top (11.6 to 8.4 m)
Pier cross section: rectangular box, with 2 ft (0.6 m)
constant w all thickness
The transition piers support the last segment of
the main structure side span and the last span of
the approaches. The pier shaft is a rectangular box
with 1 ft 4 in. (0.4 m) thick walls. Their heights are
152 ft (46 m) at one end and 164 ft (50 m) at the
other end of- the bridge. The length, parallel to the
centerline of the highway, varies from 18 to 8 ft
(5.5 to 2.4 m); the width is 38 ft (11.6 m) constant.
Atop
the pier, a 6 ft 8 in. (2 m) cap carries the perFoundations
The two center piers and two tranmanent
elastomeric bearings and all the temporary
sition piers rest on 24 in. (610 mm) diameter
jacks
and
concrete blocks that will be used at the
driven steel pipe piles. The center piers each rest
time
of
the
side-span closure pour. All four piers
upon 255 piles w ith a unit pile capacity of 140 t
are
slip-formed.
(127 mt). Footings are 81 ft (24.7 m) wide, 85 ft (26
Box Gzrder Superstructure Dimensions of the
m) long, and 15 ft (4.6 m) deep. These footings are
variable-depth
box girder were dictated by verv
surrounded by a sheet pile cofferdam and are
stringent
geometry
requirements. Vertical alignpoured on a 4 ft (1.2 m) thick subfooting seal conment
of
the
roadway
was determined by the
crete. The transition pier footings are 50 ft (15.2
maximum
allowable
grade
of the approach viam) wide, 35 ft (10.7 m) long, and 5.5 ft (1.7 m)
ducts
and
the
connection
thereof
with the roadway
thick and rest on 70 piles each of 100 t (90 mt)
system
on
both
banks.
The
clearance
required fat
bearing capacity.
the ship channel left, therefore, only a structural
depth of 2 1.8 ft (6.6 m) at the two points located
Piers The main piers provide for the stability of
250 ft (76 m) on either side of the midspan section.
the cantilevers during construction (unbalanced
The
soffit is given a third-degree parabolic shape
construction loads and w ind loads) and participate
to
increase
the structural depth near the piers in
in the capacity and behavior of the structure under
order
to
compensate
for the very lirnited height of
service loads (long-term loads due to creep and
FIGURE 2.80.
Houston
Ship
Channel
Bridge,
longitudinal
section.
Houston Ship Channel Bridge, USA
the center portion of the main span. Maximum
depth at the pier is 47.8 ft (14.6 m), with a spanto-depth ratio of 15.3. Minimum depth at midspan
is 15 ft (4.6 m), with a span-to-depth ratio of 49.
Over the 500 ft (152 m) center portion of the main
span the span-to-depth ratio is 23, compared to a
usual value between 17 and 20. Typical dimensions
of the box section are shown in Figure 2.8 1. Posttensioning is applied to the box section in three
dimensions:
69
Longitudinal prestress is provided by straightstrand tendons (twelve 0.6 in. diameter or nineteen
0.6 in. diameter strands), as shown schematically in
Figure 2.82.
Transversely, the top slab is post-tensioned by tendons (four 0.6 in. diameter strands) in flat ducts
placed at 2 ft (0.6 m) centers.
Vertically, the three webs are also post-tensioned as
prescribed in the specifications to a minimum
tzyxwvutsrqponmlkjihgfedcb
k
FIGURE 2.81.
T r a n s v e r s e
Br i dge
Houston Ship Channel Bridge, box section.
t e n d o n s
4 x O.G;
Ca n t i l e v e r
Tendons
FIGURE
pr est r ess
ov er
main
pier s
/
/
2.82.
( 1 2 x
0 6%
a..ond (19x O.6’dia.
I
Co n t i n u i t y
pr est r ess
at
Houston Ship Channel Bridge, longitudinal prestress.
mid
-span
70
Cast-in-Place Balanced Cantilever Girder Bridges
FIGURE
2.83.
Houston Ship Channel Bridge, details of travelers
compressive stress equal to 3Ji; that is, 230 psi (1.6
MPa) for a concrete strength J‘ i = 6000 psi (41.4
MPa).
Details of the form traveler are shown in Figure
2.83.
Pier segments over the main piers are of unusual
size and posed a very interesting design problem,
arising from the transfer of the superstructure un-
balanced moments into the pier shafts. Additional
vertical post-tensioning tendons are provided in
the two 2 ft (0.6 m) thick pier diaphragms for this
purpose. End segments over the transition piers
were designed to allow either the approaches or
the main structure to be completed first, as these
are two separate contracts.
It is possible to make an adjustment at the end
piers to compensate either for differential settle-
71
Other Notable Structures
zyxwvutsrqpon
(a)
2.15.1 MEDWAY
BRIDGE, U.K.
One of the first very long-span cantilever bridges
was the Medway Bridge. This structure used a series of temporary falsework bents to provide stability during construction, Figure 2.84.
2.15.2 RIO TOCANTINS
BRIDGE, BRAZIL
This structure has a center span of 460 ft (140 m)
and tw o side spans of only 174 ft (53 m), Figures
2.85 and 2.86.
2.15.3 PUENTE
FIGURE 2.84. Xlrti~av Bridge, U.K. ((I) I‘)pical
struction sequence. (h) View of’ finished bridge.
COII-
ments or for any deviation of the deflections from
the assumed camber diagram used for construction.
Provisions have been made for unexpected additional concrete shrinkage and creep problems;
empty ducts have been placed in the pier segment
diaphragms and at midspan to allow for future
possible installation of additional tendons located
inside the box girder but outside the concrete section, should the need for such tendons arise.
DEL AZUFRE, SPAIN
This bridge is located very high over a deep canyon
of the Rio Sil. Cantilever cast-in-place was the ideal
answ er to allow construction w ith a minimal contact with the environment, Figures 2.87 and 2.88.
2.15.4 SCHUBENACrlDIE
BRIDGE, CANADA
This three-span bridge with a center span of 700 ft
(213 m) crosses the Schubenacadie River, near
Truro, Nova Scotia. High tidal range, swift currents, ice, and adverse climatic conditions made
the construction of this structure very challenging,
Figures 2.89 and 2.90.
2.15.5
INCIENSO
BRIDGE,
GUATEMALA
2.15 Other Notable Structures
There are so many outstanding and interesting
cast-in-place cantilever bridges in the world today
that it is impossible to discuss the subject adequately in the space available here. Mention should
be made, however, of several notable structures
not yet covered by a detailed description.
The main three-span rigid frame structure with a
center span of 400 ft (122 m) is of cast-in-place balanced cantilever construction, and the approach
spans are of precast girders, Figures 2.91 and 2.92.
The very severe 1977 earthquake left the center
structure completely undamaged, while the usual
damage took place in the approach spans.
Cast-in-Place Balanced Cantilever Girder Bridges
72
1
FIGURE 2.85.
1.72S
j6.55
1
1.725
1
Rio Tocantins Bridge, Brazil, typical elevation and cross section.
2.15.6 SETUBAL BRIDGE, ARGENTINA
This three-span structure with a main span of460
ft (140 m) rests on two main river piers with twin
vertical walls and piles, with a transition footing at
water elevation, Figures 2.93 and 2.94.
2.15.7 KIPAPA STREAM BRIDGE, U.S.A.
This bridge is located in the Island of Oahu in the
State of Hawaii. The dual structure has an overall
width of 118 ft (36 m) to accommodate six traffic
lanes, three in each direction, and consists of two
double-cell box girders of constant depth with
interior spans of 2.50 ft (76.2 m), Figures 2.95 and
2.96. Construction was by cast-in-place cantilever
with segments 15 ft 3 in. (4.65 m) long. The bridge
has pleasant lines, which blend aestheticallv with
the rugged deep-valley site.
2.15.8
PARROTS FERRY BRIDGE, U.S.A.
This structure, built in California for the Corps of
Engineers, represents a major application of lightweight concrete for cast-in-place cantilever construction, Figure 2.97.
2.15.9
FIGURE 2.86. Rio Tocantins Bridge, Brazil, view of
the finished bridge.
MAGNAN
VIADUCT,
FRANCE
Located just off the French Riviera in Southern
France, this four-span continuous structure rests
on 300 ft (92 m) high twin piers of an I-shaped
section. Superstructure was cast in place in two
stages (first the bottom slab and webs and then the
top slab) to reduce the weight and cost of travelers.
Figures 2.98 and 2.99 show the principal dimensions and views of one cantilever and the finished
structure, Figure 2.100.
73
O t her No t able St ruct ures
6S.00
130 00
I
*+t
cm
I
I
I
Ad
*
I
FIGURE
2.15.10 PUTE4UX
2.87.
d
Puente de1 Azufre, Spain, typical elevation and sections.
BRIDGE, FRANCE
These are twin bridges crossing the Seine River
near Paris. Because of very stringent clearance and
geometry requirements, the available structural
depth was only 5.9 ft (1.8 m) for the clear 275 ft
(83.8 m) span and 4.8 ft (1.47 m) for the clear 214
ft (65.3 m) span, making both structures very slender, Figures 2.101 and 2.102. Stiff “V” piers in
both structures help reduce the flexibility of the
deck.
2.15.11 TRICASTIN BRIDGE, FRANCE
This structure spans the Rhone River with no piers
in the river, which necessitates a long center span
and two very short side spans anchored at both
ends against uplift. The center portion of the main
span is of lightweight concrete, while the two zones
over the piers where stresses are high are of conventional concrete, Figures 2.103 and 2.104.
2.15.12 ESCHACHTAL BRIDGE, GERM ANY
FIGURE 2.88.
Puente &%I Azuir e, Spun.
This bridge is located near Stuttgart, Germany.
The superstructure consists of a large single-cell
box girder with large top flange cantilevers supported by precast struts. Because of the weight involved, the central box was cast in one operation;
struts were installed and flanges cast subsequently,
Figures 2.105 and 2.106.
Elevation
16'~0"
1
16’4”
Q
I
,6,-o”
i
16’4”
Section at Midspan
II
20’~0”
Section pver
4
Piers
FIGURE 2.89. Shubenacadie Bridge, elevation and sections, from ref. 16.
FIGURE 2.90. Shubenacadie Bridge, supper t avstem
for unbalanced cantilever moment at pier (courtesy of
the Portland Cement Association).
74
FIGURE 2.91. Incknso Bridge, Guatemala, view of
the structure.
ELEVATION
@@Gp
7 50
MAIN
‘/2
SECTION
ON SUPPORT
BRIDGE
‘/2
SECTION
ON SPAN
FIGURE 2.92. Incienso Bridge, Guatemala, dimensions.
FIGURE 2.93. Setubal Bridge, Argentina, dimensions.
75
FIGURE
brid ge.
2.94.
Setubal Bridge, Argentina, view of the
Abut 2
1
3
2
29’&”
c
4
5
Elevation
-...v.&-.-.-__.
7
6
29’4”
._~ . __. i)
~~,2,,
-- - FIGURE
2.95.
Kipapa Stream Bridge, elevation and cross section.
FIGURE 2.96. Kipapa Stream Bridge, construction
view (courtesy of Dyckerhoff & Widmann).
FIGURE
2.97.
Ferry Bridge,
ref. 17.
=2?
Parrots
dimensions,
COUPE LONGITUDINALE
??
99
0 T
00
@ UC
FIGURE 2.98. Magnan
Viaduct, longitudinal section.
FIGURE 2.99.
Ilagnan Viaduct, view of a cantilever.
FIGURE 2.100. Magnan Viaduct, aerial view of the
completed bridge.
FIGURE
2.101.
Puteaux Bridge, aerial view of the completed bridge.
Ill1 rlnrn - Ml
zyxw
h
“t/ \
10.00
5 00
5.00
c
1 2 . 4 0
++-j 2 . 4 0
FIGURE 2.103. Tricastin Bridge, dimensions.
79
FIGURE 2.104. Tricastin Bridge, view of finished bridge.
FIGURE 2.105. Eschachtal Bridge, casting flange
FIGURE 2.106. Eschachtal Bridge, view of outrigger
cantilevers.
struts.
80
81
References
2.16
Conclusion
8. Ri c h ard A . D o k k e n ,
Segmental
‘I‘he I~;III\~ structures described above show the
versatilitv of’ cast-in-place balanced cantilever construction, particularl\~ in the field of vet-v-long-span
bridges with tew repetitive spans. The design aspect 01‘ these structures will be discussed in Chapter 4 attd construction problems in Chapter 11.
Bridge
“ CAL.I‘RANS
Design.”
Structures, Departmenr
California, Vol. XVII,
9 . A . P . Berzone,
Ex p e rie n c e in
Bririp Sotu, Division of
ot .[‘I-ansportation,
No. 1, March 1975.
“ Pi n e
V al l e y
C re e k
D e sig ning f o r Se g m e ntal C o nstru c tio n,”
State of
Bridge\leeting
Prepr-int 1 9 4 4 , AXE N a t i o n a l S t r u c t u r a l E n gineering hleeting, April 9-13, 1973, San Fra nc isc o .
10.
Richard
Heinen, “ Pine Valley
Creek Bridge: Use ot
Cantilever Construction,”
Meeting
Preprint 198 I,
ASCE Sational Strucrural
En g i n e e ri n g M e e ti n g ,
References
.-\pt-il
9- 13, 1973, Sail Franc isc o .
1 1. “ A . 15 e t A .86 rac c o rd e m e nt
1. H. I ‘llU l, “RlY lc Le nt M ll,”
JAIJI-g;~t~g.
2 .
Hc tt 5 .
Bdot/- uuct S~nhlh/e~r//~crtc, 6 1
\ I;ti 1966.
L‘lt-ich Fitistrrwaldet-.
“ Prestressed C o nc rete BI-idge
<:onst~~tction.” Jounrcll of tha .4ttrwicntr
tlrtr. Vol. 62. So. 9, Seprember
1965.
3 .
I)~711/,/1~-Br,-rchlr, 4-1967, S e p t e m b e r 1 9 6 7 ,
I)\ckerhof’t
4 . L‘lt-ic-h
Kc;, hlunich,
Cantilever
Germanv.
autoroutiel-
ties hauls-de-seine,”
dans le
Ministere
d e L’Equipemenr
D i r e c t i o n Departemental
de
L’Equipement
des Hauts-de-Seine, Paris, September
1976.
C o Ka m - tio ~l
‘ VfWS,
ft Po st- tensio ned Sp a n , ”
2, 1976.
Hmy~
A ug ust
IS. “N a pa Ri v e r Bri d g e , Sapa, Calif~~rnia,” Po rtland
C e m e nt A sso c iatio n, Brid g e Re p o rt, SR 194.01 E,
1977.
Construction
14. “ A lte rnate Bid d ing
f o r C alif o rnia’ s
Napa
Ri v e t
Bridge Won by
Cast-in-Place Prestressed
Concrete
Segmental
Construction,” Prestressed
Concrete
In-
or/ r
stitu te, Post-‘I‘ensioning
Dr.\rgt/ ,
Hr/f/gfj
A m eric an
r\Cl Pu b lic atio n- SP- 23,
C o nc rete
Papet
D etro it,
F. I)o\vning,
Scenery
With Cantilevered
.\‘Pu~.\-RPCO~, June 18, 1964.
“ Cantilever Segmental Prestressed
I.0 .-\ngeles. Caliti~rnia.
“ Pine \‘alle\
So\,ember
Creek Bridge,
port SK 16 1 .O 1 E.
Skokie. 111.. 19i4.
1 I- 15, 19i3.
Calit’ornia,”
Portland
Bridge
Re-
Cement Association,
D iv isio n , Sp e c ial Brid g e
Rep o rt.
1 5 . hian-Chung
“ Bridge Built :Itop the
I‘m\ elcrc,” Etrgitrwrttcg
Dale
Institu te.
Cast-in-Place Construction of’ the Pine Valley Creek
Bridge.” presented to the X.-\SHO Annual &leering,
,
departement
01 Prestressed CoIlcrete
Brid g e s an d MushroonShaped Bridges,” I;/ ,.\[ I~tprrccct~or~rtl
Svmpo.tiu~, Cow
SP23-26.
1969.
6.
K- Widmann
Fiiister~valder, “ Fi-ee
du
12. “ Brid g e H as 595
L’lricti Finster\val(ler. “ Se w D e v e lo p m e n ts in Prestreshing .\Iethotls and C o nc re te Brid g e Construcrioti.”
.5.
COPIUP~P It~cti-
nerd
‘T a ng, “ Koror-Babelthuap B r i d g e - A
World Record Span,” Preprint Paper 3441,
Convention, Chicago, October 16-20,
1978.
1 6 . D . W . Macintosh
Shubenacadie
ASCE
and R. A . W hitm an, “ The
Bridge,
nual
Conference
tion Xssociation
.Maitland,
Nova Scotia,” AnPreprints, Roads and ‘I‘ransportaof Canada, Ottawa, 1978.
15. “ C o nc re te A lte rn ate W in s C o m p e titiv e Bid d in g
Contest f’or Long Span California Bridge,” Bridge
Report,
PostGensioning
Institute,
April
1977.
zy
3
Precast Balanced Cantilever Girder Bridges
3.1
3.2
3.3
3.4
INTRODUCITON
CHOISY-LE-ROI BRIDGE AND OTHER STRUCTURES IN GREATER PARIS, FRANCE
PIERRE BENITE BRIDGES NEAR LYON, FRANCE
OTHER PRECAST SEGMENTAL BRIDGES IN PARIS
3.4.1
3.4.2
3.4.3
3.4.4
3.14 B-3 SOUTH VIADUCTS, FRANCE
3.15 ALPINE MOTORWAY STRUCTURES, FRANCE
3.16 BRIDGE OVER THE EASTERN SCHELDT, HOLLAND
3.17 CAPTAIN COOK BRIDGE, AUSTRALIA
3.18 OTHER NOTABLE STRUCTURES
Paris Belt (Downstream)
Paris Belt (Upstream)
Juvisy Bridge
Twin Bridges at GmfIans
3.18.1
Calix Bridge, France
3.18.2
Vail Pass Bridges, U.S.A.
Tent Viaduct, U.K.
3.18.3
3.18.4
L32 Tauernautobahn Bridge, Austria
3.18.5
Kishwaukee River Bridge, U.S.A.
3.18.6
Kentucky River Bridge, U.S.A.
3.18.7
I-205 Columbia River Bridge, U.S.A.
3.18.8
Zilwaukee Bridge, U.S.A.
Ottmarsheim Bridge, France
3.18.9
3.18.10 Overstreet Bridge, Florida, U.S.A.
3.18.11 F-9 Freeway, Melbourne, Australia
3.5
OLERON VIADUCT, FRANCE
3.6
CHILLON VIADUCT, SWITZERLAND
3.7
HARTEL BRIDGE, HOLLAND
3.8
RIQNITEROI BRIDGE, BRAZIL
3.9
BEAR RIVER BRIDGE, CANADA
3.10 JFK MEMORIAL CAUSEWAY, U.S.A.
3.11 SAINT ANDRk DE CUBZAC BRIDGES, FRANCE
3.12 SAINT CLOUD BRIDGE, FRANCE
3.13 SALLINGSUND BRIDGE, DENMARK
3.1
zyxwvut
Zntroduction
As indicated in Chapter 1, precast segmental construc tio n had its o rig ins (in the contemporark
sense) in France in 1962 as a logical alternative to
the cast-in-place’ method of construction. To the
advantage of segmental cantilever construction,
primarily the elimination of conventional falsew o rk, the tec hniq u e ad d s the ref inem ents im plicit in the use of precasting.
The characteristics of precast segmental construction are:
1. Fab ric atio n o f the seg m ents c an b e ac c o m plished while the substructure is under construction, thus enhancing erection speed of the
superstructure.
2.
82
BY virtue of precasting and therefore maturity.
of the concrete at the time of erection, the time
required for strength gain of the concrete is
removed from the construction critical path.
REFERENCES
3.
As a result of the maturity of the concrete at
the time of erection, the effects of concrete
shrinkage and creep are minimized.
4. Superior quality control can be achieved
factory-produced precast concrete.
for
However, geometric control during fabrication of
segments is essential, and corrections during erection are more difficult than for cast-in-place segmental construction. In addition, the connection
of longitudinal ducts for post-tensioning tendons
and the continuity of reinforcing steel, if they are
required in the design, are less easily achieved in
precast than in cast-in-place methods.
Although precast segmental had been used as
early as 1944 f o r the Lu z anc y Brid g e o v er the
Xlarne River, Figure 1.27, wide acceptance began
lvhen match-casting techniques were developed.
Basically, the principle of fabrication of precast
segments is to cast them in a series one against the
other in the order in which they are to be assem-
Choisy-le-Roi Bridge and Other Structures in Greater Paris, France
bled in the structure. The front face of a segment,
thus, serves as a bulkhead for casting the rear face
of the subsequent segment. Methods of fabrication
of precast segments will be discussed in Chapter
11.
Seg m ents are erec ted in b alanc ed c antilev er
starting from a segment over the pier, which is the
first to be placed. Modifications to the initial principle hau e further inc-rea>& the %,ex;lbcl(clty
of eye<tion procedures. Two major modifications are (1)
temporary prestress ties to secure two or more successive segments and thus free the erection equipment, and (2) cantilever prestressing tendons anchored inside the box sections instead of at the
segment face as on early structures. These refinements mean that the placing of segments and the
threading and stressing of tendons become independent operations.
Efficient application of this method has resulted
in the use of cantilever construction in moderateto small-span structures where it had previously
been considered uneconomical. Examples are the
B-3 So u th V iad u c t (Sec tio n 3.14) c o m p o sed o f
spans ranging from 98 ft (30 m) to 164 ft (50 m)
and the Alpine Motorway Bridges (Section 3.15)
where the spans range between 60 ft (18 m) to 100
ft (30 m).
It is interesting to note a constant evolution toward increased transverse dimensions and weight
o f p rec ast seg m ents. Pro b lem s in p rec asting ,
transporting, and placing segments that are constantly b ec o m ing heav ier and w id er are b eing
progressively resolved. Chapter 4 will deal with this
progressive evolution as applied to some French
precast segmental bridges and will discuss typical
cross sections of some precast segmental bridges
constructed or in the design stage in the United
States.‘.*
In continuous structures expansion joints may
be spaced very far apart. Continuous bridges up to
3300 ft (1000 m) in length have been constructed
without intermediate joints; however, this may not
be an upper limit, provided that the design of
bearings and piers is correctly integrated into the
total design of the structure. Free longitudinal
movement of the bridge due to creep and temperature change is allowed for by placing the structure
on elastomeric or sliding (teflon) bearings. We can
also u se p ier flexib ility to ac c o m m o d ate these
movements by fixing the superstructure to the
piers. In this case, flexibility can be obtained either
by pier height or by the use of single or double
thin-slab walls, thus reducing the piers flexural resistance.
83
The first precast segmental bridge to be built on
the N o rth A m eric an C o ntinent w as the Liev re
River Bridge on Highway 35,s miles (13 km) north
of Notre Dame du Laus, Quebec, with a center
span of 260 ft (79 m) and end spans of 130 ft (40
m), built in 1967. It was followed in 1972 by the
Bear Riv er Brid g e, Digby, N o v a Sc o tia (Sec tio n
3.9), with six interior spans of 265 ft (81 m) and
end spa-m of ‘Lo4 ft (65i -i-ix\. The 3FU KcnQxia(
Causeway, Corpus Christi, Texas (Section 3.10),
opened to traffic in 1973, was the first precast segmental bridge to be constructed in the United
States. In the United States, as of this writing, the
authors are aware of more than 30 precast segmental bridge projects that are either completed,
under construction, or in the design stage. Some
are listed in Table 3.1 .3
3.2
Choisy-le-Roi Bridge and Other Structures in
Greater Paris, France
The first bridge to use the precast segmental cantilever technique with epoxied match-cast joints
was built at Choisy-le-Roi near Paris between 1962
and 1964. It carries National Highway 186, a part
of the Paris Great Belt system, over the Seine River
just east of Orly Airport, Figure 3.1. This structure
is a three- sp an c o ntinu o u s b rid g e o f c o nstant
depth with end spans of 123 ft (37.5 m) and a
center span of 180 ft (55 m), Figures 3.2 and 3.3.
This bridge replaced one constructed in 1870,
which had a superstructure of six steel girders with
fiv e sp ans o f ap p ro xim ately 75 ft (23 m ). This
structure, determined to be no longer adequate as
early as 1939, was severely damaged during World
War II. It in turn had replaced an ancient bridge
of five 66 ft (20 m) oak arch spans designed by
the fam o us m athem atic ian Claud e-Lo uis-M arie
Navier.4
In 1961, a stu d y b y the A d m inistratio n o f
Bridges and Roads allowed two options, one in
prestressed concrete and the other in steel, each
having three continuous spans of 123 ft (37.5 m),
180.4 ft (55 m), and 123 ft (37.5 m). Four prestressed concrete solutions were considered. The
successful solution is illustrated in Figure 3.2.
The overall width of the superstructure for this
dual bridge is 93.2 ft (28.4 m), Figure 3.3. Each
bridge consists of two single-cell rectangular box
girders. The superstructure accommodates dual
two-lane roadways of 23 ft (7 m), two 13 ft (4 m)
sidewalks, and a 10 ft (3 m) median.4*5
Individual
box girders have a constant depth of 8.2 ft (2.5 m),
Precast Balanced Cantilever Girder Bridges
84
T ABLE
3.1.
Name and Location
Lievre River, Notre Dame
du Laus, Quebec
Bear River, Digby,
Nova Scotia
JFK Memorial Causeway,
Corpus Christi, Texas
Muscatuck River, U.S. 50,
North Vernon, Indiana
Sugar Creek, State Route 1620,
Parke County, Indiana
Vail Pass, I-70 West of Denver,
Colorado (4 bridges)
Penn DOT Test Track Bridge,
Penn Sate University,
State College, Pa.
Turkey Run State Park
Parke County, Indiana
Pasco-Kennewick, Columbia
River between Pasco
and Kennewick, Washington
(cable-stay spans)
Wabash River, U.S. 136,
Covington, Ind.
Kishwaukee River, Winnebago CO .
near Rockford, Ill.
(dual structure)
Islington Ave. Ext., Toronto,
Ontario
Kentucky River, Frankfort, Ky.
(dual structure)
Long Key, Florida (contract let
late 1978)
Linn Cove, Blue Ridge
Parkway, N.C.
(contract let late 1978)
Zilwaukee, Michigan
(dual structure)
(bids opened late 1978)
Precast Segment al Concret e Bridges in Nort h America
Date of
Construction
Method of
Construction”
Span Lengths,
tt (m)
1967
B.C.
1972
B.C.
1973
B.C.
1975
B.C.
1976
B.C.
1977
B.C.
1977
O.F.
130-260- 130
(39.6 - 79.2- 39.6)
203.75 - 6 (12 265 - 203.75
(62.1 - 6 ((I 80.77 - 62.1)
loo-200- 100
(30.5 - 6 t - 30.5)
95 190-95
(29 - 58 - 29)
90.5 - 180.5 - 90.5
(27.6 - 53 - 27.6)
134 - 200 - 200 - 134
(40.8 - 61 - 61 - 40.8)
134-200-200145
(40.X-61 -61 -44)
151-155-210-210-154
(46-47.2-64-64-47)
153-210-210154
(46.6 - 64 - 64 - 47)
124
(37.8)
1977
B.C.
1978
B.C.
1978
1.L.
1979
B.C.
1979
B.C.
1979
B.C.
S.S.
P.P.
B.C.
93.3 - 4 (a 187 - 93.5
(28.5 - 3 (@ .57 - 28.5)
170-3 @I 250- 170
(51.8 - 3 G 76.2 - 51.8)
2 @ 161 -200-5 @ 272
(2 @ 49 - 61 - 5 (if X3)
228.5 - 320 - 228.5
(69.6 - 97.5 - 69.6)
II3 - 101 fin 118 - 113
(34.4 - 101 @i 36 - 34.4)
9X.5- 163-4@ 1X0- 163-98.5
(30 - 49.7 - 4 Q 54.9 - 49.7 - 30)
26 N.B. spans total length
8.087.5 (2,465)
25 S.B. spans total length
8.0575 (2,456)
maximum span 392 (119.5)
“Method-of-construction notation: B.C.-balanced cantilever, l.L.-incremental
placement, S.S.-span-by-span.
top flange width of 21.65 ft (6.6 m), and a bottom
flange width of 12 ft (3.66 m). Webs have a constant thickness of la in. (0.26 m), and the top
flange is of constant section throughout its length
with a minimum thickness of 7 in. (0.18 m) at its
180 - 1x0
(54.9 - 54.9)
406.5 - 98 1 - 406.5
(124 - 299 - 124)
launching, O.F.-on talsework. P.P.-Progressive
crown, Figure 3.3. The bottom flange thickness is
6 in. (0.15 m), except near the river piers where
the thickness increases to 15.75 in. (0.4 m) to accommodate cantilever b e n d i n g s t r e s s e s . T h e
downstream half of the bridge (consisting of two
Choisy-le-Roi Bridge and Other Structures in Greater Paris, France
85
Precast Segmental Bridges
Choisy-le-Roi
1962-64
Courbevoie
65-66
Ring .Motorlva\
66-68
Ring .Motor\vav
6i-68
St Cloud
72-74
Juvis)
66-68
Co nflans
50-72
78
St Maurice Interchange
B-3 South L’iaduct
71-72
Marne la Vallee
7.s77
Torcv RR
78
Clichv RR
78
Cast-in-Place Segmental
Bt-idges
1974-76
13 Gennevilliers
14 North \Vest A-86 Intel-change
78
15 Clichy High\va\zyxwvutsrqponmlkjihgfedcbaZYXW
i 3 -i <i
16 Puteaus Bridges
7.3-77
17 Issv lea Moulineaus
il-54
18 CravelIe
74 -7.5
19 .Joinville
74-76
20 Neuillv sul- Marnc
6 6 - 6 8
FIGURE 3.1.
Location map of’ segmental bridges in greater Paris, France.
box girders) ~\‘as constructed first, alongside the
esisting b rid g e. A f ter rem o v al o f the existing
b rid g e. the sec o nd o r u p stream half w as c o nstructed. Each dual structure was constructed b\
the balanced cantilever method utilizing Frevssinet
tend o ns f’or the lo ng itu d inal p restressing . Bo x
girder segments \vere 8.2 ft (2.5 m) in length and
lveighed 22 tons (20 nit), except the pier segments
FIGURE 3.2. Choisv-lc-Roi
Bridge.
which were 16.4 ft (5 m) in length and weighed
60.6 tons (55 mt). The pier segments also contained two diaphragms which provided continuitv
with the inclined wall piers, Figure 3.3.
The segments were fabricated in a precasting
vard on the left bank of the Seine approximately a
mile (1.6 km) upstream of the project site, Figure
3.4. Although this bridge might be considered of
moderate importance with respect to span lengths,
its importance lies in the method of fabrication. It
was the first to use segments precast by the matchcasting technique. Segments were cast in the precasting yard as a series of 8.2 ft (2.5 m) long units,
one against the other, on a continuous soffit form
which had been carefully adjusted to the intrados
profile of the bridge with allowance for camber.
This came to be known as the “ long-line” method
(see Chapter 11). Two sets of steel forms riding the
soffit form and overnight steam curing allowed the
production of two segments per working day. To
prevent bonding of the segments to each other in
the casting form, a special peel-off bond breaker
w as sprayed over the end of the segment before
the adjacent segment was cast. The segments were
86
Precast Balanced Cantilever Girder Bridges
Elevation
Elevation and cross section of river piers
,
-..I mr-&aL&. ,66
3M
I--1-e-. ,x130
20‘ohp--- A
Cross section of superstructure
FIGURE 3.3. Choisy-le-Roi Bridge, dimensions: elevation, elevation and cross section of River piers, cross
section of superstructure.
subsequently stripped from the soffit form at their
match-castjoints and reassembled at the bridge site
in balanced cantilever on each side of the river
Diers.4
A floating crane handled the segments at the
casting yard. After the units were loaded on barges
and transported to the project site, the same crane
placed the segments over a retractable jig rolling
inside the box girder in the completed portion of
the bridge and was thus freed for another segment
placing operation. A platform mounted on jacks
on the jig, Figure 3.5, allowed for adjustment of
the segment at the desired position.4 A 1 ft (0.3 m)
wide gap was temporarily maintained between the
faces of the segments to allow workmen to apply
Choisy-le-Roi Bridge and Other Structures in Greater Paris, France
FIGURE
P------
3.4.
a7
Choisy-Iv-Koi Kritlge. view of’ the precasting yard.
-J
FIGURE 3.5.
Choisy-le-Roi
the epoxy joint material. The jig was then retracted
and prestressing tendons were placed and stressed
to connect the two symmetrical segments on each
side of the previously completed portion of the
cantilevers on either side of the pier.5
Placing of the precast segments in a cantilever
fashion on each side of the pier progressed step by
step, as indicated in Figure 3.6. Tendon layout is
illustrated in Figure 3.7. Upon completion of the
two twin cantilevers from the river piers, a cast-inplace closure pour was consummated at midspan
and a second series of prestressing tendons were
placed in the bottom flange to achieve continuity
between the two center-span cantilevers. These
tendons were given a draped profile to allow the
location of tendon anchorages in the top flange of
the box girder. Both series of tendons, cantilever
and continuity, overlap each other and contribute
Bridge,
retractable
erection
jig.
FIGURE 3.6. Choisy-le-Rot
with floating crane.
Bridge, segment placing
Precast Balanced Cantiher Girder Bridges
l3cdes1208
) / 8cablesl2#
7
FIGURE 3.7. Choisv-le-Roi Bridge, tendon lavout
to a substantial reduction in the shear forces in the
webs as a result of the vertical component of the
prestress. The side spans were constructed in a
similar manner. The three precast segments adjac ent to the ab u tm ents w ere assem b led o n
f alsew o rk. A f ter a c lo su re p o u r b etw een these
segments and the cantilever from the river pier,
positive-moment tendons were placed and stressed
in the end span to achieve continuity. Because the
midspan area of the center span had little capacity
to withstand moment reversal under ultimate load,
additional short tendons were located in the top
flang e to ac hiev e fu ll reinfo rc em ent c o ntinu ity
with the longest cantilever tend0ns.j
The same construction technique used for the
Choisy-le-Roi Bridge was used for the Courbevoie
Bridge, built between 1965 and 1967, which also
crosses the Seine in the northwest suburb of Paris,
Fig u re 3.1. The b rid g e has three sy m m etric al
spans of 130,200, and 130 ft (40,60, and 40 m) for
a total length of 460 ft (140 m), Figure 3.8. Four
box girders of constant depth carry the 115 ft (35
FIGURE 3.8. Courbevoie Bridge, elevation.
Pierre Benite Bridges near Lyon, France
m) wide deck, Figure 3.9. The available depth of
only 7.5 ft (2.28 m) made necessary a very slender
structure; depth-to-span ratio for the main span is
1/ 26.5,6
Each river pier is an assembly of two half-piers,
Figures 3.9 and 3.10, which are fixed at the level of
the foundation. Each half-pier consists of a rectangular shaft 9 by 26 ft (2.8 by 8 m), which supports
two pairs of prestressed concrete walls, above the
normal water level, in the form of a parallelogram
of 18 in. (0.45 m) thickness and 10.5 ft (3.2 m)
w idth. The w alls are arranged in a “V” in the
transverse direction of the bridge and have a dimension of 6.7 ft (2.05 m) out-to-out of walls in the
longitudinal direction.6 The girders are fixed at the
piers and supported on elastomeric bearings at the
abutments. A total of 148 precast segments of 12.5
ft (3.8 m) length were required for the superstructure. They were fabricated in four months at
the rate of two segments per day, in two sets of
steel forms, electrically heated and insulated with
polyurethane 1ining.j
Erection at the site was accomplished by a floating crane. After careful adjustment of the pier
segments, they were erected at the rate of four per
day. The temporary jig used at Choisy-le-Roi for
adjustment of the segments was replaced in this
project by two temporary steel beams bolted to the
top of each segment and connected to the completed section of the cantilever by prestressing
bars.j
The girder was prestressed longitudinally and
transversely, through three longitudinal cast-inplace strips between the top flange cantilevers of
the box girders. The completed structure is shown
in Figure 3.10.
3.3
89
FIGURE 3.9. Co urbev o ie Brid ge, cro ss sectio n at rive1
pier and abutment.
FIGURE
3.10.
Courbevoie Bridge, view of completed
brid ge.
Piewe Benite Bridges Near Lyon, France
These two large bridges carry the motorway from
Paris to the Riviera south of Lyon near the Pierre
Benite hydroelectric plant, Figure 3.11. There are
two separate bridges, one over the draft channel of
the power plant and the other over the Rhone
River. Both structures are twin bridges, each
bridge consisting of two single-cell box girders.
Typical dimensions in longitudinal and cross sections are show n in Figures 3.12 and 3.13. The same
constant depth of 11.8 ft (3.6 m) is used for all
spans of the two bridges. However, a haunch
under the intrados of the box girders increases the
FIGURE
3.11.
finished bridge.
Pierre Benite Bridge, view
of the
Precast Balanced Cantilever Girder Bridges
90
1
5600
I
I9400
I
I
84@J
!
* 7wotslr Ibxo
56m -4
I
I
m
0
0
4
ISJOO
Bridge over draft channel
(a)
259,OO
/
zyxwvuts
7500
FIGURE 3.12. Pierre Benite Bridge, longitudinal sections. (a) Bridge over draft channel. (b) Bridge over Rhone River.
16.92
- ..3.26-- (
i
t
13.00
FIGURE 3.13.
..?.O
16
30
166
Pierre Benite Bridge, typical cross section.
structural depth over the piers to a maximum of 14
ft (4.28 m) for the 276 ft (84 m) span. All piers rest
on compressed-air caissons and are made of solid
cylindrical columns 6.5 ft (2 m) in diameter which
support the cast-in-place pier segment, including
skew diaphragms between the two individual box
girders of each bridge. This pier segment served as
the starting base for precast segment placing in
balanced cantilever for the superstructure.
The 528 segments were precast near the southern
bank of the draft channel. This application of precast segmental construction was the occasion to
conceive and develop for the first time the short-
line precasting method, whereby the segments are
cast in a formwork located in a stationary position.
Each segment is cast between a fixed bulkhead and
the preceding segment, in order to obtain a perfect
match. After a learning curve of a few weeks, each
of the two short-line-method casting machines was
used to cast one segment every day. Details and
specific problems of the short-line method will be
described in Chapter 11. Figure 3.14 shows the
precast segments as they were fabricated, temporarily stored, loaded on barges by a very simple
portal structure equipped with winches, and finally
transported to the construction site.
Other Precast Segmental Bridges in Paris
91
construction site with segment placing in progress
is shown in Figure 3.16.
Both precasting and placing operations were
carried out successfully. All the segments were
placed in the structures in 13 months. The only regret was that this erection system did not provide
for precast pier segments. The geometry of the
cast-in-place pier segments was further complicated by the skew of the bridges, such that the
contractor expended as much labor on this aspect
of construction as in precasting and positioning all
the precast segments..
3.4
Other Precast Segmental Bridges in Paris
The first two match-cast bridges, Choisy-le-Roi and
Courbevoie, were followed by a series of other
crossings over the Seine River. All contracts for design and construction were obtained on a competitive basis with other types of materials or construction methods:
The next two structures were for the construction of the Paris Belt Motorway which crosses the
Seine at two locations, one downstream of the city
and one upstream; see the location map, Figure
3.1. They were followed by several others, which
are briefly described in this section.
3.4.1 PARIS BELT (DOW NSTREAM )
These twin bridges, Figure 3.17, carry four traffic
FIGURE 3.14. Pierre Henitc Bridge, precasting yard
and loading portal. (a) Precasting yard. (6) Loading portal. lanes. Dimensions are shown in Figures 3.18 and
Placing of all segments in the two twin structures
was achieved in balanced cantilever, using the
cast-in-place pier segments as a starting base. This
project used the newly developed “beam-andwinch” erection system, illustrated in Figure 3.15
together with a close-up view of a typical segment-placing operation. Electric winches are supported in a cantilever position from the completed part of the deck to allow each segment to be
lifted off the barge and placed in its final position.
Because of high-velocity river currents on one
structure, it was considered advisable to transfer
the segments from the barge to the winch system
close to the piers to allow temporary anchorage of
the barge. Therefore, segments had to be moved
longitudinally from the barge position to thtir final
location. A special trolley carried the winches and
the suspended segment while riding along rails
fixed to the finished deck. A general view of the
3.19. Maximum span length is 302 ft (92 m) and
the structural depth of the four box girders is 11 ft
(3.4 m), increased toward the piers to a maximum
of 21.3 ft (5.5 m) by straight haunches. Because of
the skew between the axis of the bridge and the
flow of the Seine, the pier shafts were given a special lozenge shape, which proved very efficient for
the hydraulic flow and is of pleasant appearance.
The limited bending capacity of the shafts called
for temporary supports during cantilever construction operations.
Precast segments were manufactured on the
bank of the Seine with two casting machines
(short-line method). For the part of the bridge
superstructure located over the river, segments
were placed with a floating crane, Figure 3.20. In
fact, almost half the bridge length was placed over
land out of reach of the floating crane. The beamand-winch equipment used at Pierre Benite Bridge
was substituted for the crane to place these segments. There was also need of additional falsework
on one bank to compensate for the unusually long
FIGURE 3.16.
Pierre Benite Bridge, under-
FIGURE 3.15. Pierre Benite Bridge, segment placing
FIGURE 3.17. Paris Belt (Downatrearn),
scheme (left and top right).
finished bridge.
92
construction.
\ itw o f
9950
FIGURE 3.18.
-------rng.3
._._ ~i6_~-~~j~~--rlps--~
___--.-
Paris Belt (Downstream), typical longitudinal section.
FIGURE 3.19. Paris Belt (Downstream), typical cross section.
:
i-
zyxw
zyxw
Precast Balanced Cantilever Girder Bridges
94
FIGURE 3.20. Paris Belt (Downstream), segment
p lacing .
FIGURE 3.21. Pal i\ Belt (C‘pstlum), \iew o f the
finished bridge.
end span, which could not be changed because of
stringent pier location requirements.
direction, Figure 3.21. The twin bridges have dimensions similar to those of the downstream
bridge, and each structure has two parallel box
girders connected by transverse prestress. Dimensions are show n in Figures 3.22 and 3.23. A circular intrados profile was used in lieu of the straight
haunches. All segments were precast on the river
bank in the immediate vicinity of the bridge, using
3.4.2 PARIS BELT (UPSTREAM )
On the other sihe of Paris another segmental
structure, also carrying the Belt Motorway over the
Seine, was designed for five traffic lanes in either
asa
56,62
I
Id GAUCHE
-.
w
FIGURE 3.22. Paris Belt (Upstream), longitudinal section.
4.50
4. 0
E
II 18zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
1
3.50 m
1
3.50 n-l
3.50 m
II
3.50 m
3.50
1
71
m
g
i Ill I
3.50 m
3.50 m
I
3.50 m
IlII
FIGURE
3.23.
3.50 m
3.50 m
-
Paris Belt (Upstream), typical cross section.
A
/\
Other Precast Segmental Bridges in Paris
S?QU l?4 CCS~~I SCK c e M Ruc na r
RUSES D’EMCUTIDN D” T -
FIGURE 3.24.
Paris Belt (Upstream), typical segment placing scheme.
the same two casting machines used previously for
the downstream bridge.
Placing segments in the structure posed some
interesting problems, as shown in the sequence
diagrams of Figure 3.24. Pier segments were too
heavy to be handled as one unit and were subdivided into two segments, assembled upon the
pier shaft before cantilever placing could start. A
crane, either on crawlers or on a barge, together
with the beam-and-winch equipment handled all
segment placing.
3.4.3 JUVISU
This bridge, Figure 3.25, is also on the Seine just
south of Choisy-le-Roi; see the location map, Figure 3.1. Dimensions are shown in Figure 3.26.
Segments were cast by the short-line method near
the site and placed w ith a floating crane. An auxiliary falsework on both banks allowed segment
placing and assembly beyond the reach of the
floating crane.
3.4.4
FIGURE 3.25. JUVISV Bridge, completed stl ucture.
BRIDGE
TWIN BRIDGES AT CONFLANS
These twin bridges, Figure 3.27, placed about 320
ft (100 m) apart to allow for interchange ramps on
both banks, are upstream of Paris where the Seine
and Marne Rivers merge; see the location map,
Figure 3.1. Dimensions and construction methods
were similar to those of the Courbevoie Bridge already described.
Precast Balanced Cantilever Girder Bridges
96
I
24:3@
1
2413’
IO’
Cl
1
I33
Ed
.-
z
FIGURE 3.26. Juvisy Bridge, cro ss sectio n.
FIGURE 3.27. Twin Bridges at Conflans, finished
bridge.
Balanced cantilever construction was accomplished utilizing a launching gantry for erection.
In the approach spans the superstructure has a
constant depth of 8.2 ft (2.5 m). Depth of the
center spans varies from 14.9 ft (4.5 m) at the piers
to 8.2 ft (2.5 m) at midspan, Figure 3.29. The rectangular box segment has a bottom flange width
of 18 ft (5.5 m) and a top flange width of 34.8 ft
(10.6 m). Webs have a constant thickness of 12 in.
(0.3 m), while the top and bottom flanges are 8 in.
(0.2 m) and 7 in. (0.18 m) thick, respectively, Figure 3.30. Typical segment length is 10.8 ft (3.3 m).
Expansion of the deck is provided in every
fourth span by a special stepped (ship-lap) joint
with horizontal elastomeric bearing pads, Figure
3.5 Oleron Viaduct, France
The Oleron Viaduct provides a link between the
mainland of France and the resort island of Oleron
off the Atlantic West Coast 80 miles (128 km) north
of Bordeaux. This structure has a total length between abutments of 9390 ft (2862 m). In the navigable central part of the structure are 26 spans of
260 ft (79 m), Figure 3.28. Approach spans consist
of two at 194 ft (59 m), sixteen at 130 ft (39.5 m),
and two at 94 ft (29 m). The superstructure is supported by 45 piers and was assembled by prestressing match-cast segments, using epoxy joints.
FIGURE 3.28. Olevon Viaduct, complcred
strllcrllre.
97
Oh-on Viaduct, France
3sllo’
3!2’
I
j
I
,
I
2916’
3!2’
I
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i
t
/
18'
34!9”
c
I
I\
Oleron Viaduct, typical cross section, from ref. 5 (courtesy of
the American C:oncrete Institute).
FIGURE 3.29.
3.30. Throughout the total length of structure
there are ten expansion joints: one at each abutment and eight intermediate ones. The latter are
located at points of contraflexure in a typical
interior span subjected to a continuous uniform
load.” The segments with the expansion joint have
the same length as typical segments and are in fact
two half-segments that are temporarily preassembled with bolts, with a special layout of temporary
and permanent prestressing tendons. It is then
possible to maintain the balanced cantilever erection procedure beyond the expansion joint to
midspan.
Later on, when continuity has been
achieved in the adjacent spans, the expansion.joint segment is ‘!unlocked”
to perform in the intended manner.
FIGURE 3.30.
The precasting plant was located in the vicinity
of the mainland abutment. Production in this plant
was scheduled so that the 24 segments required for
a typical 260 ft (79 m) central span could be fabricated in nine working days. Segments were produced by the long-line method, described in
Chapter 11. Four sets of steel forms rode a bench
that was carefully aligned to the longitudinal
profile of the roadway and the variable-depth soffit
with due provision for camber. Segments were
match-cast in the same relative order in which they
were subsequently assembled at the site.5 An aerial
view of the casting yard is shown in Figure 3.31.
Handling of segments in the casting and storage
yard w as accomplished by a special railwaymounted gantry capable of handling loads varying
Oleron Viaduct, typical center span elevation, from ref. 5 (courtesy of the American Concrete Institute).
Precast Balanced Cantilever Girder Bridges
3.32. Oleron Viaduct, construction view
showing cantilever span, from ref. 5 (courtesy of the
American Concrete Institute).
FIGURE
FIGURE
3.31.
Oleron Viaduct, aerial view of casting
yard.
from 45 tons (42 mt) for the center-span segment
to 80 tons (73 mt) for the pier segment. A lowboy
dolly riding on rails of the finished bridge and
pushed by a farm tractor transported the segments
from storage to their location for assembly.
Cantilever erection at the site was accomplished
by a launching gantry, Figure 3.32. This gantry
was the key to the successful operation of this project. Although the structure is erected over water,
the use of floating equipment would have been
difficult, expensive, and subject to uncertainty because of the great tidal range and the shallowness
of water in most of the area traversed by the
structure. Floating equipment would have been
able to reach the approach piers only at high tide.
During low tide the marsh area, which is the location of France’s famed Marennes oyster beds,
could not accept any tire-mounted or crawlermounted equipment. Consequently, it was decided
to work entirely from above with a launching
gantry. This new technique was developed for the
first time for this structure and was later refined
for other structures. For the typical central spans
the erection cycle required between eight and ten
working days.5
Construction began in May 1964, three months
after design work had started. The first segment
was cast in July and placed in August 1964. Side
spans laid on a curve were completed in December
and the launching gantry was then modified for
construction of the center spans. The last of the
870 precast segments was in place in March 1966,
and the bridge opened to traffic in May, after an
overall construction time of two years5; see the
summary of the work program in Figure 3.33. A
view of the final structure is shown in Figures 3.28
and 3.34.
The Oleron Viaduct was the first application of
the launching-gantry concept for placing segments
in cantilever. Several structures were later designed and built with the same construction
method. Mention should be made here of three
special bridges:
1. Blois Bridge over the Loire River The principal dimensions are given in Figure 3.35. The
superstructure box girders rest on the pier shafts
through twin elastomeric bearings, which allow
thermal expansion while providing partial restraint for bending-moment transfer between deck
and piers. Consequently, savings are obtained both
in the deck and in the foundations. All segments
were placed in the bridge with an improved version of the launching gantry first designed for the
Oleron Viaduct. High-strength steel and stays
were used to provide minimum weight with a satisfactory stiffness during operations, Figure
3.36. High-strength bolt connections were used
throughout to make the gantry completely capable
of dismantling and easily transportable to other
construction
sites.
2. Aramon Bridge over the Rhone River This was
the next structure where the same gantry could be
used, Figure 3.37.
3. Seudre Viaduct Located just a few miles
south of Oleron over the Seudre River, this 3300 ft
(1000 m) long viaduct was also of precast segmental construction and used the same launching gan-
Chillon Viaduct, Switxerland
99
CONTINENT
OLERON
i
PIERS
ON FOOTINGS
1 1
1 1
FIGURE
PIERS ON PILES
-------$
3.33.
PIERS ON FOOTINGS
i
Oleron Viaduct, program of work.
try. The finished structure is shown in Figure 3.38.
Foundations for the center spans were built inside
sheet pile cofferdams in spite of very swift tidal
currents.
3.6 Chillon
)-
Twin rectangular slip-formed shafts were used
for the piers, varying in height from 10 to 150 ft (3
to 45 m). Stability during construction was excellent and required little temporary bracing except
between the slender walls to prevent elastic instability.’ With the exception of three piers in each
Viaduct, Switzerland
The 7251 ft (2210 m) long dual structures of the
Chillon Viaduct are part of European Highway E-2
and are located at the eastern end of Lake Geneva
passing through an environmentally sensitive area
and very close to the famed Castle of Chillon, Figure 3.39. In addition, the structures have very
difficult geometrical constraints consisting of 3%
grades, 6% superelevation, and tight-radius curves
as low as 2500 ft (760 m). Each structure has 23
spans of 302 ft (92 m), 322 ft (98 m), or 341 ft (104
m). The variable spans allowed the viaduct to be
fitted to the geology and topography, providing
minimum impact on the scenic forest. The viaducts
are divided by expansion joints into five sections of
an approximate length of 1500 ft (457 m).
FIGURE
brid ge.
3.34.
Oleron Viaduct, aerial view of’ finished
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61.50
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Precast Balanced Cantilever Girder Bridges
100
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9.1 m
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3.35.
Blois Bridge, elevation and typical cross section.
viaduct, all piers are hinged at the top. The piers
that are less than 72 ft (22 m) high are hinged at
the base; taller piers are fixed at their base, being
sufficiently flexible to absorb longitudinal move-
FIGURE 3.36. Blois
Bridge,
operating on the superstructure.
$00
I
to 4.79 m at midspan
FIGURE
I
zoo
launching
ment of the superstructure.
The superstructure consists of a single-cell rectangular box with a cellular cantilever top flange,
Figure 3.40, and with a depth varying from 18.5 ft
gantry
FIGURE 3.37. Aramon Bridge, launching gantry.
Chillon
FIGURE
3.38.
Seudre Bridge, fItli\hcd \I
I
Viaduct, Switzerland
101
zyxwvut
FIGURE 3.39. Chlllon Viaducl.
I C 1111 e.
(5.64 m) at the longer-span piers to 7.2 ft (2.2 m) at
midspan. Widths of top and bottom flange are respectively 42.7 ft (13 m) and 16.4 ft (5 m). Dimensions of the tw& typical cantilevers are noted in
Figure 3.4 1. Maximum segment weight was 88 tons
(80 nit). A cellular cantilever top flange was used
because the overall width of the top flange ex-
aerial LICI\.
ceeded 40 ft (approx. 12 m) and the cantilever
length w as 13.15 ft (4 m). An alternative would
have been to provide stiffening ribs as used in the
Saint Andre de Cubzac Viaducts (Section 3.11) and
the Sallingsund
Bridge (Section 3.13).
Segments were precast in a yard at one end of
the structure with five casting machines, allowing
O ve r sup p o rts
(4
b
4
500
At m id- spa n
(b)
FIGURE
3.40.
Chillon Viaduct, cross sections. (a) Over supports. (b) At midspan.
i
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PORTIQUE - TYPE 48.OOm
PORTIOUE-TYPE
4 2 OOm
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FIGURE
3.41.
Chillon Viaduct, longitudinal sections of typical cantilevers.
Hartel Bridge, Holland
103
Sections I, II, and V, conventional cast-in-place
prestressed concrete box girders
Sections III and IV, precast prestressed concrete
segmental box girders
Two steel bascule bridges.
FIGURE 3.42. Ch~llon
\.~ndu~t, precasting yard.
an average production of 22 to 24 segments per
week (see aerial view, Figure 3.42).
Erection was by the conventional balanced cantilever method with a launching gantry designed to
accommodate the bridge-deck geometry in terms
of curve and variable superelevation. The overall
length of the gantry was 400 ft (122 m) and the
total weight 250 tons (230 mt). Special features of
this gantry will be discussed in Chapter Il. Cantilever placing of precast segments is shown in Figure 3.43.
This structure is truly an achievement of modern technology with emphasis upon the aesthetic
and ecological aspects of design.
3.7 Hartel Bridge, Holland
The 1917 ft (584.5 m) long Hartel Bridge crosses a
canal in Rotterdam, Figure 3.44, and consists of
the following elements:
The original design contemplated that the total
structure would be constructed as conventional
cast-in-place box girders on falsework. Substitution
at the contractor’s request of cast-in-place segmental construction by precast skgmental construction for sections III and IV saved the extensive temporary pile foundation system necessary to
avoid uneven settlement of falsework because of
initial soil conditions. The redesign proposed two
single-cell rectangular box girders as opposed to
one three-cell box girder, Figure 3.44, omitting the
center portion of the bottom flange and providing
thinner webs and a thicker bottom flange.
In the segmental box girder design the dimensions of the deck slab are constant over the entire
length, girder depth varies from 4.92 ft (1.5 m) to
17 ft (5.18 m), the webs have a constant thickness
of 13.8 in. (0.35 m), and the bottom flange thickness
varies from 10 in. (0.26 m) to 33 in. (0.85 m). Up to
a depth of 9.35 ft (2.85 m) the segments have a
length of 15.8 ft (4.8 m); over 9.3 ft (2.85 m) the
length decreases to 12.3 ft (3.75 m).
The vertical curvature of the bridge was made
constant for the full length of sections III and IV
by increasing the radius from 9842.5 ft (3000 m) to
19,029 ft (5800 m), which resulted in a repetition
of eight times half the center span. This repetition
justified precast segments.
A long-line casting bed (see Chapter 11) was constructed on the centerline of the bridge box girders
at ground level, Figure 3.45. Thus, a portal crane
was able to transport the cast segments to the storage area and also erect them in the superstructure,
Figure 3.46. The end spans have three more segments than half the center span; these were supported on temporary falsework until all the prestressing tendons were placed and stressed, Figure
3.46.
The first segment cast was the pier segment;
each of the remaining segments was then matchcast against the preceding segment. The pier segment was positioned on bearings on top of the pier,
Figure 3.47, and the two adjoining segments were
positioned (one after the other) and the joints
glued with epoxy resin. Temporary high-tensile
bars located on the top of the deck slab and in the
bottom flange were stressed to prestress the three
zyxwvutsr
FIGURE 3.43. Chillon Viatiuct,
tion with launching gantry.
c a n tile v e r c o nstrUC -
Precast Balanced Cantilever Girder Bridges
III
IV
Flevation
Cross sections of the redesign
Cross section of the
original design
3.44. Hartel Bridge, typical dimensions: elevation, cross sections of the original design, cross sections of the redesign (courtesy of Brice Bender, BVNLSTS).
FIGURE
segments together. After the epoxy had hardened,
the permanent tendons were placed and stressed.
The two segments adjoining the pier segment were
supported during erection on flat jacks on the top
FIGU?E 3.43.
of the outside struts of a steel scaffolding bearing on
the pier foundation. Thus, the flat jacks were used
for adjustment of the segments to achieve proper
geometry control. The remaining segments were
Hare1 BridgG nxrbod of castjng segments /courtesy of Brjce
Bender, BVNISTS.
Hartel Bridge, Holland
FIGURE
FIGURE 3.46. H,~~tcl H~~tigc, p t,tl (I<II~C
dling segments.
105
3.48.
H,II tel RI idgc. complctcd \tl II< ture.
for h‘in-
erected in the conventional balanced cantilever
method. The completed structure is shown in Figure 3.48.
Other structures using precast segmental construction were subsequently designed and built in
the Netherlands. Shown in Figure 3.49 is the
bridge over the I.jssel at Deventer, where segments
in the 247 f’t (74 m) spans w ere placed w ith a
launching gantry. The overall length of’ the gantry
w as 520 f‘t (156 m), allowing the legs to bear on the
permanent concrete piers and impose no loading
on the deck during construction, Figure 3.50.
FIGURE 3.49. Ikventex Bridge, placing segments
with the launching gantry.
FIGURE 3.47. Hartel Bridge, erection sequence and detail of temporary pier bracing (courtesy of Brice Bender, BVN/STS).
Precast Balanced Cantilever Girder Bridges
106
156 m (520 ft)
I-
rl
74 m (247 ft)
7 8 m (260 ft)
Real
Front
Max bridge span 74 m (247 ft)
FIGURE 3.50. Deventer Bridge, elevation of gantry.
3.8 Rio-Niteroi Bridge, Brazil
The Rio-Niteroi Bridge crosses the Guanabara Bay
connecting the cities of Rio de Janeiro and Niteroi,
thereby avoiding a detour of 37 miles (60 km). This
structure also closes the gap in the new 2485 mile
(4000 km) highway that interconnects north and
south Brazil and links the towns and cities on the
eastern seaboard, Figure 3.51. Although the route
taken by the bridge across the Bay seems somewhat
indirect, it was selected because it avoids very deep
water and is clear of the flight path from Santos
Dumont Airport.
Total project length is approximately 10.5 miles
(17 km), of which about 5.65 miles (9.1 km) is over
water. The alignment begins at the Rio side with a
3940 ft (1200 m) radius curve, then a straight section, within which are located steel box girder
navigation spans totaling 2872 ft (848 m) in length.
This is followed by an island, where the viaduct is
interrupted by a road section of 604 ft (184 m), and
finally another 3940 ft (1200 m) radius curve arriving at Niteroi.
The precast segmental concrete viaduct sections
have a total length of 27,034 ft (8240 m) representing a total deck area of 2,260,OOO sq ft (210,000
The
- Rio-Niteroi
Brii
/n
FIGURE
3.51.
_.RmdeJaneir
Rio Niteroi Bridge, site location map
Rio-Niteroi Bridge, Brazil
m*), making this bridge the largest structure of its
type. An aerial view of the crossing under traffic is
show n in Figure 3.52. The superstructure has 262
ft (80 m) continuous spans w ith an expansion joint
at every sixth span, Figure 3.53. It consists of two
rectangular box girders for a total width of 86.6 ft
(26.4 m) and a constant depth of 15.4 ft (4.7 m). A
2 ft (0.6 m) cast-in-place longitudinal closure joint
107
between the top flange cantilevers provides continuity between the two box girder segments. Typical segments have a length of 15.75 ft (4.8 m) and
weigh up to 120 tons (110 mt). The pier segments
are 9.2 ft (2.8 m) in length. Special segments are
used for expansion joints.
Longitudinal prestressing tendons consist of
twelve f in. (13 mm) diameter strands in the top
and bottom flanges with a straight profile, while
the resistance to shear stresses is obtained by vertical web prestress, Figure 3.54.
All segments were manufactured in a large precasting yard on a nearby island. Ten casting
machines (eight for the typical segments and two
for the pier and hinge segments) were laid in two
independent parallel lines, each equipped w ith a
portal crane for carrying the segments to the storage area and the loading dock. More than 3000
segments were subsequently barged to their location in the structure and erected by four launching
gantries working simultaneously on each of the two
parallel box girders and on either side of the bay,
Figures 3.55 and 3.56. The rate of segment placing
was remarkable. A typical span was assembled and
completed in five working days. Between the
months of February and July 1973, an average of
FIGURE 3.52. Rio-Sire] oi 131 dge, view of the completed structure.
Cross
section
Elevation
fb)
FIGURE 3.53. Rio-Niteroi Bridge, cross section and elevation. (a) Cross section. (b)
Elevation.
Precast Balanced Cantilever Girder Bridges
ELEVATION
PLAN
CABLAGE
SUPERIEUR
*
PLAN
FIGURE
3.54.
CABLAGE
INFERIEUR
Rio-Niteroi Bridge, typical span dimensions and tendon layout.
278 precast segments per month were installed in
the structure by the four launching gantries, representing an area of 180,000 sq ft (17,000 m’) of
finished bridge per month. At the same speed,
Oleron Viaduct could have been built in two
months. Such is the measure of the determination
and enthusiasm of engineers and constructors of
the New World.
3.9 Bear River Bridge, Canada
The Bear River Bridge is about 6 miles (9.7 km)
east of Digby, Nova Scotia, on trunk route 101
between Halifax and Yarmouth, near the Annapolis Basin; it replaces an 85-year-old structure.
Preliminary studies showed, and construction bid
prices verified, that precast segmental was more
economical than steel construction by nearly 7%.7*8
109
JFK Memorial Causeway, U.S.A.
F I G U R E 3 . 5 5 . RIO-NIICI
strut tion.
oi 131 dgc, wnrile\el
COII-
Total structure length is 1998 ft (609 m) with six
interior spans of 265 ft (SO.8 m) and end spans of
204 ft (62.1 m), Figure 3.57. The layout has very
severe geometry constraints. In plan, the east end
of the bridge has two sharp horizontal curves connected to each other and to the west end tangent by
two spiral curves; minimum radius is 1150 ft (350
m). In elevation, the bridge has a 2044 ft (623 m)
vertical curve with tangents of 5.5 and 6.0 percent.
Two sets of short-line forms employed by the contractor to cast the segments met the variable
geometry requirements admirably. The accuracy
of casting was such that only nominal elevation
adjustments were required at the abutments and
the center-span closure pours.s
The single-cell box girder superstructure is continuous for the total length of the bridge. Typical
cross-section dimensions are indicated in Figure
3.58. Prestressing tendon layout is illustrated in
Figure 3.59 for a typical interior span. Fifty-five
tendons were required for negative moments and
22 for positive-moments. The majority of nega-
FIGURE 3.56. Rio-Niteroi
tries.
RI
dgr.
launching gan-
tive-moment tendons were inclined in the web
and anchored at the face of the segments. Anchorage of six tendons at the face of the first segment
adjacent to the pier segment (three in each web)
produced a large upward shear force at the face of
the pier segment, which was not overcome until
the erection of several additional segments. The
midspan positive-moment tendons are continuous
through the cast-in-place closure joint at midspan.
These tendons, indicated by capital letters in Figure 3.59, were placed in preformed ducts upon
completion of erection of the segments in a span
and the closure pour consummated. All positivemoment tendons were anchored in the top flange.
The precast segments are typically 14 ft 2 in. (4.3
m) in length and the closure pour at midspan is 4
ft 4 in. (1.3 m) long.7,R
The precast segments are reinforced with prefabricated mild steel reinforcement cages, in addition to the primary longitudinal prestressing tendons, Figure 3.60, and transverse prestressing in
the top flange. Web shear reinforcement varies
depending on the location of the segment. The 145
precast segments were cast in a plant located near
the bridge. This plant was equipped with two casting molds, each producing one segment per day. A
12-hour steam curing period w as used and a concrete strength at 28 days of 5000 psi (34.5 MPa)
w as achieved .’
Because of the curved layout of the bridge and
its relative shortness, the use of a launching gantry
would have been uneconomical. Segments were
placed by a 200 ton (180 mt) mobile crane on land,
or on a barge over water, Figure 3.61. Construction of this bridge started in May of 1971, and it
was opened to traffic on December 18, 1972.
zyxwv
3.10 JFK Memorial Causeway, U.S.A.
A portion of the JFK Memorial Causeway represents the first precast, prestressed, segmental box
girder completed in the United States. Opened to
traffic in 1973, this 3280 ft (1000 m) long structure
spans the Gulf Intercoastal Waterway in Texas to
connect Corpus Christi and Padre Island. It was
designed by the Bridge Division of the Texas
Highway Department under the supervision of
Wayne Henneberger. The Center for Highway
Research, University of Texas at Austin, under the
supervision of Prof. John E. Breen, assisted in the
design and also built and tested a one-sixth scale
model of the bridge to check design requirements
and construction techniques.g
E LIRGS.
E PIER I
E PIER 2
203’.9”
265’.0
I
E PIER 4
E PIER 3
2 6 5 .- 0 ‘,
E PIER 4
”
265*-O”
I
E PIER 5
t P I E R 6
265’-0”
E
265,-O”
PIER 7
203’.9”
U N I T S N O . 7. 2 7 . 4 7 . 6 7 . 67,107.
1 2 7 AND 147
A R E C A S T IN P L A C E (DECK
CLOSlNG
UNITS)
ELEVATION
FIGURE 3.57. Bear River Bridge, elevation, f’ron~ ref. 8 (courtesy of the Prestressed
Concrctc Institute).
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I--- %
39’-6”
2-6’!
ROADWAY
6’4
I5’-0’
I’-()”
yI_ 4”
Ia’-0”
I
-
I
FIGURE 3.58. Bear River Bridge, typical cross section, from ref. 8 (Courtesy
of the Prestressed Concrete Institute).
HALF INTERIOR SPAN TENDON ELEVATION
HALF SECTION AT
MIDSPAN
TENDON
HALF SECTtON AT PIER
DISTRIBUTION
FIGURE 3.59. Bear River Bridge, typical center-span tendon elevation and
distribution, from ref. 8 (courtesy of the Prestressed Concrete Institute).
111
112
Precast Balanced Cantilever Girder Bridges
’
P
FIGURE
3.62. JFK Memorial Causeway, balanced
cantilever construction (courtesy of J. E. Breen).
FIGURE 3.60. Bear Kiver
Bridge, longitudinal prestress ducts in forms (courtesy of the Prestressed Concrete Institute).
FIGURE 3.61. Be,u
Ki\ cl
Bi idge,
crcc tion b v
barge-mounted crane (courtesy of the Prestressed Concrete Institute).
The structure consists of thirty-six 80 ft (24.4 m)
long approach spans of precast, prestressed bridge
beams and the 400 ft (122 m) total length segmental bridge spanning the Intercoastal Waterway.
The segmental portion of this structure has a
center span of 200 ft (61 m) w ith end spans of 100
ft (30.5 m). The segments were precast, transported to the site, and erected by the balanced
cantilever method of construction using epoxy
joints, Figure 3.62. The precast, segmental superstructure consists of constant-depth twin box
girders with a 2 ft (0.61 m) cast-in-place longitu-
dinal closure strip, Figure 3.63. Segments are
10 ft (3.05 m) in length and in cross section, are 8 ft
(2.44 m) in depth, and have a nominal top flange
width of 28 ft (8.53 m). The top flange or deck is of
constant dimension longitudinally but of variable
thickness in a transverse direction. The bottom
flange is of constant dimension transversely but
varies longitudinally from 10 in. (254 mm) at the
pier to 6 in. (152 mm) at 25 ft (7.62 m) from the
pier center.
Segments were cast with male and f-emale alignment keys in both the top and bottom flanges as
well as large shear keys in the webs, Figure 3.64.
Integral diaphragms were cast with the pier segments, Figure 3.65. Both matching faces of the
segments were coated with epoxy, and temporary
erection stressing at both top and bottom of the
segments precompressed the joint before installation of the permanent post-tensioning tendons.
The segments were erected by a barge-mounted
crane. As each segment was erected, it was tilted 21
degrees from the in-place segment, so that a pair of
hooks in the top of the segment being erected engaged pins in the segment previously erected. The
new segment was then pivoted down by the sling
until its shear key slipped into the mating shear key
of the previously erected segment.g
Figure 3.66
shows a permanent tendon being tensioned and
the temporary working platform.
The design concept on this project utilized prestressing tendons in the top flange for dead-load
cantilever stresses; after closure at midspan, continuity tendons were installed for the positive moment, Figure 3.67. Research on the model testing
of the bridge is documented in references 10
through 15 with particular emphasis in reference
14 on lessons learned during construction that
might facilitate or improve similar projects.
Saint And& de Cubzac
Sym. B Q
28 ft. (8.53 m)
L
-m
6’-8” (2.03 m)
6 ft. (1.83 ml
2
z
l---l
113
Bridges, France
8
h .?
7 ft. (2.13 m)
al
:- s
T-10” (2.39 m)
13 ft. (3.96 m)
/-
I-
FIGURE 3.63. JFK Memorial Causeway, typical cross section. Bottom slab
thickness varies from 10 in. (254 mm) at pier to 6 in. (152 mm) at 25 ft (7.62
m) from pier center.
FIGURE 3.66. JFK Memorial C;IUSC\V;I~,
prestressing
permanent tendon (courtesy of J. E. Breen).
FIGURE 3.64. JFK Memorial Causeway, precast seg-
ment in casting yard (courtesy of J. E. Breen).
FIGURE 3.65. J FK Xlemorial Causeway, construction
view showing pier segments with diaphragms (courtesy
of J. E. Breen).
3.11 Saint And& de Cubzac Bridges, France
Opened to traffic in December 1974 after a construction period of 29 months, this important
structure crosses the Dordogne River north of
Bordeaux on the South Atlantic Coast. A view of
the finished bridge is shown in Figure 3.68. The
main river crossing has a total length of 3800 ft
(1162 m) with approach land spans of 190 ft (59 m)
and main river spans of 312 ft (95.3 m), Figure 3.69.
Two intermediate expansion joints located at the
point of contraflexure in the transition spans separate the deck into three sections for concrete volume changes. The center section has a length of
1920 ft (585 m). The main piers have rectangular
hollow box shafts supported by circular opendredged caissons 30 ft (9 m) in diameter. Approach piers have an I section.
Another structure, constructid under the same
contract, consisted of twin bridges 1000 ft (307 m)
in length with typical 162 ft (49.5 m) spans in an
114
Precast Balanced Cantilever Girder Bridges
Cantilever (negative moment) tendons
8 Main pier
C$ Central span
100 ft (30.5 mJ
-I
FIGURE 3.67. JFK Memorial Causeway, system of prestressing tendons.
FIGURE 3.68.
Saint Andre de Cubzac Bridge, view of
the finished bridge over the Dordogne River.
area north of the main crossing where poor soil
conditions did not permit stability of an embankment. Altogether the deck area is 97,000 sq ft
(29,500 m2), entirely of precast segmental construction. The typical cross section is a single box
54.4 ft (16.6 m) wide with transverse ribs both in
the side cantilevers and between webs, Figure 3.69,
to provide structural capacity to the deck slab
under traffic loads. A casting yard located along
the bank of the Dordogne River produced the 456
segments for both bridges (main crossing and
north viaducts) in three casting machines (two for
the typical segments and one for the special segments such as pier, hinge, or end segments). Moderate steam curing at 86°F (30°C) for 12 hours in a
movable kiln enclosing the newly cast segment and
its match-cast counterpart allowed a one-day cycle
and proved very efficient in avoiding any geometric corrections.
Segments were placed in the structure by the
beam-and-winch method either on land (for the
northern viaducts or the approach spans of the
main river crossing) as shown in Figure 3.70 or
over water for the main spans as shown in Figure
3.71. This project was the occasion for a further
improvement in the placing scheme by beam and
winch, whereby the pier segments could be precast
and placed with the same type of equipment as
shown in principle in Figure 3.72. A provisonal
tower prestressed against the pier side face allowed
the pier segment to be installed upon the pier cap,
with the beam and winch later used for cantilever
placing. To keep the segment weight to a
maximum of 110 t (100 mt) the pier segment, representing the starting base of each cantilever, had
been divided into two halves placed successively,
Figure 3.73. Figure 3.74 shows the lifting of the
last closure segment.
3.12
Saint Cloud Bridge, France
A connection between the peripheral Paris Ring
Road and the Western Motorway (A- 13) required
the construction of a bridge over the Seine extended by a viaduct along the left bank leading to
the Saint Cloud Tunnel, Figures 3.75 and 3.76.
This structure has two traffic lanes in each direction. It will be duplicated later by a similar adjoining structure when the congested Saint Cloud
Tunnel is duplicated. O r i g i n a l d e s i g n o f t h i s
bridge contemplated a steel structure. However, an
alternative design utilizing precast segments and
+
8'
B O RD E AU X
I
I
(0
a
N
0,”
Al’iDRE
DE
CUBZAC
I)
I
I
8
---Hm
I
6,00
/
FIGURE
3.69.
1 ,
Saint Andrk de Cubzac Bridge, elevation and cross section.
3.71. Saint Andrk de Cubzac Bridge, beamand-winch segment placing over water.
FIGURE
FIGURE 3.70. Saint And& de Cubzac Bridge, beamand-winch segment placing over land.
115
WlNCt
B
01
FIGURE 3.72.
02
116
.
A
03
Saint Andrk de Cubzac Bridge, placing precast pier segments.
FIGURE 3.73. Saint Andre Cubzac Bridge, lifting
second half pier segment.
E
M
Saint Cloud Bridge, France
the balanced cantilever method of construction,
submitted by the contractor, permitted substantial
savings and was accepted by the authorities.
The bridge has a total length of 3618 ft (1103 m)
with a constant-depth superstructure. It includes
two sections: the bridge over the Seine, which is a
1736 ft (529 m) long curved structure; and a 1883
ft (574 m) long viaduct, which follows a straight
layout along the bank of the Seine and then crosses
the Place Clemenceau, on a 2260 ft (690 m) radius
curve, by an access ramp to the Saint Cloud Tunnel. It includes 16 spans divided as follows (refer to
Figure 3.76):
Seine Bridge:
160.8,288.7,333.8,296.0,150.9,and
two 219.5 ft spans (49, 88, 101.75, 90.25, 46, and
two 66.9m)
Common area: 66.4 ft (20.24 m) up to the expansion
joint, and then 153.1 ft (44.66 m), total 219.5 ft
(66.9 m)
V iaduct:
five219.5; 285.4,210.0,and 137.8ftspans
(five 66.9; 87, 64, and 42 m)
Architectural considerations led to the choice of
a 11.8 ft (3.6 m) constant-depth three-cell box
girder with slopingexternalwebs with nooverhangs,
Figure 3.77. Segments are 7.4 ft (2.25 m) in length
with a record width of 67 ft (20.4 m), their average
weight varying from 84 to 143 tons (76 to 130 mt).
Since the superstructure has a constant depth, the
bending capacity is adjusted to the moment dis-
117
tribution by varying the bottom flange thickness,
which decreases from 3 1.5 in. (800 mm) at the river
piers to 7 in. (180 mm) at midspan. To accommodate the curvature of the bridge the segments in
this area are cast, in plan, in a trapezoidal shape. A
4.5% superelevation is obtained by placing the
units over the piers in an inclined position.
Three-dimensional prestressing was used in the
superstructure: the main longitudinal prestress,
transverse prestress in the deck, and a vertical prestress in the webs to accommodate shear. After the
closure joint at midspan was cast, additional longitudinal prestress tendons were installed to provide continuity.
Superstructure segments were precast in a plant
on the right bank of the Seine. Two casting molds
were used for fabrication of the segments. Each
mold had an external formwork and an internal
retractable formwork. The adjacent, previously cast
segment was used as a bulkhead to achieve a
match-cast
joint.
For erection, segments were transported on a
trolley to a cable-stayed launching gantry of unusual size and capacity. It was of high-yield steel
construction, 402 ft (122.5 m) in length and
weighing 250 tons (235 mt), with a maximum load
capacity of 143 tons (130 mt). The constant-depth
gantry truss was supported on central and rear
legs, which were tunnel shaped to allow passage of
the precast segments endwise. At the central support, a 52.5 ft (16 m) high tubular tower topped
,.,. 59
_ls .,’
- --
FIGURE 3.76. Saint Cloud Bridge, plan view.
COUPE TRANSVERSALE
D’UN
VOUSSOIR TYPE
FIGURE 3.77.
Saint Cloud Bridge, longitudinal and typical cross section.
120
Precast Balanced Cantilever Girder Bridges
with a saddle provided a large eccentricity to
the three pairs of cable stays, which improved the
negative-moment capacity at this support location.
At the forward end of the gantry an additional leg
was used as a third support point during launching
and pier segment placing, Figure 3.78. The
launching girder was moved forward on rails
mounted on the completed superstructure, by
sliding on pads placed at the central and rear legs.
The launching girder, in cross section, was triangular in shape. The base of this triangle included
two structural steel I sections, which served as
tracks for the segment transportation trolley. The
diagonal bracing of the launching girder consisted
of tubular steel members. The girder was fabricated in ten sections, approximately 39 ft (12 m)
FIGURE 3.78. Saint Cloud Bridge, segment placing.
PLACING OF PILE UNITS
AVANCEMENT DU
+ORTlQUE D E
LANCEMENT.
MOVING
THE
MISE E N
TRUSS
PLACE
DES
P L A C I N G T H E U N I T S I N CANTELIVER
FIGURE 3.79.
Saint Cloud Bridge, sequence of operations in moving launching girder.
Saint Cloud Bridge, France
in length, so as to be transportable over the highways. These units were assembled at the job site by
prestressing bars.
The seq u enc e o f o p eratio ns in m o v ing the
launching girder forward is illustrated in Figure
3.79 and included the following operations:
The gantry was supported on
three points: the rear leg, the central leg placed
near the end of the completed cantilever, and the
Placing pier se<gment:
121
temporary front leg supported just in front of the
pier.
The gantry slid on rails at
the rear leg and rolled over an auxiliary support
p lac ed ato p the p ier seg m ent. The c entral leg ,
during this travel, crossed the gap between the
cantilever end and the pier unit.
Launching
of the gantry :
In this phase
the gantry was supported at two points: the central
leg placed over the pier and the rear leg anchored
Placing ty pical segments in cantilever:
“2
3F6
FIGURE 3.80.
the river.
Saint Cloud Bridge, sequence of operations of launching gantry over
122
Precast Balanced Cantilever Girder Bridges
FIGURE 3.81. Angers Bridge, longitudinal section.
at the end of the last completed cantilever. The
segments were lifted by the trolley at the rear end
of the girder, moved forward, after a rotation of a
quarter turn, and then placed alternatively at each
end of the cantilevers under construction.
As a result of the horizontal curvature of the
structure, the transverse positioning of a segment
was accomplished both by moving the segment
transportation trolley sideways relative to the
girder [possible side travel of 3 ft (0.9 m) on either side] and by moving the launching gantry itself sideways relative to its bearing support on the
bridge. Thus, the construction of a cantilever required one, two, or three different positions of the
gantry, according to the curvature radius and
length of span, as shown in Figure 3.80. Work
started in October 1971 and was completed in Dkcember 1973. Placing the 527 precast segments in
the 3600 ft (1097 m) long superstructure took
exactly one year.
In terms of erection speed, a more interesting
project was successfully carried out on a precast
segmental bridge awarded,to Campenon Bernard.
A unique set of circumstances arose where a bridge
over the Loire River at Angers could be fitted to
use simultaneously the dimensions and casting
machines of Saint Andre de Cubzac Bridge, which
had recently been completed, and the gantry of
Saint Cloud Bridge.
The 2577 ft (786 m) long structure rests on 10
piers and has 280 ft (85.1 m) typical spans, Figures
3.81 and 3.82, using a single box girder with ribbed
FIGURE
structure.
3.82. Angers Bridge, view of the completed
deck slab units identical to the sections used at
Saint Andre de Cubzac. The construction contract
was signed in August 1974 and the superstructure
was completed in May 1975. All segments were
placed between January and May 1975, in a little
less than five months, corresponding to an average erection speed of 26 ft (8 m) per day of finished deck.
3.13 Sallingsund Bridge, Denmark
Sallingsund in Northern Jutland between Arrhus
and Thisted is a site of great natural beauty. Construction of a bridge in such an environment was
the object of careful study, which concluded, after
an international competition, in the selection of a
precast segmental structure, Figure 3.83, resting
on piers of a unique design.
This structure has two end spans of 167 ft (5 1 m)
and 17 interior spans of 305 ft (93 m). There are
18 piers between the two abutments. The level of
the roadway reaches 100 ft (30.5 m) above the
water at the center span and 82 ft (25 m) at the
abutments. The two center spans are navigation
spans requiring 85 ft (26 m) vertical clearance over
a width of 197 ft (60 m). The bridge deck accommodates two traffic lanes, approximately 13 ft (4
m) each, two cycle paths, and two sidewalks for a
total width of 52.5 ft (16 m), Figure 3.84. The
FIGURE 3.83. Sallingsund Bridge, view of the completed structure.
t
L
f
FIGURE 3.84. Sallingsund
Bridge, typical dimensions.
124
Precast Balanced Cantilever Girder
Bridges
superstructure consists of precast concrete box
girder segments 11.7 ft (3.57 m) in length, with
epoxy match-cast joints, which are prestressed together. Segment depth varies from 8.2 ft (2.5 m) at
midspan to 18 ft (5.5 m) at the pier.
The precast superstructure segments were
match-cast by the short-line method (see Chapter
11). There are altogether 453 segments varying in
weight from 86 t (78 mt) to 118 t (107 mt). The
typical segment shown in Figure 3.85 has web corrugated shear keys together with top and bottom
flange keys. Hinge segments equipped with a
roadway expansion joint for thermal movement of
the superstructure are placed every other span
near the point of contraflexure. A hinge segment
with its diaphragm is shown in Figure 3.86. Segments are placed in the structure in cantilever with
a cable-stayed launching gantry. Transfer from the
casting area and the storage yard to the construction site and the launching gantry is achieved by a
low-bed dolly pushed by a tractor, Figure 3.87.
The gantry shown in Figure 3.88 should look
FIGURE 3.85. Sallingsund Bridge, view of a typical
segment.
FIGURE
p o rt.
3.87.
Sallingsund
FIGURE 3.86. Sallingsund
with diaphragm.
FIGURE
3.88.
Sallingsund BI idge, I,~un&ing
Bridge,
hinge
Bridge,
segment rrans-
segment
g,~ntry.
B-3 South Viaducts, France
125
Figure 3.91 presents a plan of this project and
shows a subdivision in accordance with the type of
cross sections used. It includes the following main
subdivisio ns:
1.
2.
The main viaduct VP 1-A through VP 1-J.
The main viaduct VP 2-A and VP 2-B.
3. The viaducts Vl and V2, w hich are access
ramps to the main viaduct VP 2.
4. The viaducts V3 and V4, w hich are access
ramps to the National Road RN3.
FIGURE 3.89. Sallingsund Bridge, elevation of main
piers in water.
and have 860,000 sq ft (80,000 m”) of bridge deck.
The project is in a congested area that required the
crossing of railw ay tracks, canals, and more than 20
roads; its diverse structural geometry contains
curves, superelevation ranging from 2.5 to 6% and
grades up to 5%.
.
FIGURE 3.90. B-:l South Viaduct, overall view.
The original design for this project, prepared by
the French authorities, was based on conventional
cast-in-place construction of the superstructure in
complete spans using movable formwork. The
contractor proposed a more economical design
based on the use of precast segments. The alternative design had advantages in erection, wherein
parts were erected by a launching truss and parts
by a mobile crane in conjunction with an auxiliary
truss and winch. The use of precast units allowed a
deeper and thus a more economical superstructure, because the space required for formwork did
not have to be deducted in the clearance requirements over existing roads and other facilities.
The superstructure has a constant depth of 6.5 ft
(2 m), consisting of three different cross sections,
Figure 3.91. Different width and transitions were
accommodated by varying the width of the castin-place median slab connecting the top flanges of
the precast segments. Only the V3 and V4 access
ramps were of conventional cast-in-place construction.
The webs of the precast segments have a constant thickness of 12 in. (310 mm), increased in
some cases to 20 in. (500 mm) near a pier. Webs are
stiffened by an interior rib, which also serves to anchor the longitudinal prestressing inside the box
rather than in the web at the end of a segment.
Where the webs are not thickened near a support,
they are prestressed vertically by bars to accommodate shear forces. The top flanges of the segments
are cantilevered 10 ft (3 m). In the case of segment
types 2 and 3, Figure 3.9 1, the top flange cantilever
between box sections is 9 ft (2.75 m). The top
flange follows the superelevation of the roadway.
The thickness of the cast-in-place longitudinal slab
between box girders varies from 7.9 to 13.8 in. (200
to 350 mm), depending upon its width.
The total superstructure is supported on neoprene or sliding bearings. Expansion joints are
spaced at distances up to 1970 ft (600 m) and are
Precast Balanced Cantilever Girder Bridges
126
1 5 . 2 5 in
T -- - - - - - 1
T Y P E 1 795VOUSSOlR5.L.1,50ou2,5Om
TYPE
TYPE
2
1014 VCUSSOIRS.
3 392
VCkl5501RS.L~
L= 2,SO
2,SOw
w 3,401-n
3,40m
SUD
SOUTH
RN3
+
FIGURE
3.91.
B-3 South Viaduct, plan showing segment type location.
lo c ated in sp ec ial hing e jo ints near a p ier.
Superstructure spans vary from 89.6 to 174 ft (27
to 53 m), with 90% of them being in the range of
111 to 125 ft (34 to 38 m).
This project required 2225 precast segments, all
manufactured by the short-line method (see
Chapter 1 l), which involved the following operations:
8. Transfer of the segment, eight hours after
curing, to a more permanent storage until required for erection.
1.
1.
2.
A n initial l&hour curing period at 35°C.
A two-hour temperature rise reaching 65°C.
3.
A one-hour curing period at a level of 65°C.
2.
Subassembly of mild steel reinforcing on a
template.
Storage of subassembly units.
3.
Assembly of complete reinforcement cages including tendon ducts.
4.
Placing of the cages in the forms.
Concreting and curing of the segments.
5.
6.
7.
After concreting and curing, transportation of
the segment by a dolly to a position where one
end would act as a bulkhead for the casting of
the next segment. At the same time its position
w as ad justed to co nfo rm to the p ro p er
geometric configuration of the superstructure.
Transfer of the segment that had previously
acted as the bulkhead to temporary storage for
further curing.
9. Return of the mold bottom, after temporary
storage, to the casting area for reuse.
Curing of the segments was accomplished with
low-pressure steam in the following 4&-hour cycle:
The short curing cycle can be accomplished if the
following conditions are satisfied: use of a proper
cement, preheating of the materials to 35”C, rigid
forms, and proper supervision. Casting of a segment required nine hours, allowing two segments
per day per form; the four forms used produced a
total of eight segments per day.
Erection of precast segments by the launching
gantry show n in Figure 3.92 is schematically illustrated in Figure 3.93. After being rotated 90”,
segments V2 and V’2 were placed at the same time
by means of two trolleys suspended from the bottom chord of the launching girder, Figure 3.94.
B-3 South Viaducts, France
FIGURE 3.92. H-:5
in operation.
127
V2 and V’2 were then attached to the previously
erected segments by temporary prestressing.
During the erection operation of V2 and V’2 a
transport dolly delivered segment V’ 3, then V3,
and so on. In this manner the erection of segments
could be carried out without being delayed by
transportation of the segments from the storage
area. In addition, the threading and stressing of
the permanent prestressing tendons were independent of the erection cycle, since the tendons
were anchored in the internal ribs and could be
prestressed inside the box girder.
Where the span length w as less than 125 ft (38
m), the pier segments were placed by the gantry in
its normal working position. The pier segment position was adjusted from a platform fixed to the top
of the pier to avoid delaying the placement of cantilever segments at the preceding pier. For the few
South Viaduct, launching gantry
The matching faces of the segments being erected
and the previously erected segments, V 1 and V’l,
were coated with epoxy joint material. Segments
(b)
FIGURE 3.93. B-3 South Viaduct, erection sequence. (a) Placing the units: The two
trolleys bring the units V2 and V’2 which will be placed, after rotation at 90” , against
the units VI and V’l. During this time, the lorry carries the units V’3, then V3, and so on.
(b) Launching the truss: The rear and the central legs are lifted above the piers PO and Pl.
‘The truss is supported by trestles and trolleys in Pl and P2 and moves forward by the
action of the trolley motors until the legs reach Pl and P2. Thus the truss has advanced
along one span length and can place the pile-unit in P3 and the cantilevers from P2.
Precast Balanced Cantilever Girder Bridges
--
t.*
structure segments were placed simultaneously
by two different methods. The launching gantry
previously described placed 57% of the segments and a mobile crane in conjunction w ith a
movable winch frame erected the remaining ones.
The latter method was used where access was
available for a truck-mounted crane and the segment transportation dolly. The truck-mounted
crane could easily be used along the centerline of
the structure to place segments at outboard cantilever ends. However, its use became complicated
in the midspan area, particularly when it was used
to place the closure segments. To solve this problem, an auxiliary truss equipped w ith a w inch w as
used in conjunction with the mobile crane. This
truss was supported at one end over the pier where
cantilever construction proceeded and at the other
end over the last completed cantilever arm, which
might or might not require a temporary support
pier, Figure 3.95. The segments were lifted by a
trolley-mounted winch traveling along the truss.
This truss was also used to stabilize the cantilevers
during erection, since it was fixed to the pier and
the completed portion of the superstructure. After
the pier segment was positioned by the mobile
crane, the frame was launched with the trolley in a
counterweight position at the rear of the frame.
When the span exceeded 65 ft (20 m), the front of
the frame was held by the crane,
This structure exemplifies an innovative application of precast balanced cantilever segmental
construction to a difficult urban site and shows its
adaptability to almost any site conditions.
zyxwvutsrqpo
FIGURE 3.94. B-3 South Viaduct, placing two segments in balanced cantilever.
larger spans, the pier segment was placed after closure of the preceding completed spans and advancement of the launching gantry. The center leg
was advanced out onto the last completed halfspan cantilever, but it remained in the proximity of
the pier. Launching of the gantry to the next span
was achieved by using the two segment transportation dollies temporarily fixed on the completed
superstructure by two auxiliary steel trusses. The
high degree of mechanization of the gantry together with the repetitive nature of the project allowed speedy erection. A typical 130 ft (39 m) span
was erected and completed in two working days.
To maintain the construction schedule 2nd
minimize required erection equipment, the super-
FIGURE 3.95. B-3 South Viaduct, auxiliary truss for segment assembly (crane placing).
(1) Auxiliary truss, (2) winch for segment lifting, (3) precast segment, (4) possible tempnrary support (as required), and (5) concrete cantilever stability device.
Alpine
Motorway
3.15 Alpine Motorway Structures, France
The new Rhone-Alps Motorway system in South
East France includes 220 miles (350 km) of tollways, of which 60 miles (100 km) are an optional
section, between the cities of Lyons, Grenoble,
Geneva, and Valence in order to improve communications between Germany and Switzerland on
one hand and South France and Spain on the
other. The motorway is situated among the beautiful western slopes of the Alpine mountain range
(see the location map, Figure 3.96). The first 160
miles (250 km) include the following structures:
Ten viaducts varying in length between 500 and
1300 ft (150 to 400 m)
Two hundred overpass bridges
Fifty underpasses
Such a project afforded an exceptional occasion to
Structures,
France
129
optimize the structures in terms of initial investment and low maintenance costs.
The underpasses had to accommodate a variable
and often considerable depth of fill to reduce the
constraints of the longitudinal profile in this
mountainous region. The ideal answer was found
in the use of reinforced concrete arch structures,
which proved extremely well adapted and had a
cost approximately half that of conventional girder
bridges.
Apart from the first section of the motorway
(East of Lyons), which had to be built immediately
and therefore called for conventional solutions
(cast-in-place prestressed concrete slab), and except for certain special situations (excessive skew,
railroad crossing, and so on), a careful study
showed that the remaining 150 overpass bridges
should be of precast concrete segmental construction, which were 20% more economical than other
methods and practically maintenance free. The
study further showed that segmental construction
FIGURE 3.96. Alpine Motorway, location map.
130
Precast Balanced Cantilever Girder Bridges
should be extended to viaduct structures and that
all segments for both overpasses and viaducts
could be economically built in a single factory located near the center of gravity of the motorway
network. The maximum carrying distance was no
more than 75 miles (120 km) and the average was
40 miles (60 km). Figures 3.97 and 3.98 are views
of a typical viaduct and a typical overpass in the
motorway network.
The two-span and three-span overpass bridges
have spans ranging from 59 to 98 ft (18 to 30 m). A
variety of standardized precast cross sections were
developed for this project, depending upon span
and width requirements. The first structures used
single and double-cell trapezoidal box sections, although later on voided slab sections were preferred, as illustrated in Figure 3.99a. This solution
proved aesthetically pleasing and very simple to
manufacture and assemble. The viaducts had to
satisfy a wide range of environmental requirements. It was found that span lengths could be
limited at all sites to a maximum of 200 ft (60 m),
4
zyxwvut
2.60
c
\
+..
I
---. 4.m
-.
-4
(.
FIGURE 3.99. Alpine Motorway, typical sections of
overpass and viaducts. (a) Overpass segments. (b) Viaduct segments.
FIGURE 3 . 9 ’ 7 .
FIGURE 3.98.
Alpine Motorway, view of a viaduct
Alpine Motorway, view of an overpass.
which allowed a constant-depth superstructure
with precast segments, Figure 3.996.
Segment manufacture was carried out in a factory close to the new motorway with easy access to
the existing highway system, which was used to
haul all segments to their respective sites. The factory had two parallel bays, Figures 3.100 and
3.10 1, one for the overpass segments and one for
the viaduct segments. Segments for the overpasses,
Figure 3.100, were match-cast by the short-line
method with their longitudinal axis in a vertical
position. The bottom segment was a previously cast
unit. The segment at the top was then match-cast
against the segment on the bottom. After the unit
being cast had reached the required strength, the
bottom unit was removed for storage, and the en-
-zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
40'
PORTAL
(RAIitzyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
fb)
FIGURE 3.100. Alpine Motorway, precasting factory.
132
Precast Balanced Cantilever Girder Bridges
FIGURE 3.101. Alpine Motorway, general view of
precast factory and segment storage.
tire process repeated. Figure 3.102 is a view of a
segment in a vertical match-casting position.
Erection procedure for a typical three-span
overpass structure was as follows:
1 . After the foundations and pier columns had
been constructed, precast concrete slabs were
placed on sand beds adjacent to the piers to form
foundations for the steel falsework towers. The
precast slabs and towers were reusable for subsequent bridges. The erection commenced with
placement of the first segment on top of four partially extended 25-ton jacks, Figure 3.103~.
2. The second and third segments were placed
and p restressed to the first seg m ent, Fig ure
3.103b. The joints between the segments were
epoxy coated as the segments were erected. The
prestressing of the second and third segments to
the first segment consisted of temporary bars
above the top surface of the segments, and other
temporary tendons within the segments near the
bottom of the segments. The four 25-ton hydraulic
ja c ks under the first segment were then replaced
by four partially extended loo-ton hydraulic ja c ks
positioned under segments two and three. The
ja c ks were supported on teflon sliding bearings.
3. The remaining segments were then erected,
forming cantilevers on each side of the falsework
towers, Figure 3.103~.
The prestressing of the
segments consisted of temporary tendons positioned above the segments, as indicated in Figure
3.103.
4. The erection of the segments could ta ke
place simultaneously at both piers, or one could
precede the other, Figure 3.103d. Observe that at
this stage of erection each assembly of segments
was independently supported on four large hydraulic ja c ks and hence could be raised, lowered,
FIGURE 3.102. A lp ine hlororway,
casting o f segments.
vertical
m atc h
or rotated if required to adjust its position with respect to its pier or to its counterpart at the opposite
pier. This method eliminated the need for a castin-place closure joint at midspan
of the central
span. Through the adjustment of the hydraulic
ja c ks, perfect m ating o f the tw o centerm o st
match-cast segments could be achieved when the
assemblies of segments were slid together as indicated. The time required to erect the superstructure was significantly reduced by avoiding the use
of a cast-in-place closure joint.
5. At this point in the erection, the first group
of permanent prestressing tendons were inserted
in preformed holes through the segments, after
which they were stressed and grouted, Figure
3.103e.
6. The process proceeded with the erectiomof
the remaining segments, Figure 3.103f
7. After installation of precast match-cast
abutments, a second group of permanent tendons
was installed, and finally the temporary falsework
and temporary prestressing was removed, Figure
3.103g.
Alpine Motorway Structures, France
133
SECMENlS 18 b 2ST
SPAN
I8 te 30 m
zyxwvutsrq
J
lb)
TEMPORARY TIE URS
FIGURE 3.103. Alpine Motorway Bridges, erection
scheme for typical three-span overpasses. (a) Placing the
first and second segments. (b) Transfer to loo-ton jacks.
(c) First half completed. (d) Joining precast assemblies by
sliding. (e) Threading and stressing cables. v) Placing
the end segments. (g) Threading and stressing last cables.
Overpass structures of two spans could be
erected using the technique illustrated above for
three-span structures, Figure 3.104. As would be
expected, the longer spans required the use of additional falsework towers. An overpass bridge,
foundations plus piers and superstructure, could
be constructed in less than two weeks. Figure 3.105
shows a typical segment being placed in the over-
pass bridge with a mobile crane. Temporary prestress over the deck slab is shown in Figure 3.106.
The viaducts required the manufacture of larger
segments in the same precasting factory used for
the overpass segments, but with casting proceeding
in the usual short-line horizontal fashion. Three
casting machines were used simultaneously to produce all viaduct segments.
134
Precast
HYORAULIC
SLIDE
Balanced
JAcu5
Cantilever
Girder
Bridges
EMFORARY PRE5TR6’59
_
5ilDE
FIGURE 3.104. Alpine Motorway Bridges, erection scheme for two-span overpass
bridges.
Erecting segments in the various structures required the use of a launching gantry of an exceptionally light and elaborate design, allowing easy
transportation and erection from site to site, Figure
3.107. A typical 200 ft (60 m) long cantilever in-
eluding 25 segments, one pier segment weighing
48 t (44 mt), and 24 typical segments weighing 36 t
(33 mt) could be accomplished in six to eight
working days, including launching the gantry to
the following pier and achieving continuity with
the preceding cantilever. The maximum rate of
segment placing w as 12 units in a single day.
This project is another interesting application of
mass-production techniques and the standardization of segmental construction.
3.16
Bridge over the Eastern Scheldt, Holland
The bridge over the Eastern Scheldt, otherwise
known as the Oosterschelde Bridge, Figure 3.108,
FIGURE 3.105. Alpine Motorway, segment placing in
overpass with crane.
FIGURE 3.106. Alpine Motorway, provisional prestress over deck slab.
135
Br-idge Over the Eastern Scheldt, Holland
3.107. ,\lpine >lotol
viaducts with launching gantry.
FIGURE
wn, segment placing
in
time restraints for construction, and scarcity of
labor, prefabrication was required to a very high
degree. Since the precast pile elements would be
large and heavy, it was decided that the pier and
superstructure segments should be equally large
and heavy, in the range of 400 to 600 tons.i6
A casting yard, Figure 3.110, capable of producing all the various precast elements for the
structure was constructed near one end of the
bridge. This facility provided all the advantages of
yard production techniques and the potential for
high quality control.
The 14 ft (4.27 m) diameter cylinder piles have
14 in. (0.35 m) thick walls and were cast vertically in
20 ft (6 m) lengths. They were then rotated into a
horizontal position where they were aligned,joints
concreted, and the pile post-tensioned. In this
manner piles were produced in required lengths
up to 165 ft (50 m). The assembled pile was then
transported by barge to the site, where a derrick
picked it up at one end and rotated it into its verti-
600 tons
FIGURE 3.108. Bridge over the Eastern Scheldt,
overall view of the structure.
is part of a project known as the Delta Works,
which closed the mouths of many rivers and
streams southwest of Rotterdam to protect the
coastline from flooding. The bridge consists of
fifty-five 300 ft (9 1.4 m) spans, a roadway width of
35 ft (10.7 m), and a vertical navigation clearance
of 50 ft (15.2 m). Parameters considered in the
choice of structural type and span were economics,
foundation restraints, and ice loads.
Substructure consists of three cylinder piles with
a caisson cap and an inverted V pier, Figure 3.109.
The superstructure was assembled from seven
precast elements, one pier segment, and two each
of three progressively smaller segments to produce
one double cantilever span of 300 ft (91.4 m). The
bridge design, therefore, consists of very large prestressed cylinder piles, precast pier elements posttensioned together, and precast superstructure
elements erected and post-tensioned together to
form a double cantilever system with a joint at each
midspan location. Because of open-sea conditions,
C ytindrm l
ho llo w
FIGURE 3.109. Bridge over the Eastern Scheldt,
schematic of precast elements in the structure (courtesy
of the Portland Cement Association).
3.110. Bridge o ve r the Eastern Scheldt, view
of precasting plant (courtesy of the Portland Cement Association).
FIGURE
136
Precast Balanced Cantilever Girder Bridges
cal position. Cylinder piles weighted from 300 to
550 tons (270 to 500 mt). The pier cap w as also
precast at the same yard, where it was post-tensioned circumferentially and vertically. The inverted V portion of the pier was also precast with
provision for on-site post-tensioning to achieve final
assembly.16
Figure 3.111 shows the bridge under construction. The temporary enclosures between each section are to protect the cast-in-place joint concrete
against cold weather. Cast-in-place joints 16 in. (0.4
m) wide were used, with faces of the precast elements serrated to act as shear keys.
The superstructure segments were all set from a
traveling steel gantry, Figure 3.111, that extended
over two and one-half spans at a time. Segments
were barged to their final location, then hoisted in
symmetrical order about each pier. The joints were
concreted and the primary stressing completed be-
fore the next series of segments were hoisted into
position. Erection sequence is depicted in Figure
3.112. An aerial view of various stages of construction is shown in Figure 3.113. A typical cycle for
two spans of superstructure, not including the pier
segment, involving the raising, concreting, and
stressing of 12 segments, was three weeks.
3.17 Captain Cook Bridge, Australia
FIGURE 3.111. Bridge over the Eastern Scheldt, view
of launching truss and enclosure for cast-in-place joints
(courtesy of the Portland Cement Association).
This structure carries a six-lane highway over the
Brisbane River in Brisbane, Australia, as part of
the Riverside Expressway and South-West Freeway
designed to relieve the city’s overloaded traffic
system.
The navigation requirements were for a 300 ft
(91.4 m) wide horizontal clearance with a vertical
clearance of 45 ft (13.7 m) across 200 ft (61 m) and
40 ft (12 m) at either extremity. However, a 600 ft
(183 m) span became necessary because of the
skew crossing. Adequate bearing rock, at a reasonable depth, was found at the south bank such that
the pier could be founded on a spread footing. At
the north end, because of the steeply rising bank,
the anchor span w as limited to a span of 140 ft
(42.7 m) and the abutment was designed as a
counterweight connected to the superstructure by
a prestressed tie-down wall, Figure 3.1 14.17
Once the navigation span requirements had
been met, the remaining span lengths were selected to meet design requirements, while the
superstructure depth boundaries had to fall within
a maximum allowable grade requirement of 3%
and the flood level. The superstructure is a dual
FIGURE 3.112. Bridge over the Eastern Scheldt,
schematic of erection sequence (courtesy of the Portland
Cement A sso ciatio n).
FIGURE 3.113. Bridge over the Eastern Scheldt, aerial view of construction showing various phases (courtesy of the Portland Cement Association).
ELEVATION
FIGURE 3.114. Capt. Cook Bridge, plan and elevation, f‘rom ref.. 17.
Precast Balanced Cantilever Girder Bridges
138
structure of prestressed concrete segmental twocell boxes, Figures 3.115 and 3.1 16.17
Steel rocker bearings were used to support the
superstructure at piers 1, 3, and 4, and largediameter single steel roller bearings were used at
pier 2. Lubricated bronze bearings sliding on
stainless steel were used at the north abutment and
for the movable bearings at the suspended spans.
Steel finger joints, allowing a 10 in. (250 mm)
maximum movement, were provided at each slid-
II
H-=---
FIGURE 3.115. Capt. Cook Bridge, cross section at
pier 3, from ref. 17.
FIGURE 3.116. Crpt. Cook Bridge, two-cell box girder segment being erected (courtesy of G. Beloff, Main
Roads Department).
ing bearing location and rubber and steel finger
joints at the remaining locations.”
The box girder segments have a maximum
depth of 32 ft (9.75 m) and a minimum depth of 6
ft (1.83 m). Segment length is 8 ft 8 in. (2.64 m). A
16 in. (0.4 m) cast-in-place, fully reinforced joint
was used between segments. Maximum segment
weight is 126 tons (114 mt). A total of 364 precast
segments were required in the superstructure with
the two segments over the tie-wall in the south
abutment being cast in place.”
The ContracEor chose to locate the precasting
operation on the river bank near the south abutment. This casting yard consisted of a concrete
mixing plant, steam-curing plant, three adjustable
steel forms, segment tilting frame, and a gantry
crane to transport the segments to a storage area
along the river bank. Segments were designed so
that the top flange and upper portion of the webs
had a constant thickness. The depth and lower
portion accommodated all variations, allowing the
contractor to cast in two sets of adjustable forms.
Segments were cast with their longitudinal axis in a
vertical position for ease of concrete placement
around the prestressing ducts. Separate interior
forms were constructed for each box to permit
variations in the bottom flange and web thickness
and size of fillets. Aft.er casting and curing, segments were lifted into a tilting frame to realign the
segment into its normal position ready for handling and storage.i7
A floating crane, designed and built by the contractor, was used for erection of the segments. It was
essentially a rectangular pontoon with mounted
A-frame lifting legs rising to 120 ft (36.6 m) with
adequate clearance to service the finished deck
level, while the stability was sufficient to transport
the segments to the erection position, Figure 3.117.
An extended reach was required to position segments on the first two spans in the shallow water
near the bank.17
Segments on each side of the pier were supported on falsework anchored to the pier shafts,
Figure 3.118. From this point additional segments,
as they were erected, were supported on a cantilever falsework from the completed portion of
the structure. This falsework was fixed under the
completed girder and supported from deck level,
Figure 3.119. When the capacity of the pier to
carry the segment unbalanced load was reached, a
temporary prop support on driven piles was constructed before cantilever erection could continue..
Segment erection then proceeded on each side
until either the joint position of the suspended
Other
Notable
Structures
139
FIGURE 3.119. Capt. Cook Bridge, cradle support
trusses and temporary support tower (courtesy of G.
Beloff, Main Roads Department).
FIGURE 3.117. Cap. Cook Bridge, segment being
transported by barge derrick to final position (courtesy
of G. Beloff, Main Roads Department).
span was attained or the closure gap in span 3 was
reached. The completed structure was opened to
traffic in 1971, Figure 3.120.
3.18 Other Notable Structures
In Sec tio ns 3.2 thro u g h 3.15 the histo ric al d e-
velopment of precast segmental bridges with
match-cast joints has been illustrated by examples,
ranging from the first structure at Choisy-le-Roi to
the largest applications such as the Rio Niteroi and
Saint Cloud bridges. Emphasis has been placed
on North American experience as well as on the
advantages of precast segmental construction for
urban structures (B-3 Viaducts) or repetitive applications (Alpine Motorways). Two particularly
outstanding structures, deserving special mention
because of their size and characteristics where pre-
cast segmental was used with conventional joints
(not match-cast) were the Oosterschelde and Captain Cook Bridges (Sections 3.16 and 3.17). Before
closing this important chapter, let us briefly give
due credit to several other contemporary matchcast segmental bridges.
3.18.1
CALIX BRIDGE, FRANCE
This 14-span superstructure has a maximum span
leng th o f 512 f t ( 156 m ) o v er the m aritim e
FIGURE 3.118. Capt. Cook Bridge, support for segments on each side of pier (courtesy of G. Beloff, Main
Roads Department).
FIGURE 3.120. Capt. C o o k Brid g e, c o m p leted
structure (courtesy of G. Beloff, Main Roads Department).
Precast Balanced Cantilever Girder Bridges
140
1.39
3.h2
13.42
e
FIGURE
L
3.121. Calix Viaduct, near Caen, France general dimensions.
waterway and typical 230 ft (70 m) spans in the ap-
proaches on both banks. Dimensions are show n in
Figure 3.12 1. The deck consists of two parallel box
girders connected by a precast prestressed slab
strip. All segments, with a maximum weight of 49 t
(43 mt), were cast in a long bench and placed with a
tower crane traveling between the box girders in
the approaches. Segments were barged in for the
main span, and a beam and w inch system w as used
for hoisting them into place, Figure 3.122.
3.18.2 VAIL PASS BRIDGES, U.S.A.
FIGURE 3.122. Calix Viaduct, placing precast segments in superstructure.
These bridges are located on Interstate I-70 over
Vail Pass near Vail, Colorado, in a beautiful setting at an altitude betw een 9000 and 10,000 ft
(2700 and 3000 m) above sea level where winter
conditions are critical and the construction period
is very short. Dimensions are shown in Figure
3.123, and a view of one finished bridge appears in
Figure 3.124.
3.18.3 TRENT VIADUCT, U.K.
Section near midspan
FIGURE 3.123. Vail Pass Bridge, cross-section general dimensions.
This structure carries the M-180 South Humberside motorway over the River Trent and consists of
dual roadways of three lanes each, with a central
median. Precast segmental construction was selected against a steel plate girder design w ith a
reinforced concrete deck slab. The bridge is sym-
Other Notable Structures
141
FIGURE 3.124. \‘A Pass bridge, a completed precast
segmental structure (courtesy of International Engineering Company, Inc.).
F I G U R E 3 . 1 2 6 . I‘rellt Bridge,
finishing the deck.
metrical with four spans of 159, 279, 279, and 159
ft (48.5, 85, 85, and 48.5 m).
Each roadway is supported by an independent
superstructure of twin concrete box girders varying in depth from 16 ft (4.9 m) at the piers to 7 .ft
(2.1 m) at midspan of the center spans. Principal
dimensions are shown in Figure 3.125. Each box
girder is made up of 91 precast segments 10 ft (3
m) long, varying in weight between 38 t (35 mt) to
82 t (75 mt). All segments were placed in balanced
cantilever with a launching gantry shown in operation in Figure 3.126, with precast units being delivered on the finished deck.
3.18.4
L-32 TAUER,~AUTOBAHN
l a u n c h i n g ganrry
ft (33.5, twenty at 55, and 33.5 m). Box piers have a
maximum height of 330 ft (100 m). The constantdepth superstructure of 12.5 ft (3.8 m) is made up
of 722 segments match-cast in a job-site factory
equipped with four casting machines, Figure
3.127. A launching gantry was used to place all
segments in the two bridges in balanced cantilever,
Figure 3.128.
3.18.5 KISHWAUKEE RIVER BRIDGE, U.S.A.
This dual structure carries U.S. Route 51 over the
Kishwaukee River near the city of Rockford, Illinois. Dimensions are shown in Figure 3.129. Prestressing is achieved in the transverse and longitudinal directions by bar tendons. All segments
were placed in the structure by a launching gantry,
shown in Figure 3.130, which represents the first
application of this method in the United States.
BRIDGE, AUSTRIA
This structure is located between Salzburg and
Villach, Austria, as part of a new motorway connecting Germany and Yugoslavia. The 22-span
twin bridge has a total length of 3820 ft (1167 m)
distributed as follows: 110, twenty at 180, and 110
17.400
MOTORWAY
CENTRAL
L RESERVE
4
-
zyxwvutsrqpo
INSITU
JOINT
\
t
WEST
I-
zyxwvutsrqponm
I-
NAVIGATION CNANNEL
EAST
El e v at i o n
FIGURE 3.125. Trent Bridge, typical dimensions.
FIGURE 3.127. L-32 Tauernautobahn Bridge, casting machine.
3.18.6 KENTUCKY RIVER BRIDGE, U.S.A.
This structure crossing the Kentucky River is located in Franklin County just south of Frankfort,
Kentucky. It is a three-span structure with a 323 ft
(98.5 m) center span and 228.5 ft (70 m) side spans.
In cross section the superstructure consists of two
rectangular boxes. It is,the first precast segmental
bridge to be constructed in the United States using
the long-bed casting method, Figure 3.131. A view
during construction is show n in Figure 3.132.
FIGURE 3.128. L-32
launching gantry.
Tauernauto bahn
Bridge,
3.18.7 I-205 COLUMBIA RIVER BRIDGE, U.S.A.
This large project represents one of the major applications of precast segmental construction in the
United States. The 5770 ft (1759 m) long structure
carries Interstate I-205 from Vancouver, Washington, across the North Channel of the Columbia
River to Government Island near Portland, Oregon. Twin structures carry two 68 ft (20.7 m) wide
roadways with span lengths varying between 600 ft
(183 m) and 242 ft (74 m). Typical dimensions ofzyxwvutsrq
170’-0’
I_--
ELEV 694.0
-TRANSVERSE
POST-TENSIONIN
21
0
-!k-Ao.‘-L+fb)
._
fcJ
FIGURE 3.129. Kishwaukee River Bridge, superstructure elevation and cross sections.
(a) Elevation. (b) Section at midspan. (c) Section at pier. (From ref. 18.)
Other Notable Structures
143
FIGURE 3.131. Kentucky River Bridge, long-line
casting bed.
FIGURE 3.130. Klrhwaukte River Bridge, v~elv during construction showmg launching truss.
the main spans over the river are shown in Figure
3.133. Dimensions of the cross section, as designed,
are shown in Figure 3.134. However, the contractor, under a value engineering option in the contract documents (see Chapter 12), elected to redesign the cross section to a two-cell box section,
Figure 3.135. The contractor exercised the op-
FIGURE
3.133.
FIGURE
struction.
3.132.
kcntut
k\ Rncxt 131 idgc. (In1 111% con-
tion allowed in the bidding documents to select
his own construction method and proceeded with
casting in place in conventional travelers the two
cantilevers adjacent to the main navigation channel (piers 12 and 13), w hile all other spans are
of precast segmental construction. Figure 3.136
shows a rendering of the structure.
I-205 Columbia River Bridge, elevation and plan.
.
Precast Balanced Cantilever Girder Bridges
144
ll'-10"
I
67'-10"
1
I
67'-11"
I
I
I
FIGURE 3.134. I-205 Columbia River Bridge, cross sections.
CROSS
FIGURE
3.135.
SECTION
OF
PRECAST
SEGMENTS
!
372’ ,
I-205 Columbia River Bridge, revised
cross section.
3.18.8
ZILWAUKEE BRIDGE, U.S.A.
This bridge is another important example of precast segmental construction in the United States.
Located in central Michigan, this 8080 ft (2463 m)
long structure carries dual four-lane roadways
over the Saginaw River near Zilwaukee, Michigan.
Principal dimensions are shown in Figure 3.137.
366’i
FIGURE
FIGURE 3.136. I-205 Columbia Rner Bridge, ren-
dering of the structure.
389’
3.137.
377’ !
392’
368’ j
372’ 1351’
Zilwaukee Bridge, typical dimensions.
The 5 1 spans vary m length from 155 ft to 392 ft
(47 to 119 m). An additional three-span ramp carries some traffic onto the southbound high-level
bridge. Navigation clearance is 125 ft (38 m) above
the Saginaw River.
For a total deck area of 1,180,OOO sq ft (110,000
145
O t her No t able St ruct ures
I
11.70
FIGURE
3.138.
Ottmarsheim
Bridge,
general
dimensions.
m*) t h e r e a r e 1 5 9 0 l a r g e s e g m e n t s v a r y i n g i n
length from 8 to 12 ft (2.4 to 3.65 m) with a
maximum weight of 160 t (144 mt). Segments were
produced in a production-line operation with
short-line casting and placed in the structure in
balanced cantilever with a large launching gantry
accommodating two successive spans.
weighing a maximum of 50 t (45 mt) are designed
to be placed in balanced cantilever with an auxiliary overhead truss (and winch system) in the
approach spans to stabilize the deck over the flexible piers during construction.
3.18.9 OTTM ARSHEIM BRIDGE, FRANCE
This very important project is a recent application
of precast segmental construction to urban elevated structures. The constraints relating to location of piers and construction over highway and
railway traffic are comparable to the conditions encountered at the B-3 South Viaducts in Paris,
France.
The principal project dimensions are shown in
Figure 3.142. All segments will be placed in the
twin bridge using two launching gantries, which
incorporate the latest technological developments
in safety and efficiency.
This bridge in East France close to Germany and
the Rhine River at the Ottmarsheim hydroelectric
plant is today the longest clear span of precast
segmental construction and the first major application of lightweight concrete to this type of
structure. Principal dimensions are shown in Figure 3.138. As shown in the longitudinal section,
lightweight concrete was used only in the center
portion of the two main spans over the navigable
waterway and over the outlet channel of the power
plant. Figure 3.139 is a view of the completed
structure.
3.18.11 F- 9 FREEW AY, M ELBOURNE, AUSTRALIA
3.18.10 OVERSTREET BRIDGE, FLORIDA, U.S.4.
This structure crosses the lntracoastal Waterway
near Panama City in Western Florida. Dimensions
are shown in Figures 3.140 and 3.141. The main
navigation span is 290 ft (88 mm) long between
piers to avoid any construction in the water fender
system during operation. Approach spans are 125
ft (38 m) long and rest on I-shiped piers bearing
on precast piles. The main piers consist of twin I
piers of the same design. The total length of
structure is 2650 ft (808 m) divided as follows: 95,
seven at 125, 207.5, 290, 207.5, seven at 125, and
95 ft (29, seven at 38, 63, 88, 63, seven at 38, and
29 m). Precast segments 10 ft (3 m) long and
FIGURE 3.139. Ottmarsheim Bl-idge,
completed
structure.
vic\v of’ the
2650’-0”
Overall Length of Bridge
21t
207’b!i”
;125,-O&l
25’-0225’.0’2
25’-0’~25’-0~125-0’~125-0’~i’-0r;l
1
2’-6”
-4-
zyxwvutsrqponmlkjih
Sand Cement 3 :
Riprap. (Typ.1
FIGURE 3.140. Overstreet Bridge, blot-da,
elevation
L2ig”
t,
LOLO”
I
al
FIGURE 3.141. Overstreet Bridge, Florida, cross sections.
References
References
1. Jean Muller, “ Ten Years of Experience in Precast
Segmental Construction,” Journal of the Prestressed
Concrete Institute, Vol. 20, No. 1, January-February
1975.
2. C. A. Ballinger, W. Podolny, Jr., and M. J. Abrahams, “ A Report on the Design and Construction
of Segmental Prestressed Concrete Bridges in Western Europe- 1977,” International Road Federation, Washington, D.C., June 1978. (Also available
from Federal Highway Administration, Office of
Research and Development, Washington, D.C., Report No. FHWA-RD-78-44.)
3. Walter Podolny, Jr., “ An Overview of Precast Prestressed Segmental Bridges,” Journal of the Prestressed
Concrete Institue, Vol. 24, No. 1, January-February
1979.
4. J. Mathivat, “ Reconstruction du Pont de Choisy-leRoi,” Travaux, Janvier 1966, No. 372.
5. Jean Muller, “ Long-Span Precast Prestressed Concrete Bridges Built in Cantilever,” First International
Symposium, Concrete Bridge Design, Paper SP 23-40,
AC1 Publication SP-23, American Concrete Institute, Detroit, 1969.
6. Andre Bouchet, “ Les Ponts en Beton Precontraint
de Courbevoie et de la Grande-Jatte (Hauts-deSeine),” La Technique des T r a v a w , Juillet-Aout
1968.
7. “ Bear River Bridge,” STUP Bulletin of Information,
November-December 1972.
8. “ Nova Scotia’s Bear River Bridge-Precast Segmental Construction Costs Less and the Money
Stays at Home,” Bridge Bulletin, Third Quarter 1972,
Prestressed Concrete Institute, Chicago.
9. “ John F. Kennedy Memorial Causeway, Corpus
Christi, Texas,” Bridge Report SR 162.01 E, Portland Cement Association, Skokie, Ill., 1974.
10. G. C. Lacey, and J. E. Breen, “ Long Span Pre-
147
stressed Concrete Bridges of Segmental Construction State of the Art,” Research Report 12 l-l,
Center for Highway Research, The University of
Texas at Austin, May 1969.
1. S. Kashima and J. E. Breen, “ Epoxy Resins for
Jointing Segmentally Constructed Prestressed Concrete Bridges,” Research Report 121-2, Center for
Highway Research, The University of Texas at Austin, August 1974.
2. G. C. Lacey and J. E. Breen, “ The Design and Optimization of Segmentally Precast Prestressed Box
Girder Bridges,” Research Report 121-3, Center for
Highway Research, The University of Texas at Austin, August 1975.
13. R. C. Brown, Jr., N. H. Burns, and J. E. Breen,
“ Computer Analysis of Segmentally Erected Precast
Prestressed Box Girder Bridges,” Research Report
121-4, Center for Highway Research, The University of Texas at Austin, November 1974.
14. S. Kashima and J. E. Breen, “ Construction and Load
Tests of a Segmental Precast Box Girder Bridge
Model,” Research Report 121-5, Center for Highway Research, The University of Texas at Austin,
February 1975.
15. J. E. Breen, R. L. Cooper, and T. M. Gallaway,
“ Minimizing Construction Problems in Segmentally
Precast Box Girder Bridges,” Research Report
121-6F, Center for Highway Research, The University of Texas at Austin, August 1975.
16. Ben C. Gerwick, Jr., “ Bridge over the Eastern
Scheldt,” Journal of the Prestressed Concrete Institute,
Vol. 11, No. 1, February 1966.
17. “ A Pro u d A c hiev em ent- The C ap tain C o o k
Bridge,” Issued by the Commissioner of Main
Roads-1972, Main Roads Department, Brisbane,
Queensland, Australia.
18. “ Prestressed Concrete Segmental Bridges on FA 412
over the Kishwaukee River,” Bridge Bulktin, No. 1,
1976, Prestressed Concrete Institute, Chicago.
zy
4
Design of Segmental
4.1
4.2
4.3
4.4
4.5
4.6
4.7
4.8
INTRODUCTION
LIVE LOAD REQUIREMENT?3
SPAN ARRANGEMENT AND RELATED PRINCIPLES
OF CONSTRUCTION
DECK EXPANSION, HINGES AND CO 4.4.1 Hinges at Midspan
4.4.2 Continuous Su~ tructures
4.4.3 Expansion of Long Bridge
TYPF, SHAPE AND DIMENSIONS OF THE SUPERsTRu4.5.1 Box Sections
4.5.2 Sbape of Superst~ cture in Elevation
4.5.3 Choice of Typical Cross Section
4.5.4 Dimensions of the Typical Cuss Section
TRANSVERSE DISI’RIBUI’ION OF LQADS BETWEEN
BOX GIRDERS IN MULTIBOX GIRDERS
EFFECT OF TEMPFXATI-JRF,
GRADIENTS IN BRIDGE
suPFRsl-RucrUREs
DESIGN OF LONGITUDINAL MEMBERS FOR FLEXURE AND TENDON PROFILES
4.8.1
4.8.2
4.8.3
4.8.4
4.8.5
Principle of Pre&ess Iayout
Draped Tendons
Shaight Tendons
Summary of Tendon Profiles and Anchor Locations
Special Problems of Continuity PresWss and Ancbonge Thereof
4.8.6 Iayout of Pmskess in Strucaups with Hinges and
Expansion Joints
4.8.7 Redistribution of Moments and Stresses Through
concrete creep
4.1
Introduction
Design of concrete highway bridges in the United
States conforms to the provisions of The American
Association for State Highway and Transportation
Officials (AASHTO) “ Standard Specifications
for Highway Bridges.” For railway structures,
specifications of the American Railway Engineers
Association (AREA) should be consulted. For the
148
4.9
4.10
Bridges
4.8.8
Prediction of Preskess Losses
ULTIMATE BENDING CAPACITY OF LONGITUDINAL MEMBERS
SHEAR AND DESIGN OF GROSS SECITON
4.10.1
Introduction
4.10.2 Shear Tests of Reinforced Concx~te
Beams
4.103
DifIiculties in Actual Structmw
4.10.4 Design of h@dinal Members for Shear
4.11 JOINTS BETWFEN MATCH-CAST SEGMENTS
4.12 DESIGN OF SUPERSTRUCl-URE CROSS SECl’ION
4.13 SPECIAL PROBLEMS IN SUPmUCIWRE
DESIGN
4.14
4.15
4.16
4.17
4.13.1
Diapluagms
4.13.2 Superstructure over Piers
4.13.3 End Abutments
4.13.4 Expansion Joint and Hinge Segment
DEFLECITONS O F CAN TI LE V E R B R I D GE S AN D
CAMBER DESIGN
FATIGUE IN SEGMENTAL BRIDGES
PROVISIONS FOR FUTURE PmIN G
DEhGN FXAMPLE
4.17.1 Longitudinal Beding
4.17.2 Redktribution of Moments
4.17.3 Stresses at Midspan
4.17.4
shear
4.17.5 Design of the Cross-Section Frame
4.18 QUANTITIES OF MATERIALS
4.19 POTENTL4L PROBLEM ARF.AS
REFERENCES
most part, the provisions in these specifications
were written before segmental construction was
considered feasible or practical in the United
States.
Before discussing design considerations, the
authors wish to emphasize that no preference for
either cast-in-place or precast methods of construction is implied here. The intent is simply to
present conditions that the designer should be
Span Arrangement and Related Principles
aware of to produce a satisfactory design. Both
concepts are viable ones, and both have been used
to produce successful structures.
In general, the segmental technique is closely
related to the method of construction and the
structural system employed. This is why segmental
construction, either cast in place or precast, has
been often identified with the cantilever construction use d in so many applications. It is logical to
ta ke bridge structures built in cantilever as a basis
for the design considerations developed in this
chapter. Where other methods, such as incremental launching or progressive placement, require
special design considerations, such problems are
discussed in the appropriate chapters.
4.2 Live-Load Requirements
In comparing practices in other countries to those
in the United States, an important parameter to
keep in mind is that of live-load requirements. Figure 4.1 illustrates the considerable differences
among code requirements in various countries.’
For a simple span of 164 ft (50 m) and width of
24.6 ft (7.5 m), the German specification requires a
live-load design moment 186% greater and the
French requires one 290% greater than that of
AASHTO. Some Canadian provinces use the
AASHTO specifications but arbitrarily increase the
live load by 25%.
4.3
Span Arrangement and Related Principles
of Construction
zyxwvutsr
CPC
290
177
f
5000
P
t
A
//
4 0 0 0 l- q M a x . M
8
I
0
France-
DIN l07i
10
20
30
40
/
/
/
/
50
60
70
80
90
loo
ah)
Span
FIGURE 4.1. M aximum
149
In the balanced cantilever type of construction,
segments are placed in a symmetrical fashion about
a pier. The designer must always remember that
construction proceeds with symmetrical cantilever
deck sections centered about the piers and not with
completed spans between successive piers.2
For a typical three-span structure, the side spans
should preferably be 65 percent of the main center
span instead of 80 percent in conventional castin-place structures. This is done to reduce to a
minimum the length of the deck portion next to
the abutment, which cannot be conveniently built
in balanced cantilever, Figure 4.2~.
Where span lengths must vary, as between a
main span and an approach span, it is best to introduce an intermediate span whose length will average the two flanking spans, Figure 4.26. In this
manner the cantilever concept is optimized.
Individual cantilever sections are generally made
continuous by insertion of positive-moment ten-
A A S H T O IRC
50
100
138
100
loo
138
M km)
Construction
The depth-to-span and width-to-depth ratios for
segmental construction presently advocated in the
United States have been adopted from European
practice. The lighter live loads used in the United
States should permit further refinements in our
design approach.
AASHTO 100%
DIN 1072
186
173
P(m)
of
live-load moment (simple span)
(F. Leonhardt, New Practice in Concrete Structures, IABSE, New
York, 1968).
Design of Segmental Bridges
0.65-07OL
065-O
70L
(a)
LI
,
I
‘ 2 (LITL2)
L2
..I
(b)
Section A-A
FIGURE 4.4. End restraint at abutment.
fc)
FIGURE 4.2. Cantilever construction showing choice
of span lengths and location of expansion joints.
dons upon closure. It is preferred not to have any
permanent hinges at midspan. Continuous decks
without joints have been repeatedly constructed to
lengths in excess of 2000 ft (600 m) and have
proved satisfactory from the standpoint of maintenance and riding quality.
For very long viaduct-type structures, intermediate expansion joints are inevitable to accommodate volume changes. These joints should be located near points of contraflexure, Figure 4.2c, to
avoid objectionable slope changes that occur if the
joint is located at midspan. This consideration will
be discussed in more detail in Section 4.4.
In many cases it may not be possible to provide
the desirable optimum span arrangement. Thus,
the end span may be greater or less than the optimum span length desired.2 In the case of a long
end span, the superstructure might be extended
over the abutment wall to provide a short additional span. As shown in Figure 4.3, a conventional
bearing (1) is provided over the front abutment
wall. A rear prestressed tie (2) opposes uplift and
permits cantilever construction to proceed outward from the abutment to the joint t’Jl), where a
connection can be effected with the cantilever from
the first intermediate pier. Figure 4.4 shows an alternative scheme with a constant-depth section, as
opposed to a haunched section, where the deck has
been encased within the abutment wing walls for
architectural purposes. For the normal end span, a
special segment is temporarily cantilevered out so
as to reach the first balanced cantilever constructed
from the next pier, Figure 4.5. Alternatively this
portion could be cast in place on falsework, if site
conditions permit.
In a short-end-span situation, cantilever construction starts from the first pier and reaches the
abutment on one side well before the midspan section of the adjacent span, Figure 4.6. An uplift
reaction must be transferred to the abutment
during construction and in the completed structure. Consequently, the webs of the main box
girder deck are cantilevered over the expansion
FIGURE 4.3. End restraint in abutment.
Deck Expansion, Hinges and Continuity
151
FIGURE 4.5. Conventional bearing on abutment.
FIGURE 4.6. Anchorage for uplift in abutment.
joint into slots provided in the main abutment
w all, Fig u re 4.7. The neo p rene b earing s are
placed above the web cantilever rather than below
to transfer the uplift force while allowing the deck
to expand f-reely.
Interesting examples of such concepts are given
in the three following bridges:
G iv o rs Brid g e o v er the Rho ne Riv er, Franc e,
shown in Figure 4.8. The main dimensions are
given with the typical construction stages of the
superstructure.
duced by the use of lightweight concrete in the
center of the main span.
Puteaux Bridges over the Seine River, near Paris
(Section
2.15.10).
A few bridges have even been built in cantilever
entirely from the abutments. The Reallon Bridge
in Frarice is one such structure, Figure 4.10, where
very special site conditions with regard to bridge
profile and shape of the valley were best met with
this concept.
Another set of circumstances may be encountered when it is not possible to select the desired
span lengths to optimize the use of cantilever construction. Such was the situation of the bridge over
the Seine River for the Paris Ring Road, where a
side span on the left bank could not be less than 88
percent of the main river span over the river, while
very stringent traffic requirements governed the
placement pattern of precast segments on the right
bank, Figure 4.11.
Tric astin Brid g e o v er the Rho ne Riv er, Franc e
(Section 2.15.11). No river piers were desired for
the structure, which dictated a main span of 467 ft
(142.50 m), and there w as no room on the banks to
increase the side spans so as to avoid the end uplift.
Two very short side spans of only 83 ft (25.20 m)
provide the end restraint of the river span. The
uplift is transferred to the abutments, which are
earth tilled to provide a counterweight, Figure 4.9.
The magnitude of the uplift force has been re-zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
4.4
Deck Expansion, Hinges and Continuity
4.4.1 HINGES AT MIDSPAN
Prestressing
FIGURE 4.7.
Longitudinal section.
units
Historically, the first prestressed concrete bridges
built in cantilever were provided with a hinge at
the center of the various spans. Such hinges were
designed to transfer vertical shear between the tips
of two adjacent cantilever arms (which could develop under the live loading applied over one arm
only in half the span length) while enduring a free
exp ansio n o f the c o nc rete d ec k u nd er v o lu m e
changes (concrete creep and seasonal variations of
temperature). Continuity of the deflection curve
R I M CiMCHE 6
s
FIGURE 4.8.
152
3
2
Givors Bridge over the Rhone River, France, span dimensions and typical
construction stages. (1) Construction of left bank river pier segment. The eight segments
either side of the pier are erected, and pier stability is assured by temporary props. (2)
The connection between deck and abutments is made. Temporary props are removed
and the seven remaining segments are placed in cantilever. (3) The above operation is
repeated on the right bank. The central pier segments are poured. Two segments are
erected on either side of each pier, supported by scaffolding. (4) The last segment is
placed in the central span, continuity is achieved between the two cantilevers, and the
scaffolding is removed. (5) The remaining 16 segments on either side of the central piers
are placed. (6) The 110 m spans are completed by pouring the closure segments and
tensioning the continuity prestress. The superstructure is now complete.
1
RP.‘E
CRXTE
Elevation
Section
A-A
’
’
I
I
Plan
FIGURE 4.9.
Tricastin Bridge over the Rhone River, France.
FIGURE 4.10. Reallon Bridge, France.
PHASE
1
construction of central cantilever
1
2
n
PHASE 2
construction of right bank cantilever
1
3
PHASE 3
*
\
1
e
2
1
&
3
-1
j
c
closure of central and right bank cantilever
\
154
&z---4
1
2
!
il
d
f
3
n
”
5
PHASE 4
joining of right bank cantilever with abutment
PHASE 5
construction of left bank cantilever
PHASE 6
closure of left bank and
PHASE 7
joining of left bank cantilever with abutment
central
cantilever
Deck Expansion, Hinges and Continuity
155
I
Cc)
4.11. Paris Belt (Downstream). (0) Typical
construction stages. (b) Segment assembly-right bank.
(c) Segment assembly-left bank.
FIGURE
was thus obtained in terms of vertical displacement
but not insofar as rotation at the hinge point was
concerned.
Remember that in this type of structure the deck
is necessarily fixed at the various piers, which must
be designed to carry the unbalanced moments due
to unsymmetrical live-load patterns over the deck.
On the other hand, these structures are simple to
design because they are statically determinate for
all dead loads and prestressing, and the effect of
live load is simple to compute. Because there are
no moment reversals in the deck, the prestressing
tendon layout is simple.
Some disadvantages were accepted as the price
of simplicity of design:
The deck has a lower ultimate capacity as compared with a continuous structure, because there is
no possible redistribution of moments.
Hinges are difficult to design, install, and operate
satisfactorily.
There are many expansion joints, and regardless
of precautions taken in design, construction, and
operation they are always a source of difficulty and
high maintenance cost.
The major disadvantage, revealed only by experience, related to the exceeding sensitivity of such
structures to steel relaxation and concrete creep.
Because of the various hinges at midpoints of the
spans, there is no restraint to the vertical and angular displacements of the cantilever due to the effect of creep. Steel relaxation and the corresponding prestress losses tend to make matters
worse, while concrete creep is responsible for a
progressive lowering of the center of each span.
With time, there is an increasing angle break in the
deck profile at the hinge. The magnitude of the
deflection has been reported to be in excess of one
foot (0.03 m).
The difficulties experienced with this type of
construction are such that most government officials in Western Europe will no longer permit its
use.3
4.4.2 CONTINUOUS SUPERSTRUCTURES
Further research concerning the exact properties
and behavior of materials for such structures having a midspan hinge would enable more accurate
prediction of the expected deflection and thus
better control. A far more positive approach is to
eliminate the fundamental cause of the phenomenon by avoiding all permanent hinges and achieving full continuity whenever possible.
To show the relative behavior of a continuous
structure and one with hinges at midspan, a numerical application was made for the center span
of the Choisy-le-Roi Bridge in two extreme cases:
156
Design
TABLE
4.1.
of Segmental Bridges
Comparison of Crown Deflections (Hinged versus Continuous Structure)
Cast-in-Place
Hinged Structure
No. Load Stage
1
2
3
4
5
6
7
8
9
10
Girder weight
Initial prestress
Cumulative
5% Deviation of prestress
Co ntinuity p restress
Superimposed load
Finished structure (initial)
C o nc retecreep
and lo sses
Finished structure (final)
Live lo ad s
Precast
Continuous Structure
E
?
0
E
?
( lo6 psi)
(in.)
(in. X 103/ in.)
( lo6 psi)
(in.)
2.4
-2.0
0.4
5.1
5.1
5.1
6.4
6.4
1.50
-0.90
0.60
4.3
4.3
4.3
6.4
2.1
6.4
1.80
- 1.50
0.30
23%
0.30
0.60
1.10
1.70
0.90
0.4
0.8
1.4
2.2
1.1
6J
(in.
X
103/ in.)
2.0
- 1.2
0.8
7%
-0.30
0.10
0.40
-0.10
0.30
0.30
2.1
6.4
0
0
0.8
0
0.8
0
Explication of symbols:
E = modulus of elasticity for each particular loading stage
y = vertical deflection at crown
o = total angular break at crown (expressed in thousandths of inch per inch)
Derivation of results:
girder weight and initial prestress
(3) = (1) + (2)
(7) = (3) + (5) + (6) finished structure (initial stage)
finished su-ucture (final stage)
(9) = (7) + (8)
Cast-in-place cantilever with a hinge at midspan,
and
Precast
segmental
continuous
construction.
Results comparing the two structures are shown in
Table 4.1 and in Figures 4.12 through 4.14.
The study shows no significant difference between the two types of structures with respect to
the theoretical behavior of the cantilever method
under combined dead load and initial prestress,
Figure 4.12. In fact, the angle change at midspan is
even slightly less for the hinged structure, because
the p restress o f f sets a g reater p erc entag e o f
dead-load moments, 83 percent instead of 58 percent.
f
CIilCREl;t
;
ClEEP
11.5
I
ClnIInw
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONML
CISI II r ace
E
d
nim
rrecw Slrrcw
sw uciur e
FIGURE 4.13. Comparison of deflection caused by
creep (hinged versus continuous structure).
LILI
LIYE’
3
=
=
,=
I.0
-
2’
CISI II me
nln181
struclnrc
cloIIoIIII
I
rrecna
Wrrlrre
FIGURE 4.12. Comparison of deflection under dead
load and prestressing (hinged versus continuous structure).
E
=
m 11 rince
llnltd Strwri
I
Clntlnrl~:
rrecul ltrrclrre
FIGURE 4.14. Comparison of deflections caused by
live load (hinged versus continuous structure).
Deck Expansion, Hinges and Continuity
When the effect of concrete creep is considered,
however, there is a significant difference between
the tw o ty p es o f stru c tu res, Fig u re 4.13. The
hinged structure has a vertical deflection of 1.1 in.
(28 mm) and a corresponding total angle break of
0.0028 in./ inch. This value is twice that shown in
Table 4.1 and Figure 4.13 for the angle change of
one cantilever, the value of 2.8 being the total angle
break of the two abutting cantilevers. The continuous structure indicates a camber of 0.1 in. (3 mm),
and no angle break will ever appear because of full
continuity.
Further, the effect of deviation of actual prestress load from the design prestress load points
out an important difference in the sensitivity of the
two systems. Assuming the actual prestress in the
structure to differ from the design assumption by
5%, the corresponding maximum deflection is increased by 23% in the hinged structure but only
157
7% in the c o ntinu o u s stru c tu re. Theref o re, the
continuous structure is three times less sensitive to
possible deviations from the assumed material
properties.
Live-load deflections of the continuous structure
are three times more rigid than the hinged structure, Figure 4.14. The deflection of a typical span
of the Oleron Viaduct in France is compared with a
continuous span and with a crown hinged span in
Figure 4.15.
From these data it is obvious that the fullest use
o f c o ntinu ity and the elim inatio n o f hing es at
midspan w henev er p o ssib le is b enef ic ial to the
structural behavior of the bridge, to safety and
comfort of traffic, and to the structure’s aesthetic
appearance.
In practice, the continuity of the individual cantilever arms at midspan is obtained by another set
of prestressing tendons, usually called continuity
\
I
/’
\(
’
0 , 6 I. %
Y
I
2 6 0 I-L
FIGURE 4.15. Comparisons between live-load deflections for continuous or
hinged structures.
158
Design of Segmental Bridges
prestressing, which is installed along the span in a
continuous structure. Details of the design aspects
of this prestress will be discussed in Section 4.8.
4.4.3 EXPAMSIO,V
O F LOAVG
Maximum deflection under live load is reduced in
the ratio of 2.2 to 1.
Maximum angle break under live load is reduced
in the ratio of 3.0 to 1.
BRIDGES
When the continuity of the superstructure is selected as optimum for the behavior of the structure, one must keep in mind that proper measures
should be concurrently taken to allow for expansion due either to short-term and cyclic volume
changes or to long-term concrete creep.
The piers may be made flexible enough to allow
for such expansion or may be provided with elastomeric bearings to reduce the magnitude of horizontal loads to acceptable levels when applied to
the su b stru c tu re. This im p o rtant asp ec t o f the
o v erall b rid g e d esig n co ncep t is co nsid ered in
Chapter 5.
Several structures are currently made continuous in lengths of 1000 to 2000 ft (300 to 600 m)
and in exceptional cases even 3000 ft (900 m). For
longer
structures, full c o ntinuity b etw een end
abutments is not possible because of the excessive
magnitude of the horizontal movements between
superstructure and piers and related problems.
Therefore, intermediate expansion joints must be
provided. For long spans they should not be placed
at the center of the span, as in the early cantilever
bridges, but closer to the contraflexure point to
minimize the effect of a long-term deflection. Such
a concept was developed initially for the Oleron
Viaduct and is currently used on large structures
such as the Saint Clo ud Brid g e in Paris, Sallingsund Bridge in Denmark, and the Columbia
River and Zilwaukee Bridges in the United States.
Detailed computations were made in the case of
the Oleron Viaduct to optimize the location of the
expansion joint in a typical 260 ft (80 m) span, Figure 4.15 shows the shape of the deflection curve
for a uniform live loading with the three following
assumptions:
For dead-load deflections the difference is even
more significant, such that there is no substantial
difference between the actual structure and a fully
continuous one.
The variation of the angle break at the hinge
point versus the hinge location along the span
length is shown in Figure 4.16. There seems to be
little doubt that the structure is improved by selection of a proper location for the hinge and the expansion joint.
Theoretically, the ideal hinge position is between
points ,4 and B, which are the contraflexure points
f o r d ead and liv e lo ad s. Fro m a c o nstru c tio n
standpoint, such a location f-or the hinge complicates the erection process, for the hinge must be
tem p o rarilv b lo c ked and subsequentlv released
w hen the sp an is c o m p lete and continuitv is
achieved. We will consider this subject in detail
after exam ining the lay o u t o f lo ng itu d inal prestress in cantilever bridges (Section 4.8.6).
It was recently discovered, in the designing of
the Sallingsund Bridge, that the optimum location
Fully continuous span
Span with a center hinge
Span with an intermediate hinge located at 29 percent of the span length from the adjacent pier (actual case)
LOCATION OF HINGE BETWEEN
The advantages of having moved the hinge away
from the center toward the quarter-span point are
obvious:
MID- SPAN AND
PIER
FIGURE 4.16. Variation of angle break at the hinge
with hinge location along the span.
Type, Shape, and Dimensions of the Superstructure
of the hinge to control the deflections under
service-load conditions does not simultaneously
permit achievement of the overall maximum
capacity under ultimate conditions. This question
will be discussed later in this chapter.
The preceding discussion of hinge location
applies particularly for very long spans or for slender structures. For moderate spans with sufficient
girder depth it has been found that careful detailing of the prestress in the hinged span can allow
the hinge to be maintained at the centerpoint for
simplicity (spans less than 200 ft with a depth to
span ratio of approximately 20). Such was the case
for the cantilever alternatives of the Long Key and
Seven Mile Bridges in Florida.
4.5
Type, Shape, and Dimensions of
the Superstructure
4.5.1 BOX SECTIONS
The typical section best suited for cantilever construction is the box section, for the following reasons:
1. Because of the construction method, deadload moments produce compression stresses at the
bottom fiber along the entire span length, and
maximum moments occur near the piers. The
typical section therefore must be provided with a
large bottom flange, particularly near the piers,
and this is achieved best with a box section.
The efficiency of the box section is very good,
and for a given amount of concrete provides the
(0)
Longitudinal section
FIGURE 4.17.
159
least amount of prestressing steel. The efficiency of
a section is usually measured by the following dimensionless coefficient:
r2zyxwvutsrqponmlkjihgfedcbaZYXWVUTSR
p=C&2
with the notations as given in Figures 4.17 and
4.18, where some basic formulas are presented.
The efficiency would be p = 1 if the concrete
were concentrated in thin flanges with webs of
negligible thickness. On the other hand, a rectangular section has an efficiency of only l/3. The
usual box section efficiency is p = 0.60, which is
significantly better than that of an I girder.
2. Another advantage of the large bottom
flange is that the concrete area is sufficiently large
at ultimate load to balance the full capacity of the
prestressing tendons without loss in the magnitude
of the lever arm. ’
3. The elastic stability of the structure is excellent both during construction and under service
conditions, because the closed box section has a
large torsional rigidity.
4. In wide bridge decks where several girders
must be used side by side, the large torsional stiffness of the individual box girders allows a very
satisfactory transverse distribution of live loads
without intermediate diaphragms between piers.
5. Because of their torsional rigidity, box
girders lend themselves to the construction of
curved bridge
and provide
superstructures
maximum flexibility for complicated tendon trajectories.
6)
Typical tramverSe
section
Typical characteristics of a box section: Total section height: h; crosssection area: A; moment of inertia: I; position of centroid; c,, c2; radius of gyration: r
given by rp = Z/A; efficiency ratio: p = r%,c,; limits of central core: r*/c, = PC,; r%, = pc2;
for the usual box girder: p = 0.60.
160
Design of Segmental Bridges
dl
px
,c2/ I
h
- F
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCB
Cl
Y
(a)
h
Ph
-
I
Cc)
FIGURE 4.18. ‘rypical prestress requirements of a
box girder. (u) For maximum negative moment over the
pier (LX + LL): total moment = M; required prestress =
F = M/z with z = c, - cf, + cp; usually over the piel- z =
0.75 12. (b) For maximum positive moment at midspan
(LX + IL): total moment = ‘M; required prestress = F =
M/i with z = cp - cf2 + c ,; usually at midspan z = 0.70h. (c)
For variable moments (LL): total moment variation =
A M (sum of positive and negative L.C. moments); required prestress = F = hM/ph (p = 0.60).
zyxwvuts
(b)
The optimum selection of the proportions of the
box section is generally a matter of experience. A
careful review of existing bridges provides an excellent basis for preliminary design. The various
parameters that should be considered at the start
of a design are:
Constant versus variable depth
Span-to-depth
ratio
Number of parallel box girders
Shape and dimensions of each box girder, including number of webs, vertical or inclined webs,
thickness of webs, top and bottom flanges
All these factors are closely related to each other,
and they also depend largely upon the constructio n req u irem ents- f o r exam p le, the siz e o f the
p ro jec t that w ill req u ire a larg e inv estm ent in
sophisticated casting equipment.
4.5.2 SHAPE OF SUPERSTRUCTURE IiY
ELEVz4TlOh’
Constant depth is the easiest choice and affords the
best solution for short and moderate spans, up to
200 ft (60 m). However, constant depths have been
used for aesthetic reasons for spans to 450 ft (140
m), such as the Saint Cloud Bridge in Paris and the
161
Type, Shape, and Dimensions of the Superstructure
Pine Valley and Columbia River Bridges in the
United States, Figure 4.19~.
When the span increases, the magnitude of
dead-load moments near the piers normally requires a variation of structural height and a curved
intrados. When clearance requirements allow, a
circular intrados is the easier and more aesthetically pleasing choice, although in some cases (such
as the Houston Ship Channel Bridge) a more complex profile must adjust to the critical corners of
the clearance diagram. Between the constantdepth and the curved-intrados solutions, Figure
4.19, intermediate options may be used, such as:
Increase thickness
at pier ,
Ii?!’
A ,_ ,,,.,,
The semiconstant depth, where the concrete required in the bottom flange near the piers is placed
outside the typical section rather than inside the
box (constant dimension for the interior cell). This
solution has been used on two bridges in France
and is aesthetically satisfactory, Figure 4.196.
Straight haunches (bridge for the Ring Road in
Paris). In this case caution must be exercised to insure compatibility of the local stresses induced by
the abrupt angle change of the bottom soffit at the
start of the haunch, where a full diaphragm is usually needed inside the box, Figure 4.19~.
_.
l/15< h<1/30
optimum l/18 to l/20
,,
-Yw“’
.““.
-zyxwvutsrqponmlkjihgfedcbaZ
”
_.,
1116
<h,lL
< l/20
optimum
1118
1/16<h,lL<1/20
o p t i m u m l/l8
II22 <hr,lL < l/ 28
1/30<holL
< l/50
I
Circular intrados or
third-degree parabola
Cd)
FIGURE 4.19 Longitudinal profile for segmental bridges. (k) Constant depth.
(b) Semiconstant depth. (c) Straight haunches. (d) Variable depth.
Design of Segmental Bridges
162
4.53
CHOICE OF TYPICAL CROSS SECTION
Web spacing is usually selected betw een 15 and 25
ft (4.5 and 7.5 m) to reduce the number of webs to
a minimum, simplifying construction problems
while keeping transverse bending moment in the
top and bottom flanges within reasonable limits.
A superstructure up to 40 ft (12 m) in width is
thus normally made up of a single cell box girder
with two lateral cantilevers, the span of which is
slightly less than one-fourth the total width (7 to 8
ft for a 40 ft width).
For wide bridges, multicell box girders may be
used:
Three webs, two cells: as in the B-3 South Viaduct
and the Deventer Bridge
Four webs, three cells: as in the Saint Cloud Bridge
and the Columbia River Bridge
Alternatively, large lateral cantilevers and a large
span length between webs are accepted with special
provisions to carry the deck live loads transversely:
Transverse flange stiffeners as in the Saint Andre
de Cubzac, Vejle Fjord, and Zilwaukee Bridges
10 in. (250 mm) when small ducts for either vertical or longitudinal post-tensioning tendons occur
in the web
12 in. (300 mm) when ducts for tendons (twelve 3
in. diameter strands) occur in the web
14 in. (350 mm) when an anchor for a tendon
(twelve 4 in. diameter strands) is anchored in the
web proper
Most codes underestimate the capacity of twoway slabs, such as the roadway slab or top flange of
a box girder bridge, whether prestressed transversely or mild-steel reinforced. There is a great
reserve of strength due to the frame action between slabs and webs in the transverse direction.
The minimum slab thickness to prevent punching shear under a concentrated w heel load is approximately 6 in. (150 mm). However, it is recommended that a slab thickness of not less than 7 in.
(175 mm) be used to allow enough flexibility in the
layout of the reinforcing steel and prestressing
ducts and obtain an adequate concrete cover over
the steel and ducts.
Recommended minimum top flange thickness
versus the actual span length between webs should
be:
Side boxes as in the Chillon Viaduct
Alternatively several boxes may be used side by
side to make up the superstructure. Figures 4.20
through 4.24 give the dimensions of a few structures selected at random from various countries
throughout the world.
4.5.4 DIM ENSIONS OF THE TYPICAL
C R O SS SEC TI O N
Three conditions must be considered in determining the web thickness:
Shear stresses due to shear load and torsional moments must be kept within allowable limits
Concrete must be properly placed, particularly
where draped tendons occur in the web
Tendon anchors, when located in the web, must
distribute properly the high prestress load concentrated at the anchorages
Following are some guidelines for minimum web
thicknesses:
8 in. (200 mm) when no prestress ducts are located
in the web
Span less than 10 ft (3 m)
7 in. (175 mm)
Span betw een 10 and 15 ft
(3 to 4.5 m)
8 in. (200 mm)
10 in. (250 mm)
Span betw een 15 and 25 ft
(4.5 to 7.5 m)
Over 25 ft (7.5 m), it is usually more economical to
substitute a system of ribs or a voided slab for a
so lid slab.
Early bridges used very thin bottom flanges in
order to reduce critical weight and dead-load moments. A 5 in. (125 mm) thickness was used in
bridges, such as the Koblenz Bridge in Germany. It
is very difficult to prevent cracking of such thin
slabs due to the combined effect of dead load carried between webs and longitudinal shear between
web and bottom flange. For this reason, it is now
recommended that a minimum thickness of 7 in.
FIGURE 4.20. Typical dimensions of some cast-inplace segmental cantilever bridges in France. Year of
construction and maximum span length (ft): (a) Moulin
a Poudre (1963), 269. (6) Morlaix (1973), 269. (c) Bordeaux St. Jean (1965), 253. (d) Givors (1967), 360. (e)
Oissel (1970), 328
(fl)
-I--=-+
(b)
(4
t
s,
zyxw
(e)
163
164
Design of Segmental Bridges
FIGURE 4.20 (Continzx~) (f) Viosne (1972), 197. (g) J o i n v i l l e (twin deck) (1976), 354. (h)
Gennevilliers (1976), 564.
(175 mm) be used, regardless of the stress requirements. Where longitudinal ducts for prestress
are distributed in the bottom flange, a minimum
thickness of 8 to 10 in. (200 to 250 mm) is usually
necessary, depending on the duct size.
Near the piers, the bottom slab thickness is progressively increased to resist the compressive
stresses due to longitudinal bending. In the Bendorf Bridge, 680 ft (207 m) span, the bottom
flange thickness is 8 ft (2.4 m) at the main piers
and is heavily reinforced to keep the compressive
stresses w ithin allo w able limits.
After this brief review of the various conceptual
choices for dimensioning the deck members, con-
sideration should be given to the design of such
members with particular emphasis on the following points:
Distribution of load between box girders in multibox girder bridges
Effect of temperature gradients in the structure
4.6
Transverse Distribution of Loads Between Box
Girders in Multibox Girders
We noted earlier that wide decks can conveniently
consist of two or even three separate boxes trans-
3 4.60
0
(1))
10.92
I
1
1
’
I
!
-t
5.50 g I
10.60
’
10.60
1
7
,
FIGURE 4.21. Typical dimensions of some precast segmental cantilever bridges in
France. Year of construction and maximum span length (ft): (a) Choisy-le-Roi (1965),
180; (b) Courbevoie (1967), 197; (c) Oleron Viaduct (1966), 260; (d) Seudre (1971), 260;
(e) B-3 South Viaduct (1973), 157; cf) St. Andre de Cubzac (1974), 312; (g) St. Cloud
(1974), 334; (h) Ottmarsheim (1976), 564.
165
%I
900
,96,
%1
1
9cKl
196
1
(4
c_- 9 50
(e)
t
I
c
zyxwvutsrqp
.-
-
zyxwvutsrqpo
(h)
166
Cc)
zyxw
(4
(ft): (a) Koblenz, Germany (1954), cast in place,
374; (b) Bendorf, Germany (1964), cast in
place, 682; (c) Chillon, Switzerland (1970), precast, 341; (d) Sallingsund, Denmark (1978),
.precast, 305; (e) Vejle Fjord, Denmark (1979),
cast in place, 361.
167
(b)
(c)
zyxwvutsrqpo
FIGURE 4.23. Typical dimensions of some segmental cantilever bridges in Europe.
Year of construction and maximum span length (ft): (a) Felsenau, Switzerland (1978),
cast in place, 512; (6) Tarento, Italy (1977), cast in place, 500; (c) Kochertal, Germany
(1979), cast in place, 453.
168
Transverse Distribution of Loads Between Box Girders in Multibox
Typical Cross Section
!22'-6"!
!22'-;q6,
t
4
,
r
20'4
1
r
3a'-6"
t
4
,
17'-5"
,
'
38'-6"
T
4
t
,
T
36'
59'-3"
1
r
m
4
I
1
381
1
T
Typical dimensions of some segmental
cantilever bridges in the Americas. Year of construction
and maximum span length (ft): (n) Rio Niteroi, Brazil
(1971), precast, 262: (h) Pine Valley, U.S.A. (1974), cast
in place, 450; (c) Kipapa, U.S.A. (1977). cast in place,
250; (n) Kishwaukee,
U.S.A., precast, 250; (e) Long Key,
U.S.A., precast, 118;(r) Seven Mile, U.S.A., precast, 135;
(y) Columbia River, U.S.A., cast in place and precast,
600: (h) Zilwaukee, U.S.A., precast, 375; (i) Houston
Ship Channel, U.S.A., cast in place, 750.
FIGURE 4.24.
versely connected by the top flange. A detailed
analysis was made of such decks with regard to the
distribution of live load between the various boxes.
It was found that in normal structures of this type,
the combined effect of the flexural rigidity of the
roadway slab acting transversely as a rigid frame
with the webs and bottom slab of the various box
Girders
169
girders, on one hand, and the torsional rigidity of
such box girders on the other hand, would result in
a very satisfactory transverse distribution of live
loads between box girders. There is no need for
diaphragms between girders as normally provided
for I-girder bridgers.
Comprehensive programs of load testing of several bridges, including accurate measurements of
deflections for eccentric loading, fully confirmed
the results of theoretical analysis. This analysis has
been reported in various technical documents, and
only selected results will be presented in this section.
The first bridge analyzed in this respect was the
Choisy-le-Roi Bridge. A knife-edge load P is considered with a uniform longitudinal distribution
along the span, Figure 4.25. When this load travels
crosswise from curb to curb, each position may be
analyzed with respect to the proportion of vertical
load carried by each box girder, together with the
corresponding torsional moment and transverse
moment in the deck slab. These analyses have
made it possible to draw transverse influence lines
for each effect considered, such as longitudinal
bending m o m e n t s ( o v e r t h e s u p p o r t o r a t
midspan), torsional moments, or transverse moments.
For longitudinal moments it is convenient to use
a dimensionless coefficient, Figure 4.25c, which
represents the increase or decrease of the load carried by one box girder in comparison with the
average load, assuming an even distribution between both girders. Numerical results show that
the transverse distribution of a knife-edge load
placed on one side (next to the curb) of a twin box
girder produces bending moments in each box that
are 1.4 and 0.6 times the average bending moment.
For the same configuration, a typical deck with I
girders would have an eccentricity coefficient of
approximately 4 compared with 1.4 for the box
girders. There are, however, two side effects to
such an encouraging behavior, which relate to torsion stresses and transverse bending of the deck
slab.
Torsional M oments in the Box Girder An unsymmetrical distribution of live loads in the transverse
direction tends to warp the box girders and cause
shear stresses. It is their high torsional rigidity
which produces a favorable distribution of loads
between girders. However, the maximum torsional
moments usually occur when only one-half the
structure (in cross section) is loaded, and the resulting stresses do not cumulate with the shear
stresses produced by the full live-load shear force.
Design of Segmental Briees
170
Span length,
L
P
(A)
4
I
41
I
(z
4
I +
ho (at midspan)
Center of span
h, (over support)
(
2d
c---h
*
2d’
4
67)
FIGURE 4.25. Principle of transverse distribution of loads between box
girders. (a) Dimensions. (b) Influence line of the shear in the connecting slab.
(c) Transverse influence line of longitudinal moment. (d) Transverse bending influence line at section A.
Transuer M oments in the Deck Slab The deck slab
cannot be considered as a continuous beam on
fixed supports because of the relative displacements on the two boxes due to unsymmetrical
loading. Figure 4.25d shows the consequence. If
the slab w ere resting o n f ixed su p p o rts, the
influence line for the moment in a section such as
(A) would be the typical line (1). Because the box
girders undergo certain deflections and rotations,
the effect is to superimpose the ordinates of another line such as (2).
Numerically, the difference is not as great as
may be expected at first sight, because line (1) pertains to the effect of local concentrated truck loads
w hile line (2), b eing the resu lt o f d ifferential
movements between box girders, pertains to the
effect of uniformly distributed loads. In summary,
deck moments are increased by only 20 to 30%
over their normal values if flexibility of the box
girders is ignored. As a matter of practical interest,
actual num erical v alues fo r sev eral b rid g es in
France with either two or three box girders that
have all shown excellent performance for more
than 10 years are presented in Figures 4.26 and
4.27.
4.7
Effect of Temperature Gradients
Bridge Superstructures
in
Experience has shown the sensitivity of long-span
cantilever bridges to concrete creep. This resulted
in the p referenc e f o r c o ntinu o u s rather than
hinged cantilevers. How ever, tw o more problems
arose from this significant change in design approach, both being the immediate result of continuity. These problems are (1) effect of temperature gradient in bridge decks and (2) redistribution
Effect of Temperature Gradients in Bridge Superstructures
Spans
Bridge
2d
2d'
(ft)
(ft)
29.5
6.6/18.0
13.1
26.2
11.1/18.0
15.4
33.9
9.2115.7
14.1
23.3
5.9114.'
14.6
24.9
5.2/10.7
0
Givors
7360' 1 ,- (1lOiy
15.'
0
D/S Paris
Ring Parkway
-300'by
0
U/S Paris
Ring Park-
'
I
(90m)
7
295'
r
Eccen.
Coeff.
ho/h1
way
co
Corde
0
l(79m)T
5
260'
T
,
220'
yp
Juvisy
@
Choisy-leRoi
\
y
180'
7
1
(55m)
1.23
t
__- -.11.1
22.3
1
co;;fant
1 1.28
i
zyxwvuts
?d' I
tzyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
2d
t
I
:
1
FIGURE 4.26. Transverse distribution of loads between box girders, numerical values
for several two-box girders.
of internal stresses due to long-term effects (steel
relaxation and concrete creep). The importance of
these two new problems was discovered experimentally. All structures are designed, according to
the provisions of the various codes, for changes of
temperature that are assumed to apply to the entire section. Significant bending moments in the
superstructure occur only as a result of the frame
action with the piers where a rigid connection is
achieved between sub- and superstructure. Actual
measurements on existing structures confirm this
assumption. The average concrete section undergoes a progressive shortening due to shrinkage
and concrete creep superimposed naturally with
the usual seasonal temperature variations, Figure
4.286. The to tal c o nc rete strain o f 120 X 10m6
in./ in. w as v ery m o d erate f o r a p erio d o f f o u r
years.
Daily readings, on the same bridge, of strains
and magnitude of reactions over the abutment
172
Design of Segmental Bridges
I
1
2
Calculated deflection
(E = 6.4 X lo6 psi)
I
Measured deflection
Measured deflection
Calculated deflection
(E = 6.9 X lo6 psi)
0
,
I
I
1
I
II
/I
I
---
,
4.27.
1
deflection
--
I-
FIGURE
..~
Measurec
Calculated deflection
(E = 7.4 X lo6 psi)
Transverse distribution of loads between box girders.
brought to light a factor that had previously been
ignored. This was the differential exposure of the
bridge deck to the sun on warm summer days. This
situation is aggravated for bridges crossing a river,
where the bottom flange is kept cool by the water
and the usual black pavement placed over the top
flange concentrates the sun’s radiation. Within a
24-hour period the reaction over the abutment
c o uld v ary as m uc h as 26%, Fig u re 4.28~. The
equivalent
temperature difference between top
and b o tto m f lang es reac hed 18° F ( 10° C ) . The
maximum stress at the bottom flange level, due
Design
of Longitudinal Members for Flexure and Tendon Profile
173
The effect is usually computed by assuming the
gradient to be constant throughout the bridge
superstructure length, which is not necessarily the
case.
Figure 4.29 shows the result for the case of a
typical span built-in at both ends (this is the case of
a long structure with many identical spans). The
stress at the bottom fiber depends only upon the
following two factors:
Variation of height between span center and support (ratio hi/ h,)
Position of the center of gravity within the section
(ratio c,lh,)
The lowest stress is obtained for a symmetrical section and a constant-depth girder.
The stress increases rapidly when the variation
in depth is more pronounced. For normal proportions the effect of gradient is increased by 50% in
v ariab le- d ep th g ird ers c o m p ared to constantdepth girders (240 psi versus 160 psi for a 9°F gradient and a modulus of 5 x lo6 psi).
4.8
Design of Longitudinal Members for Flexure
and Tendon Pro$lezyxwvutsrqponmlkjihgfedcbaZY
Cc)
FIGURE 4.28. Champigny Bridge, observed values of
concrete strains and deck reactions. (a) Typical dimensions. (b) Long-term shortening of bridge deck due to
concrete creep superimposed with temperature variations. (c) Daily temperature variations as exemplified by
change in reactions over abutments.
only to this temperature gradient, reached 560 psi
(3.9 MPa), a value completely ignored in the design
assumptions.
Various countries of Western Europe have now
incorporated special provisions on temperature
gradients as a result of this knowledge. In France,
the following assumptions are required:
1. Add the effect of a 18°F (10°C) temperature
gradient to the effect of dead loads and normal
volume changes (such as shrinkage, creep, and
maximal temperature differences). The effect
of gradient is computed with an instantaneous
modulus of elasticity (usually 5 million psi).
2.
Add the effect of a 9°F (5°C) temperature gradient to the combined effect of all loads (includ ing liv e lo ad and im p act) and v o lum e
changes, again using an instantaneous modulus of elasticity.
4.8.1 PRINCIPLE OF PRESTRESS LAYOUT
The longitudinal prestress of a cantilever bridge,
whether cast in place or precast, consists of two
families of tendons:
1.
As construction in cantilever proceeds, the increasing dead-load moments are resisted at
each step of construction by tendons located in
the top flange of the girder and symmetrically
placed on either side of the pier, Figures 4.30
and 4.31~. These are known as cantilever tendons.
2. Up o n co m p letio n o f ind iv id ual cantilev ers,
continuity is achieved by a second family of
tendons essentially placed at the center of the
various spans, Fig u re 4.316. Bec au se g ird er
lo ad m o m ents are sm all, exc ep t thro u g h
long-term redistribution, because of the construction procedure, the continuity prestress is
designed to resist essentially the effect of:
a. Su p erim p o sed lo ad s ( p av em ent, c u rb s,
and the like).
b . Liv e lo ad s.
C.
Temperature
gradient.
Design of Segmental Bridges
174
I
I
2.0
1.5
ELEVATION
OF
SPAN
2.5
3.0
ho
-L
SECTION AT CENTER
FIGURE 4.29. Effect of thermal gradient on box girder decks.
d. Subsequent redistribution of girder load
and cantilever prestress.
Tensile stresses are large at the bottom flange
level, but seldom will continuity prestress gain the
full advantage of the available eccentricity because
of the stress conditions at the top flange level. Usually this prestress is divided into tendons, B 1 or B2,
located in the bottom flange, and a few tendons
such as B3 which overlap the longer cantilever
tendons, Figure 4.3 lb.
For the best selection of prestressing methods, it
is essential to use prestressing units of a capacity
large enough to reduce the number of tendons in
the concrete section, particularly in very long
sp ans. O n the o ther hand , there m u st b e a
sufficient number of tendons to match with the
number of segments in the cantilever arms. Also,
units w ith an excessive unit capacity w ill pose serious problems for the transfer of concentrated high
loads, particularly for cast-in-place structures,
where concrete strength at the time of prestress is
always a critical factor within the construction
cycle.
In practical terms, prestress bars are as well
adapted to short and medium spans as strand ten-
dons (such as twelve 3 in. diameter strands). For
very long spans (above 500 ft) large-capacity tendons (such as nineteen 0.6 in. diameter strands)
with a final prestress force of about 700 kips afford
a very practical solution for cantilever prestress.
For continuity prestress the size of tendons is governed by the possibility of locating the tendon anchors in such areas and w ith such provisions as to
allow a proper distribution of the concentrated
load to the surrounding concrete section. Units
such as twelve 3 in. diameter or twelve 0.6 in.
diameter are usually well adapted with careful detailing for this purpose.
4.8.2
DRAPED
TENDONS
In early applications, both families of prestress
were given a draped profile in the web of the box
section to take advantage of the vertical component
of prestress to reduce the shear stresses. In such a
configuration there is a considerable overlapping
of tendons in the web, because the cantilever prestress is anchored in the lower part of the web and
the continuity prestress is anchored at the top
flange level; see the layout in Figure 4.31~. Fo r a
constant-depth section and for segments of equal
Diagrams of moments in a cantilever
4.30. Typical cantilever moments and prestress. When placing
unit 8, the increase of bending moment is represented by the hatched area
and the resultant curve is transferred from position 7 to position 8. Additional sets of cantilever prestressing tendons are placed each time a pair of
segments is erected. This procedure allows the magnitude of prestress to
follow very closely the various steps of construction.
FIGURE
length, it is easy to completely standardize the layout of prestress in various segments.
Mechanization of the casting operations is a very
desirable feature, all prefabricated reinforcing
cages being identical, with ducts always at the same
locations. A substantial amount of repetition may
still be obtained in variable-depth members as seen
in Figure 4.32, which represents a typical span of
the Oleron Viaduct. The two disadvantages of such
a prestress layout are:
Cantilever tendon anchors are located in the web
and it is difficult to prevent web cracking, particularly in cast-in-place structures, except through the
use of thicker webs and smaller tendons.
Continuity tendons extend above deck level at both
ends. The installation of the anchor w ith the
block-out for stressing is difficult in the casting
form, and good protection against water seepage
to the tendons in the finished structure is a critical
factor.
4.83 STRAIGHT TENDONS
Tendons are in this configuration located in the
upper and lower flange of the box girder and anchored near the web in their respective flanges.
There is no draped profile for the tendons within
the web and consequently no reduction of shear
stresses due to a vertical component of prestress.
This is a disadvantage of this scheme, w hich may
often require vertical prestress to maintain shear
stresses within allowable limits. On the other hand,
the two advantages are:
Simplicity in both design and construction
Design of Segmental Bridges
176
(AI )
span
Average
length
of
L
cantilever
tendons
0.52 L
I
fd
Average
length
of
contlnuit?
tendons
: 0 35 - 0.50 L
(b)
.4 : cantilever tendons
B : continuity tendons
0
A
Q
A
Typical layout of longitudinal prestress. (a) Cantilever tendons. (b)
Continuity tendons. (c) Standardized layout of tendons for constant-depth segments.
FIGURE 4.3 1.
Significant reduction in friction losses of the prestress tendons for both curvature and wobble effects, and consequent savings on the weight and
cost of the longitudinal prestress of at least lo%, all
else being equal
The Rio N itero i Brid g e (d esc rib ed in Sec tio n
3.8) used straight tendons, Figure 4.33. Typical
characteristics of the deck are as follows:
Span
length
262 ft
Width of a box
42 ft
Two webs at
14.2 in. each
Longitudinal
prestress
cantilever
42 (12 4 in. diam.
strands)
Longitudinal
prestress
continuity
14 (12 f in. diam.
strands)
Vertical
prestress
1 in. diameter bars
Cantilever prastress
30412 x 1.2” 6)
+8-(12x .315”9)
15
I<
13
I?
1’1 ,o
Continuity prestress--14-(12x l/2” @I
+ 4(12 x .315” $1
Detail B
Detail A
Transverse
prestress
r
Longitudinal
FIGURE 4.32. Oleron
prestress
Viaduct, longitudinal prestress.
distribution
Design of Segmental Bridges
178
Vertical bars 25 mm 9 (typ. )
TOP
PRESTRESS
12 strand 12.7 mm # cablas
FIGURE 4.33. Rio-Niteroi Bridge, typical prestress layout.
Critical stresses near the pier are:
Longitudinal compression
850 psi
400 psi
Vertical compression
Maximum shear stress
Diago nal stresses
580 psi
- 110 psi (tensile),
an d
1360 psi
(compressive)
Typical details of tendon profiles and anchorages are portrayed for Linn Cove Viaduct in North
Carolina, U.S.A., in Figures 4.34, 4.35, and 4.36.
4.8.4
SUM M ARY OF TENDON PROFILES AND
A N C H O R L O C A TI O N S
In the two preceding configurations, tendons were
anchored in the following manner:
1. For cantilever prestress:
a. On the face of the segment in the fillet
between top flange and web.
b.
c.
On the face of the segment along the web.
In a block-out near the fillet between top
flange and web, but inside the box.
2. For the continuity prestress:
a . At the top flange level.
b.
In a block-out near the fillet between web
and bottom flange.
c.
In a block-out in the bottom flange proper
away from the webs.
Configurations lc, 2b, and 2c all permit prestressing operations to be performed safely and
efficiently inside the box, Figure 4.37, permitting
such operations to be removed from the critical
path of actual placement or construction of the
seg m ents. O nly tho se tend o ns req uired fo r
balancing the self-weight of the segments need to
be installed at each step of construction. The balance of the required prestressing may thus be installed later, even after continuity is achieved between several cantilever arms. Tendons for the
additional prestress may then be given a profile
comparable to that used in cast-in-place bridges
with a length extending over several spans. The
practical limit to this procedure is excessive sophistication and related high friction losses in the
tendons.
4.8.5 SPECIAL PROBLEM S OF CONTINUITY
PRESTRESS AND ANCHORAGE THEREOF
Tendons for continuity prestress may not, or even
should not, always be located in the fillet between
web and bottom flange. They may be located in the
bottom flange proper. When a variable-depth
member is used, the bottom flange has a curvature
in the vertical plane, which must be followed by the
prestress tendons. Unless careful consideration is
Design
of Longitudinal Members for Flexure and Tendon Profile
5 s&u e 7 “: Z-‘/1 .’
\
c 8’ .I. e ; ”
FIGURE 4.34.
3
SW 0
7”,1!4’
Linn Cove Viaduct, typical cross section showing prestress ducts.
given to that fact at the concept and detailed design
stages, difficulties are likely to develop; we may see
this by looking at Figures 4.38 and 4.39, which
show the free-body diagrams of stresses in the
bottom flange due to the curvature, together with a
numerical example. Curvature of a tendon induces
a downward radial load, which must be resisted by
transverse bending of the bottom flange between
the webs.
Longitudinal compressive stresses in the bottom
flange similarly induce an upward radial reaction
in the flange, counteracting at least in part the effect of the tendons. Unfortunately, when the full
live load and variable effects, such as thermal gradients, are applied to the superstructure, the lon-
gitudinal stresses vanish and consequently the partial negation of the effect of tendon curvature is
lost. Therefore, the effect of tendon curvature
adds fully to the dead-load stresses of the concrete
flange. The corresponding flexural stresses are
four to five times greater than the effect due to
dead load only, and if sufficient reinforcement is
not provided for this effect, heavy cracking is to be
expected and possibly failure. Practically, the situation may be aggravated by deviations in the location of the tendon ducts in the segments compared
to the theoretical profile indicated on the drawings.
At the point between segments, ducts are usually
placed at their proper position; but if flexible tubing is used with an insufficient number of sup-
F
ANCHORAGE
zyxwvuts
/
A
.?:o*
I
I
i
zyxwvutsrqpon
__------------
t
FIGURE 4.35.
180
Linn Cove Viaduct, top flange prestress details.
:‘s
HOLES FOR TEMPORARY PRESTRESSING 1’1)‘8
~-~--~
BARS
D U C T 3 ‘/I$’ 0
I
.--__~__--.-.--p.p.
FIGURE 4.36.
Linn Cove Viaduct, bottom flange prestress details.
181
182
Design of Segmental Bridges
PARTIAL
CROSS-SECTION
FIGURE 4.37. B-3 South Viaduct, prestressing operations in box girder.
porting chairs or ties, the duct profile will have an
angle break at each joint. In addition to the increased friction losses, there is a potential danger
of local spalling and bursting of the intrados of the
bottom flange, Figure 4.40. Rigid ducts properly
secured to the reinforcement cage and placed at
the proper level over the soffit of the casting
machine or traveler will avoid this danger.
Another item concerning potential difficulties in
continuity prestress relate to the projection of the
anchor block-out in the bottom flange and where
anchor blocks are not close to the fillet between
web and bottom flange. When this method is used
in conjunction with a very thin bottom flange (a
Assumed
COMPRESSIVE
STRESSES
FREE BOOY DIAGRAM
FIGURE 4.38. Secondary stresses due to curved tendons in the bottom flange.
flange as thin as 5 or 6 in. has been used in early
bridges), it is almost impossible to distribute the
concentrated load of the anchor block in the slab
without subsequent cracking. For a 7 or 8 in. flange
it is recommended that no more than two anchor
Longitudinal
Radius
(12 x l/2"+
1,000'
tendons) Typ.
a. = 15.67'
lo'-0"
4
I_
lo'-0"
FIGURE 4.39. Secondary stresses due to curved prestressing tendons, numerical example. Assumed longitudinal radius = 1000 ft. Weight of bottom slab
= 100 psf. Effect of compressive stresses: unloaded bridge,f, = 2000 psi, compressive radial load: f,tlR = (2000 x 8 x 12)/1000 = 200psf; loaded bridge, 0
psi. Effect of prestressing tendons: stranded tendons (twelve f in. dia strands) at
10 in. interval with a 280 kip capacity, corresponding radial load: F/R =
280,000/[( 10/12)1000] = 336, say 340 psf. Total loads on bottom slab: (1) during
construction, load = 100 psf; (2) unloaded bridge, load = 100 - 200 + 340 =
240 psf; (3) loaded bridge, load = 100 + 340 = 440 psf, moment = we2112 = 9
kips ft/ft, stress in bottom slab: f = M/S = (9000 x 12)/[( 12 x 64)/6] = 840 psi.
Design of Longitudinal Members for Flexure and Tendon Profile
tendons may be made continuous through the
expansion joint or equipped with couplers.
Pa ma l e le va tio n
d. Resume normal cantilever segment placing
and prestressing to the center of the span, with
tendons crossing the joint.
e. Achieve continuity with previous cantilever by
pouring closure joint and stressing continuity
tendons. Layout of these tendons includes anchors in the special hinge segment to transfer
the shear forces in the completed structure.
f.
FIGURE 4.40. Ef‘fect
of’ misalignment of’ continuity
prestress.
blocks for (I2 f in. diameter strands) tendons be
placed in the same transverse section in conjunction with additional reinforcing to resist bursting
stresses. Wherever possible, the anchor blocks for
continuity tendons should be placed in the fillet
between the web and flange where the transverse
section has the largest rigidity.
4.8.6 L,-iYOI’T O F PRESTRESS I.Y STRUCTURES
IZ’ITH HI.Z’GES ,4SD EXP,4.\‘SION
JOINTS
Section 4.4.3 explained how the expansion joints in
the superstructure should be located preferably
near the contraflexure point of a span rather than
at midspan as in previous structures. However,
there is a resultant complication in the construction
process, because cantilever erection must proceed
through the special hinge segment. A typical construction procedure and the related prestress layout are presented in Figure 4.4 1. For the geometry
of the structure in this figure, the construction
proceeds as follows:
a.
b.
C.
Place the first five segments in balanced cantilever and install cantilever prestress for resistance against dead load.
Place the lower half of the special segment and
the corresponding tendons.
Install the upper half of the special hinge segment with permanent, or provisional bearings,
and provisional blocking to permit transfer of
longitudinal compressive stresses. Cantilever
Remove temporary blocking at hinge. Release
tension in cantilever tendons holding segments
7, 8, and 9 or cut tendons across the hinge
after grouting.
4.8.7 REDISTRIBUTION OF MOME,VTS
AND
STRESSES THROUGH CONCRETE CREEP
In a statically indeterminate structure the internal
stresses induced by, the external loads depend
upon the deformation of the structure. In prestressed concrete structures such deformations
must include not only short-term but also longterm deformation due to relaxation of prestressing
steel and concrete creep. In conventional structures such as cast-in-place continuous superstructures, the effect is not significant if all loads and
prestress forces are applied to the statical design of
the completed structure, which is the common case
of construction on scaffolding. The behavior of
cantilever bridges, particularly cast-in-place structures, is quite different, because the major part of
the load (the girder load often represents 80% of
the total load in long spans) is applied to a statical
concept that is different from the completed design. As soon as continuity is achieved, the structure tends to resist the new situation in which it has
been placed; this is one aspect of a very general law
in mechanics whereby consequences always oppose
their cause.
A very simple example is presented in Figure
4.42, which will provide the basis for a better appreciation of the problem. Assume two identical
adjacent cantilever arms built-in at both ends and
free to deflect at the center. The self-weight produces a moment:
at both ends with a corresponding deflection and
rotation at the center of y and o.
184
Design of Segmental Bridges
I
Cantilever &$qqq
5
3
4
2
tendons
1111
67)
Tendon
Cantilever tendons
for construction
FIGURE 4.41. Construction procedure and prestress in a span with an
expansion joint.
If the load is applied for a short time, the value
of E to take into account is Ei (instantaneous modulus). Assuming that continuity is achieved between the cantilevers as shown in Figure 4.42c,
there cannot be an angle break at the center, but
only a progressive deformation of the completed
span. After a long time the concrete modulus has
changed from its initial value Ei to a final value E,,
which may be approximately 2.5 times less than Ei .
Because the external loads are unchanged and
the structure is symmetrical, the only change in (he
state of the structure is an additional constant moment M, developing along the entire span and increasing progressively with time until the concrete
creep has stabilized. At all times the magnitude of
this moment adjusts in the structure to maintain
the assumed continuity at the center.
The additional deflection at midspan,
y2, takes
place in a beam with fixed ends under the effect of
its own weight and only because of the progressive
change of the concrete modulus from the value Ei
to the value E,.
Considering the concrete strain at any point of
the structure, the total strain q is the sum of two
terms:
Ef = E, + Ep
where cr = strain before continuity is achieved,
E2 = strain after continuity is achieved.
Hooke’s law relating stress and strain at a particular point in time states:
E, =g
1
Design of Longitudinal Members for Flexure and Tendon Profile
zyxwv
185
In other words, the effect of concrete creep is to
place the final stresses in the structure in an internal state (either of moments, shear forces, deflections, or stresses) intermediate between:
The initial statical design with free cantilevers, and
The completed design with continuity.
Assume, for example, EfIEi
= 0.40. Thus:
f= 0.4Of, + O.SOf,
The relationship is equally true for moments,
shear forces, or deflections.
Moments over the support are:
In the free cantilevers, M = M,
In the continuous structure, M = 3M,
MO-MI
Cd)
Ml
The final moment is therefore:
M, - M, = 0.40M, + 0.6O(fA4,)
c
= 0.80M,
an d
M, = 0.20 M,.
FIGURE 4.42.
crete
Redistribution of stresses through con-
creep.
At midspan,
moments are:
In the free cantilevers, M = 0
Similarly there is a relationship between the additional strain e2 and the corresponding stressfi produced at the same location by the same loads
applied in the continuous structure. One may
w rite:
E$
f2
=EC
where E,, the creep modulus, is given by:
1
-=
-1 - 1 E,
Ef
Ei
or
1
l 2=f2 --+
i Ef
1 1
Thus:
The corresponding total stress in the structure then
becomes:
In the continuous structure, M = MO/3
and the actual final moment:
M, = 0.60 +- = 0.20M,
The a b o ve derivation applies not only to external loads but also to the effect of prestressing.
Continuity prestress applied to a continuous
structure gives little internal redistribution of moments except in multispan structures, where the
spans react with one another according to the actual construction procedure. Cantilever prestress,
which acts to offset an appreciable part of the
dead-load moments, tends to reduce the distribution of moments due to external loads, Figure 4.43.
Up to now the concrete modulus has been assumed to ta ke only the two values Ei and E,
(short-term and long-term values). In fact, because
construction of a cantilever takes several weeks (or
even several months in the case of cast-in-place
structures), account must be taken of the concrete
strains versus the age and the duration of loading.
186
Design of Segmental Bridges
a = L/2
I
MGL
Pe
MO=
ia 1 dx
n
a dx
{
T
M
= Girder Load Cantilever Moment
=
=
Cantilever
MGL
-
Prestress
Moment
Pe Moment Inducing Redistribution
MO
= Moment at
6 under M in continuous beam
I
= Moment of Inertia (variable)
FIGURE 4.43. Computation of moment redistribution due to dead load and
cantilever prestress.
Such relationships are presented for normalweight prestressed concrete and average climate in
Figure 4.44.
Concrete strains are presented for convenience
as a dimensionless ratio between the actual strain
and the reference strain of a 28-day-old concrete
subjected to a short-term load.
We see that short-term strains vary little with the
age of the concrete at the time of loading except at
a very early age. However, long-term strains are
significantly affected by the age of the concrete.
For example, a three-day-old concrete will show a
final strain 2.5 times greater than a three-monthold concrete. This is particularly important for
cast-in-place structures with short cycles of construction (two pairs of segments cast and prestressed every week, which has now become common practice).
Two other factors play an important role in the
redistribution of stresses in continuous cantilever
bridges:
1. Relaxation of prestressing steel and prestress
losses. Because the stress in the prestressing
steel varies with time (a part of that variation
being due precisely to the concrete creep), the
internal moments that produce the deformation of the structure and therefore originate
the redistribution of stresses varv continuallv.
This factor is important because’the resultant
moments in the cantilever arms (dead load and
prestress) are given bv the difference of two
large numbers, and a variation on one usual+
has an important effect upon the result, Figure
4.43.
2. Change of the mechanical properties of
concrete section. For the sake of simplicity
gross concrete section is usually adopted
computation of bending stresses. In fact,
section to be used should be:
the
the
for
the
a. The net section (ducts for longitudinal
prestress deducted from the concrete sec-
Design of Longitudinal Members for Fkxure and Tendon Profile
zyxwv
187
Because it is difficult for some engineers to depend fully upon computer solutions in approaching a design problem, it is desirable to have orders
of magnitude of the moment redistribution for
preliminary proportioning and dimensioning of
the structure. The following guidelines are based
on experience and judgment.
L -mp”pDAYS
-v
MONT,,-3
YEARS
1. Consider the case of a symmetrical span
made up of two equal cantilevers fixed at the ends
and built symmetrically. Compute girder load
moments of the typical cantilever and prestress
moments using the final prestress forces and the
transformed concrete sections with n = 10 (average).
2. Compute the moment at midspan due to the
difference of the above two loading cases (Figure
4.43). More generally, compute in the final structure the moments in the various spans due to the
difference between cantilever girder load and
moments and final prestress moments, including
the restraint due to piers if applicable.
3. Reference is made now to the formula given
previously and repeated here for convenience:
FIGURE 4.44. Concrete strains versus age and dura-
tion ot loading. Note that strain is given as a dimensionless ratio beuqeen the actual strain and the reference
str-ain of a 2%dav-old concrete subjected to shot-t-term
hid.
b.
tion) for effect of girder load and prestress
up to the time of tendon grouting.
The transformed section (with incorporation of the prestress steel area with a suitable coefficient of transformation) after
grouting, where the coefficient of equivalence n = E,JE,, ratio of the modulus of
steel and concrete, should be taken as a
variable with time, from 5 to 12 or even 15.
The above discussion indicates the complexity of
the problem with respect to the material properties
and indicates the unreliable results of the early designs.
The only acceptable solution is the global approach, whereby a comprehensive electronic computer program analyzes step by step the state of
stresses in the structure at different time intervals
and whenever any significant change occurs, thus
following the complete history of construction.
Such programs are now available and have
proven invaluable in helping us understand
the behavior of segmental bridges. They provide
efficient tools for the final design of the structure.
w h e r e f = final stress (or moment or shear load
in the structure at any point),
,ft = stress at the same point obtained by
adding all partial stresses for each
construction step using the corresponding statical scheme of the
structure,
f2 = stress at the same point assuming all
loads and prestress forces to be
applied on the final structure with
the final statical scheme,
Ei = initial or intermediate modulus of
elasticity (short-term or for the duration of loading before continuity),
E, = final modulus (long-term).
Using different assumptions on the construction
sequence of bridge decks and the corresponding
strains as given by Figure 4.44, we find that the
average value of EfIEi would vary from 0.50 to
0.67. It is recommended that the conservative
value of 0.67 be used in this approximate method.
Thus the actual moment due to redistribution
should be 0.67, the value computed under paragraph 2. This moment must be added to the effect
of live load and thermal gradient at midspan.
Design of Segmental Bridges
188
tween cantilevers of different ages, and the redistribution of support moment may thus vary in wide
proportions, Figure 4.45. To keep on the safe side,
it is not recommended that the reduction in support moment be taken into account in designing
the prestress forces.
It is interesting at this stage to give some orders
of magnitude of moment redistribution by considering some fundamental formulas given as reference in Figure 4.46.
It has been assumed:
That the secondary moment due to the stressing of
continuity tendons is 6% of the total moment over
the support,
That the distance, n, between the center of gravity
of the cantilever tendons and the top slab is equal
to 0.05h.
MOMENTS
DUE 10
REDISTRIBUTION
That the center of gravity, depending upon the
section dimensions, may vary between (c,lh = 0.4
and c,lh = 0.6) and (c,lh = 0.6 and c,lh = 0.4).
ft-kips
+1260
a
BOTH CANTILEVER5 OF SAME AGE (BUILT IN
@
CANT(l)
CANT(2)
BUILT O-100 DAYS
BUILT 100-200 DAYS
@
CANT(2)
ONE
YEAR
OLDER
THAN
100
That the efficiency factor is p = 0.60.
DAYS)
From the data indicated above and in Figure
4.46, the percentage of prestressing steel,p, may be
determined as follows:
CANT(I)
FIGURE 4.45. Variation of redistribution moment in
cantilever construction with the construction procedure.
4. Correspondingly, the support moment (over
the piers) is decreased by the same amount. In fact,
the construction of cantilevers in successive stages
is such that continuity is achieved in each span be-
r2/c2
-I
1
I
i
r2/c1
assuming a final stress in the tendons of 160 ksi
assuming a maximum compressive stress in the
bottom flange of 2000 psi:
P =
T
Cl
r\”
A J&, = 2 0 0 0 +
limit of the central core
r2
r2
-= PC1
CZ
= PC2
s;
p = efficiency factor
c2
average
stress = 2000 g
.fi
\
~200~ psi
FIGURE 4.46. Approximate moment redistribution (moments over support). Total
moment: MT = MGL + MSL + M L,,where MGL = girder load moment, MSL = superimposed load moment, ML, = live-load moment (including impact). Assumed secondary
moment due to continuity prestress: 0.06 M,. Final prestress force: P = 0.94M,l[r +
(r*/c,)] = 0.94MJ(e + pc2). Prestress moment (1): Pe = 0.94M,l[l + (pcJ~)]. Momentinducing redistribution: MGL - Pe, given by (2): (MGL - Pe)IM, = M,,lM, - 0.94/[1 +
hJe)l.
zyxw
zy
zyxwvutsrqponm
Design of Longitudinal Members for Flexure and Tendon Pro@
AS
P = A,--8;))
For a symmetrical section, cr = 0.5h, andp would,
thus, be equal to 0.63%, a reasonable and common
value. The transformed percentage area of the
steel with n = 10 is equal to:
np =
sections is plotted versus the position of the centroid with or without transformed area.
It is interesting to study the effect of an accidental variation in the prestress load due to excessive friction in the ducts. Assume, for example, a
reduction of 5% in the prestress load for the case
c,lh = 0.5 (symmetrical section over the support)
and M,,lM, = 0.80.
The intial values of (M,, - Pe)IM, are changed
as follow s:
0.125 +
All mechanical properties of the section change
to make the denominator of equation (2) in Figure
4.46 increase and, consequently, the momentinducing redistribution increase also. This fact,
which was completely overlooked for many years,
is clearly seen in Figure 4.47, where the percentage
of moment-inducing redistribution in the various
0.600 -
0.500 -
I
I
Gross area
Transformed
area
100%
95%
Prestress
Prestress
Percent
Variation
0.264
0.292
1.12
1.10
0.236
0.265
The combined effect of tendon grouting and of
added friction losses increases the redistribution of
moments by 25%.
I
-.--GrossArea
Transformed
I
I
Area
Cl lh
0.35
I
0.40
I
0.45
I
0.50
I
0.55
I
0.60
0.65
czlh
0.65
0.60
0.55
0.50
0.45
0.40
0.35
Figure 4.47.
189
M oment redistribution, numerical values over support.
D esign of Segmental Bridges
190
4.8.8
PREDICTION OF PRESTRESS LOSSES
The prediction of losses in prestressed concrete
has always been subject to uncertainty. This is due
to the high stress levels used for the prestressing
steel, the variable nature of concrete, and its propensity to creep and shrink. As recently as 1975,
AASHTO made a major revision to its code to provide improved methods for predicting prestress
losses. The Structural Engineers Association of
California has an excellent report on creep and
shrinkage control for concrete in general. The report concludes that special attention should be
given to material selection and proportioning. For
creep and shrinkage calculations many European
engineers recommend the guidelines of the Federation Internationale de la Precontrainte, Comiti
Europeen du B&on (FIP-CEB).
The design computations for segmental prestressed concrete bridges are very involved for the
construction phase. Every time a segment is added
or a tendon is tensioned, the structure changes,
and it must be reanalyzed. As the segment ages, the
concrete and prestressing steel creep, shrink, and
relax. Thus, each segment has its own life history
and an elastic modulus that depends upon the age
and composition. To accurately compute all of
these effects by hand, throughout the life of the
structure, would be very difficult, particularly
during the construction phase. Comprehensive
computer programs such as “ BC” (Bridge Construction) and others have been recently developed
and are now available to aid the design engineer.
In addition to construction analysis, these programs will check the completed bridge in accordance with AASHTO specifications. It is possible
to revise them to satisfy other codes or loadings,
such as A REA .
Not only are all prestress losses properly evaluated and taken into account, but redistributions of
moments due to concrete creep and steel relaxation are automatically incorporated in the design
analysis.
4.9
Ultimate Bending Capacity
Longitudinal M embers
of
Basically, the d esig n ap p ro ac h o f seg m ental
bridges is one of service load. It is important, however, not to lose sight of the ultimate behavior of
the structure to ensure that safety is obtained
throughout.
In simply supported structures, the ultimate
capacity is very simply analyzed by comparing in
the section of maximum moment:
The total design load moment including girder
load and superimposed load (DL) and live load
(LL )
The ultimate bending moment of the prestressed
section M,
Depending on the governing codes and the usual
practice in various countries, this comparison may
be done in various ways:
Apply a load factor on DL and LL and a reduction
factor for materials on M,
Apply a single factor K on (DL + LL) and compare
w ith M,
Apply a single factor K on LL only and compare DL
+ KLL with M,
In all cases, the designer must first compute the
ultimate capacity of the section considering the
concrete dimensions and characteristics of prestressing tendons (and possible conventional reinforcement). From previous studies it may be shown
that the ultimate moment of a prestressed section is
computed very simply by considering a dimensionless factor called the weight percentage of prestressing steel, q (see Figure 4.48).
To account for the fact that the concrete characteristics are less reliable than those of the prestressing steel, which are well known and very constant, fs is usually taken equal to the guaranteed
minimum tensile strength, whereas,fi is assumed to
be only 80% of the 28-day cylinder strength.
Considering now the case of- segmental superstructures, which are most generally continuous
structures, one may take the conventional approach of considering the various sections of the
m em ber (fo r exam p le, sup p o rt sectio n and
midspan sections in the various spans) as independent from one another in much the same way
as for simple members. Such simplification overlooks the capacity of the redundant structure to
redistribute, internally, the applied loads, which
seems to be a conservative assumption.
In fact, it is not always as conservative and safe as
it looks, as will be shown by an example computed
numerically for a typical span of the Rio Niteroi
Bridge. For such a span the design moments are as
follows (in foot-kips x 1000):
191
Ultimate Bending Capacity of Longitudinal Members
As prestressing
steel
FIGURE 4.48. Ultimate moment of a prestressed section. (1) Dimensionless coefficient,
q’ = (A,lbd) f&/ f:), whereA, = area of prestressing steel, 6 = width of section,d = effective
depth of section (distance between centroid of prestress and extreme compression fiber),
f,i = ultimate tensile strength of prestressing steel,fi = ultimate compressive strength of
concrete. (2) Value of ultimate moment: for q’ < 0.07, M, = 0.96A&d; for 0.07 < q’ <
0.50, M, = (1 - O.Gq’)A,J-?‘d.
Support
Girder load
Superimposed load
Midspan
116
10
Total dead load (DL)
Total live load (LL)
0
5
5
22
-
126
29
Total (DL + l.L)
155
Live-load moment in simple span: 37
27
The ultimate moments have been computed for all
sections for both positive and negative bending.
The envelopes of ultimate moments are shown in
Figure 4.49.
Neglecting any moment redistribution, the situation would be the following over the support and
at midspan:
Section
Mo ment
Support
Midspan
IV”
256
79
DL
LL
M, =
or .M, =
126
29
1.65(DL + LL)
DL + 4.5 LL
5
22
2.93(DL + LL)
DL + 3.4LL
The picture is substantially different when looking
at redistribution due to plastic hinges. Assuming
an overall increase of both dead and live load
simultaneously (loading arrangement A), we ob-
tain the overall safety factor by comparing the sum
of ultimate moments over the support and at
midspan:
256 + 79 = 335
and the sum of simple span moment due to DL and
LL:
DL: 126 + 5 = 131
LL
37
Total
iii
The overall safety factor is thus:
K = $ =2.0
approximately 20% higher than for the support
section considered alone. In fact, it is more important and more realistic to consider only an increase
of the live load, which is the only variable factor in
the structure. Proceeding as before, the safety
factor on LL only would be:
K
=
335
-
37
13’
=
j
j
.
However, this is not the actual safety factor of the
structure, because there exists a more aggressive
loading arrangement than that where all spans are
live loaded. In the case where the live load is
applied to only every second span [arrangement
r
Design of Segmental Bridges
192
260
!-
Elevation
I
I
I
I Live-load arrangement (A)!
I
I
zyxwv
260
260
Support
I
FIGURE
4.49.
I
I
-
*
I
I
!
, support
I
1
I
i-----f
izyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPO
I
I
Ultimate bending capacity of’ a continuous deck.
(b) in Figure 4.491, the first plastic hinge will appear at the center of the unloaded spans w ith a
negative moment (tension at the top fiber) and the
support moment reaches the following limiting
v alue:
Ultimate negative moment at midspan:
38
Actual dead-load moment in simple span:
126 + 5 = 131
169
This value of 169 is substantially low er than the ultimate moment at that support section considered
by itself (M, = 256).
The failure appears when the second plastic
hinge appears at the center of the loaded span
under positive moment (tension at the bottom
fiber). The limiting value of the safety factor K is
such that:
169 + 79 = 131 + K . 37 and K = 3.2
In such structures a very important characteristic
must be emphasized. At the time of ultimate load
failure, due either to negative moment in the unloaded spans or positive moments in the loaded
spans, the maximum moment over the support has
only slightly increased above the value at design
load ( 169 against 155) and is far below the ultimate
moment of the section (256). Three interesting
consequences may be derived from this fact:
1. Because the overall safety of the structure is
not dependent upon the ultimate moment
near the supports, it is not necessary to dimension the bottom flange of the concrete section
in this area to balance the ultimate capacity of
the prestressing tendons.
2. The global safety factor of the structure depends directly on the capacity of the sections
near midspan for both positive and negative
moments. The capacity for positive moments is
given by the continuity tendons placed in the
Shear and Design of Cross Section
3.
193
bottom flange for service-load conditions. The
capacity for negative moments depends upon
the tendons placed at the top flange level to
overlap the cantilever tendons of the two indiiidual cantilever arms. The magnitude of this
overlap prestress does not appear as a critical
factor when designing the structure for service
loads, yet it plays an important role in the ultimate behavior of the structure.
the beneficial effect of longitudinal compression
(either in columns subject to axial load or in prestressed members) is taken into account.
At ultimate load, it was shown that the areas of
the members close to the supports are subjected to moments only slightly in excess of design load moments and in most cases below
cracking moments. No early failure due to
combined shear and bending is anticipated.
4.102 SHEAR TESTS OF REINFORCED
CONCRETE BEAMS
In long structures where hinges and expansion
joints are provided in certain spans, the same design principles may be applied to analyze the ultimate capacity. Hinges represent singular points
through which the moment diagrams must go regardless of the loading arrangement under consideration. It was found that the optimum location of
the hinge with regard to ultimate safety is somewhat different from the location allowing the best
control of long-term deflections. It may be of
interest therefore to move the hinge slightly toward the center of the span, which has a further
advantage of simplifying construction.
4.10
Shear and Design of Cross Section
4.10.1
I,\‘TRODUCTIO,V
Designing prestressed concrete members for shear
represents a challenging ta sk for the engineer, because there are many differences of opinion and
large variations in the requirements of the various
codes. In particular t h e AC1 c o d e a n d t h e
AASHTO specifications differ in several ways
from the FIP-CEB and other European codes.
It is common practice in many countries to design reinforced concrete and prestressed concrete
members for shear by allowing the concrete to
carry a proportion of the shear loads while stirrups
(formerly in conjunction with inclined bars) carry
the rest. A complete agreement has not yet been
reached on this aspect of design for shear:
The French codes (CCBA, for example) allow
nothing to be taken by the concrete and the total
shear to be carried by the transverse steel, which is
certainly an overconservative approach. Obviously,
The recent FIP-CEB code allows some proportion
of the shear to be carried by the concrete.
AC1 code allows a larger proportion of shear to
be carried by the concrete with a consequent savings in stirrup requirements.
Tests were recently carried out in France in order
to increase the knowledge of this phenomenon,
both on simply reinforced concrete and on prestressed members.4 Static tests on reinforced concrete I beams showed that the steel stress in stirrups increases linearly with the load and is three
times smaller than it would be if the concrete carried no shear, Figure 4.50. In this respect, all codes
are fully justified in taking the concrete into account as a shear-carrying component.
However, dynamic testing on the same beams
showed a very different behavior. A cyclic load was
applied between one-third and two-thirds of the
ultimate static load for one million cycles, whereupon the beam was statically tested to failure, Figure 4.51. Before cracking, the elastic behavior of
the homogeneous member kept the steel stress in
the stirrups very low. However, before 10,000 cycles, a crack pattern had appeared that remained
to the end of the test and became more and more
pronounced with a continuous increase of the inclined crack width. Crack opening reached &r in.
(1.5 mm) at the end of the dynamic test. Most
probably stirrup rupture took place about 600,000
cycles, although the ultimate static capacity of the
FIGURE 4.50. Static test of reinforced concrete Ibeam steel stress in stirrups.
Design of Segmental Bridges
lbre de cycles
~(log N) ,
1,
+
aa 106
FIGURE 4.51. Dynamic test of reinforced concrete
I-beam web cracking and variation of steel stress in stirrups.
beam after dynamic testing was substantially the
same as for the other beams, which were tested
only under static loads. Such tests show that the
conventional approach of designing web reinforcement for static loading with a large part of the
shear carried by the concrete may not provide adequate safety in the actual structures as soon as web
cracking is allowed to develop.
4.103
DIFFICULTIES IN ACTUAL STRUCTURES
Another source of information is afforded by the
behavior of existing structures. Fortunately,
examples of difficulties imputable to shear in cantilever box girder bridges are scarce. The authors
are aware of only two such contemporary examples, which are summarized here for the benefit of
the design engineer.
The first example relates to a box girder bridge
deck constructed by incremental launching and
shown in Figure 4.52. Permanent prestress was
achieved by straight tendons placed in the top and
bottom flanges, as required by the distribution of
moments. During launching an additional uniform
prestress was applied to the constant-depth single
box section, which produced an average compressive stress of 520 psi (3.60 MPa). Near each pier
there was a vertical prestress designed to reduce
web diagonal stresses to allowable values.
During launching a diagonal crack appeared
through both webs between the blisters provided in
the box for anchorage of top and bottom prestress.
The corresponding shear stress w as 380 psi (2.67
MPa), and there was no vertical prestress in that
zone. The principal tensile stress at the centroid of
the section w as 200 psi (1.40 MPa), w hich is far
below the cracking strength of plain concrete. In
fact, the webs of the box section were subjected to
additional tensile stresses due to the distribution of
the large concentrated forces of the top and bottom prestress. The truss analogy shown in Figure
4.52 indicates clearly that such tensile stresses are
superimposed on the normal shear and diagonal
stresses due to the applied dead load and may
therefore produce cracking. This could have been
prevented by extending the vertical prestress in the
webs further out toward midspan.
The second example concerns a cast-in-place
variable-depth double box girder bridge with
maximum span lengths of 400 ft. Because the
bridge was subsequently intend ed to carry
monorail pylons, two intermediate diaphragms
were provided at the one-third and two-thirds
points of each span, as show n in Figure 4.53. Prestress was applied by straight tendons in the top
and bottom flanges and vertical prestress in the
webs to control shear stresses. Diagonal cracking
was observed in the center web only near the intermediate diaphragms with a maximum crack
opening of 0.02 in. (0.6 mm). Repair was easily accomplished by adding vertical prestress after
grouting the cracks.
A complete investigation of the problems encountered revealed that cracking was the result of
the superposition of several adverse effects, any
one of which was almost harmless if considered
separately: (1) The computation of shear stresses
failed to take into account the adverse effect (usually neglected) of the vertical component of continuity prestress in the bottom flange of a girder
with variable height. (2) The distribution of shear
stresses between the center and side webs was
made under the assumption that shear stresses
were equal in all three webs. In fact the center web
zyxwv
zyxw
Shear and Design of Cross Section
,,i:‘,: t:“:“, \ I’
,A
NO
VERTICAL
PRESTRESS
IN
THAT
ZONE
195
TOP PRESTRESS
/
TYPICAL tIAlF (ROSS SKTlOIi A- A
FIGURE
4.52.
Example of web cracking under application of’ high prestress forces.
f
FIGURE 4.53. Example of web cracking in a 400 ft span. (a) Typical cross
section. (6) Partial longitudinal section.
carries a larger proportion of’ the load, and shear
stresses were underestimated for this web. (3) The
vertical web prestress was partially lost into the intermediate diaphragms, and the actual vertical
compressive stress was lower than assumed. (4)
Present design codes do not provide a consistent
margin of safety against web cracking when vertical prestress is used. This margin decreases
significantly when the amount of vertical prestress
increases. In the present French code, the safety
Design of Segmental Bridges
196
factor against web cracking is 2 when no vertical
prestress is used and only 1.3 for a vertical prestress of400 psi. (5) At present, vertical prestress is
usually applied with short threaded bars, and even
when equipped with a fine thread they are not
completely reliable unless special precautions are
taken under close supervision. Even a small anchor
set significantly reduces the prestress load, and it is
not unlikely that the actual prestress load is onlk
three-fourths or even two-thirds of the theoretical
prestress.
It sho u ld , ho w ev er, b e em p hasiz ed that the
difficulties mentioned above have led to progress
in this field, and the increase in knowledge has ensured that these examples remain rare exceptions.
Practically all existing box girder bridges have perf o rm ed exc ep tio nally w ell u nd er the ef f ec ts o f
shear loads and torsional moments.
The essential aspects of
are:
this important problem
Dimensioning of the concrete section particularly
in terms of web thickness
Design of transverse and/ or vertical prestress and
of conventional reinforcement
The twvo
FIGURE 4.54. C o m p u tatio n o f net ap p lied ~IGIIload. (0) \Terrical comporicrit of’ pr-esrr.css. (h) k:f fvc t 01
inclined bottom flange (Resal effec-t). (0 Net \hr;rt- ~OI-cc.
Shear f’orce due to applied loada = I.: dcdrrct \cr-tic al
component of‘dr-aped tendons = - 1 P 5irr a,; aclcl \-erticdl component of continuitv te ndo ns = * 2 t’ 4irr a?: ticduct Red effect = - f,v.tlB ta n p: total ih ner applied
shear t’k~ = I’,,.
are o b tained bv c o nsid ering stresses on sections
perpendicular to the top flange (\vhich
is usually
the orientation of joints betlveen segmenta) and
projecting the loads on the section for determining
shear stresses. The total
. net shear force is the sum
of the following terms:
major considerations are:
Shear force due to applied loads.
At the design stage (or, in modern code language,
serviceability limit state) prevent or control cracking so as to avoid corrosion and fatigue of reinforcement.
Reduction due to vertical component of
tendons where used.
c\t the ultimate stage (or load factor design concept
state o r u ltim ate lim it state) p ro v id e ad eq u ate
safetv.
Reduction due to the inclined principal compressive stresses in the bottom flange (usuallv called the
Resal effect after the engineer who fi;st studied
members of variable depth). Because the direction
of the principal stresses in the web is not fullv determined, it is usual to neglect the added reduction
of shear force derived from web stresses.
For the box sections used in cantilever bridges the
behavior under shear must be investigated:
draped
Increase due to inclination ofcontinuitv tendons in
the bottom flange for variable-depth girders.
In the webs.
At the connections between web and top flange (including the outside cantilevers) and web and bottom flange. Figures 4.54 and 4.55 show a suggested
method to compute shear loads and shear stresses.
Modern computer programs analyze the box
girder cross sections perpendicular to the neutral
axis and ta ke into account all loads projected on
the neutral axis and the section. Equivalent results
Shear stresses m ay f u rther b e c o m p u ted f ro m
shear force and torsion r;loment using the conventional elastic methoils.
Tests have shown that the presence of draped
tendon ducts in the webs, even if grouted after tensioning, changes the distribution of’ shear stresses.
To take this effect into account, it is suggested to
compute all shear stresses using a net web thickness
that is the actual thickness minus one-half the duct
197
Shear and Design of Cross Section
t
b
b’ Gross web thickness
d diameter of duct
(a)
FIGURE 4.55. Computation of shear stress. Typical box section:
net web thickness = h = b’ - Id; shear stress due to shear force V, net
applied shear load = v = V,;Q/[(Xb).I], w here Q = statical moment at
centroid, b = net web thickness, I = gross moment of inertia, V,, = net
applied shear load; shear stress due to the torsion moment = v =
CI(P.b.3, w here C = torsion moment, b = net web thickness, S = area
of the middle closed box. Note: check the shear stress at centroid level.
diameter. Ducts for vertical prestress need not be
taken into account because they are smaller and
parallel to the vertical stirrups, which compensates
for the possible small effect of the prestress ducts.
Web-thickness dimensioning depends upon the
magnitude of shear stress in relation to the state of
compressive stress. In the case of monoaxial compression (only longitudinal prestress and no vertical prestress) the diagonal principal tensile stress
must be below a certain limit to insure a proper
and homogeneous margin of safety against web
cracking with its resulting long-term damaging effects. Figure 4.56 suggests numerical values based
on the latest state of the art that are believed to be
realistic and safe. Numerical values for.allowable
shear stresses under design loads are given in Figures 4.5’7 and 4.58 for 5000 and 6000 psi concrete.
Web thickness must therefore be selected in the
v ario u s sec tio ns alo ng the sp an to keep shear
stresses within such allowable values. It may be
that construction requirements or other factors
make it desirable to accept higher shear stresses.
It is necessary in this case to use vertical prestress
to create a state of biaxial compression. Figure
4.566 ind ic ates the c o rresp o nd ing p ro c ed u re.
The vertical compressive stress must be at least 2.5
times the excess of shear stress above the value
for monoaxial compression.
When vertical prestress is used, the beneficial effect of increasing the length of the horizontal component of the potential crack in the web created by
the horizontal compression due to prestress is partially lost. In fact, if both horizontal and vertical
compressive stresses are equal,f, = fU, the direction of the principal stress is given by /3 = 45” as in
Y
fx
v
,‘p
fx
-'\ I tu7
(a)
q1 l-&’u 1;”
(b)
FIGURE 4.56. Allowable shear stress for mono- and
biaxial compression in box girders. (a) Monoaxial compression: allowable shear stress = v = 0.05f:. + 0.2Of,;
co rresp o nd ing d iago nal tensio n = fP given by v2 = fPcfs +
f,). (6) Biaxial compression: allowable shear stress = zl =
0.05f:. + 0.2Of, + 0.4Of,; corresponding diagonal tension = fp given by v’ = cfs +f,) (fU +fP).
198
Design of Segmental Bridges
stresses higher than a limiting value of lo* be
accepted prior to careful investigation based on
specific experimental research.
In this respect, a very interesting case arose for
the construction of the Brotonne Viaduct in
France (described in Chapter 9), where an exceptionally long span called for minimum weight and
consequently high concrete stresses. The most
critical condition for shear stresses developed in
the 8 in. (0.20 m) webs near the piers of the approach spans, where a maximum shear stress of
640 psi (4.5 MPa) was accepted together with an
unusually low longitudinal compression stress of
500 psi (3.45 MPa). Vertical prestress was used in
this case. The chart for a 6000 psi concrete, Figure
4.58, would give:
In monoaxial state withf, = 500 psi, V = 400 psi.
FIGURE 4.57. r\llo\<able
shear stresses forf:. = 5000
psi.
In biaxial state withf, = 550 psi, V = 620 psi, which
is substantially equal to the actual shear stress of
6 4 0 psi.
A test was conducted to study the behavior of the
precast prestressed web panels in the normal design load stage and up to failure, Figure 4.59. Results are shown in Figure 4.60. The ultimate
capacity of the web was very large and probably far
in excess of the needs. It is believed that web
FIGURE 4.58. Allowable shear stresses forf:.
= 6000
psi.
ordinary reinforced concrete. If a higher vertical
stress is use d, a crack with p > 45” could develop,
with a *consequent reduction of the horizontal
length over which concrete and reinforcement
must carry the total shear. To prevent such a situation, it is deemed preferable to use a vertical compressive stress not greater than the longitudinal
compressive stress, fu < fz.
Finally, considering present knowledge on the
behavior of prestressed concrete beams under high
shear stresses, it is not recommended that shear
PLAN
TRANSVERSE
FIGURE
4.59.
cast web panels.
VIEW
SECTION
Brotonne Viaduct, test set-up for pre-
Joints Between Match-Cast Segments
Strrssrs at &sign stug~ (approach viaduct):
500 psi
550 psi
6 4 0 psi
Horizontal compressive stress
Vertical compressive stress
Shear stress
Rrsulb of test at rufduw:
Normal Lo ad
630 t
84o’t
Ultimate shear
Horizontal compressive stress
Vertical compressive stress
Shear stress (elastic theory)
uniform
840x 1.3
630
1650 psi
5 8 0 psi
3 3 0 0 psi
2 2 0 0 psi
.JoinI destroyed and multiple keys sheared
off. Panels intact.
FIGURE 4.60. Brotonne Viaduct, results of precast
\\,eb panel tests.
cracking control can be obtained only by proper
stress limits at the design load level.
When designing longitudinal bridge members
for shear, another important factor remains to be
considered, which has sometimes been overlooked
t& inexperienced designers. It concerns longitudinal shear stresses developing between the webs and
the top and bottom flanges as shown in Figure
4.6 1. When web stresses have been verified at the
level of the centroid, it is not necessary to make a
detailed study at other points of the web [such as
levels (d) and (e)], although the principal tensile
stress near the pier may be slightly higher at point
(d) than at the center of gravity. On the other
hand, to keep the integrity of the box girder, it is
verv important to verify that shear and diagonal
stresses in sections (a), (b), and (c) are within the
//
/
/
FIGURE 4.61. Longitudinal shear between web and
flanges.
199
same allowable values as set forth previously for
the webs and that a proper amount of reinforcing
steel crosses each section.
This leads to the design of transverse reinforcement in the cross section to resist shear stresses.
According to the provisions of the AC1 Code and
the AASHTO specifications, the web shear steel
requirements are controlled by the ultimate stage.
The net ultimate shear force is given by the following formula, based on the current partial load
factors:
V, = 1.3OV,,, + 2.17V,, + V,
w h e r e V, = net shear force at ultimate stage,
VDL = actual shear force due to the effect
of all dead loads, including the reduction due to variable depth where
applicable,
VLL = shear force due to live loads including impact,
VP = unfactored vertical component of
prestress where applicable.
Effects of temperature gradients and volume
changes are usually small in terms of shear load
and may be neglected except in rigid frames. On
the contrary, shear due to moment redistribution
and secondary effects of continuity prestress must
be included. A partial safety factor on material
properties is applied to the ultimate load state.
4.11 Joints Between Match-Cast Segments
Joints between match-cast segments are usually
filled with a thin layer of epoxy to carry normal
and shear stresses across the joint. In the early
structures, a single key was provided in each web of
the box girder to obtain the same relative position
between segments in the casting yard and in the
structure after transportation and placing. This
key was also used to transfer the shear stresses
across the joint before polymerization of the
epoxy, which has substantially no shear strength
before hardening. Figure 4.62 summarizes the
force system in relation to a typical segment both
during erection and in the completed structure.
Provisional assembly of a new segment to the
previously completed part of the structure is usually achieved by stressing top (and sometimes bottom) longitudinal tendons, which induce forces F,
(and F2). The resultant F of F, + F2 resolves with
the segment weight W into a resultant R. The vertical component of R can be balanced only by a reac-
200
Design of Segmental Bridges
FIGURE 4.62. Typical segment in relation to the force system. (0) Provisional assembly
of segment(s). (b) Segment(s) in the finished structure.
tion such as R, given by the inclined face of the key,
while the balance of the normal force is R, which
produces a distribution of longitudinal compressive stresses. In the finished structure, all normal
and shear stresses are naturally carried through
the joints by the epoxy material, which has compressive and shear strengths in excess of the segment concrete.
A series of interesting tests were performed for
the construction of the Rio-Niteroi Bridge in Brazil
to verify the structural behavior of epoxy joints
between match-cast segments. A l-to-6 scale model
was built and tested to represent a typical deck
sp an near the su p p o rt and the c o rresp o nd ing
seven segments as shown in Figure 4.63.
ELEVATION
FIGURE
4.63.
A crack pattern developed in the web when the
test load was increased above design load, as shown
in Figure 4.64. The epoxy joints had no influence
on the continuity of the web cracks, and the behavior of the segmental structure up to ultimate
was exactly the same as that of a monolithic structure. Failure occurred for concrete web crushing
when the steel stress in the stirrups reached the
yield point. The corresponding shear stress was
970 psi (6.8 MPa) for a mean concrete cylinder
strength of 4200 psi (29.5 MPa).
The first bending crack had previously occurred
for a load equal to 93 percent of the computed
cracking load, assuming a tensile bending strength
of 550 psi (3.9 MPa). Other tests w ere performed
DETAIL OF JOINT
Rio-Niteroi Bridge, partial elevation and joint detail.
201
Joints Between Match-Cast Segments
4.64.
FIGURE
Rio-h’itcroi Bridge. ~vel) crack pattern at ultimate in model test.
in order to study the transfer of diagonal principal
compressive stresses across the segment joints as
shown in Figure 4.65. Prismatic test specimens
I
were prepared, some with and some without shear
keys across the joint, and tested for various values
of p, the angle between the principal stress and the
neutral axis of the girder. In the case of the Rio
Niteroi Bridge the value of /3 is between 30 and
35”. For a reinforced concrete structure p = 45”.
A preliminary test showed that the epoxy joint
had an efficiency of 0.92 as compared to a
monolithic specimen with no joint (ratio between
the ultimate load P on the prismatic specimen with
an epoxy joint and with a monolith specimen). For
various directions of the joint the results are as
follows:
P
0”
15”
30”
fd zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
Efficiency
0.94
0.92
0.98
PRlSMATlC
PRISMATIC
WITH KEYS
P
P
I
I
t
1
P
P
(b)
FIGURE 4.65. Rio-Niteroi Bridge, test specimens for
web. (a) Crack pattern in web and related test specimen.
(b) Actual test specimens.
45”
0.95
60”
0.70
It can be seen that for values of p smaller than 45”
(which covers the entire field of prestressed concrete members) the compressive strength is hardly
affected by the presence of the inclined joint. All
these tests confirmed earlier experimental studies
to show that epoxy joints are safe provided that
proper material quality together with proper mixing and application procedures are constantly obtained.
Several early incidents in France, and some more
recently in the United States, have shown that
these conditions are not always achieired. The logical step in the development and improvement of
epoxy joints was therefore to relieve the epoxy of
Design of Segmental Bri&es
202
any structural function. The multiple-key (or
castellated-joint) design embodies this concept and
provides for simplicity, safety, and cost savings.
Webs and flanges of the box section are provided
with a large number of small interlocking keys designed to carry all stresses across the joint with no
structural assistance from the resin. Figure 4.66
shows the comparison between the structural behavior of an early joint with a single web key and a
joint with multiple keys, assuming that the epoxy
resin has improperly set and hardened. It is now
recommended that multiple keys be used in all precast segmental projects, as shown in Figure 4.67.
With the current dimensions used for depth and
height of multiple keys, the overall capacity of the
joint is far in excess of the required minimum to
transfer diagonal stresses safely up to the ultimate
load state.
C o ntinuo us tra nsfe r
o f she a r stre sse s
(It)
Design of Superstructure Cross Section
4.12
The typical cross section of a box girder deck is a
closed frame subjected to the following loads, Figure 4.68:
Girder weight of the various components (top and
bottom flanges, webs)
Superimposed loads essentially applied to the top
flange (barrier, curbs and pavement) and sometimes to the bottom flange, as when utilities are installed in the box girder
Live loads applied on the deck slab
A typical box girder element limited by two parallel
cross sections, Figure 4.686, is in equilibrium because the applied loads are balanced by the difference between shear stresses at the two limiting
sections. To design the typical cross section the assumption is usually made that the shape of the section remains unchanged and that the closed frame
may be designed as resting on immovable supports
such as A and B. Bending moments are created in
the various sections of the frame due to the applied
loads. Maximum moments occur in the deck slab
due to live loads in sections such as (a), (b), and (f).
Cb)
FIGURE 4.66. Joint between match-cast segments,
comparison between single- and multiple-key concepts.
Cd)
FIGURE 4.67.
Precast segment with multiple keys.
FIGURE 4.68. Design of deck cross section. (a) Typical loading on cross section. (b) Free-body diagram.
Special Problems in Superstructure Design
Because the webs are usually much stiffer than the
flanges and the side-deck slab cantilevers and the
center-deck slab between webs are built into the
webs, most of the deck-slab moments are transferred to the web, with a maximum value in section
(d) at the connection between web and top flange.
In bridges where transverse or vertical prestress or
both are used, the design of the deck cross section
is not greatlv affected by the fact that moments and
normal forces computed in the frame superimpose
their effects on the shear stresses due to longitudinal bending mentioned in Section 4.10.
The case is more critical when only conventional
transverse reinforcing steel is used in both flanges
and webs. A common method, based on experience, is to compute the steel area required on
either face at critical sections such as (a) through
(e), shown in Figure 4.68, for the following:
1.
Shear stresses in the longitudinal members.
2.
Transverse bending of the frame.
The minimum amount of steel should not be less
than the larger of the following:
item 1 plus one-half of item 2,
item 2 plus one-half- of item 1, or
0.7 times the
4.13
,
Special
sum
of item 1 and item 2.
Problems
in
Superstructure
Design
All design aspects covered in the preceding sections pertain to the design of deck members for
bending and shear regardless of the local problems
encountered over the piers or abutments and at
intermediate expansion joints when required. This
section w ill now deal w ith such local problems,
which are of great practical importance.
4.13.1
DIAPHRAGM S
It was mentioned in Section 4.6 that the combined
capacities of the deck slab in bending and the box
girder in torsion allow a very satisfactory transverse distribution of live loads between girders in
the case of multiple box girder decks. It has therefore been common practice to eliminate all transverse diaphragms between box girders except over
the abutments. Diaphragms inside the box section
are still required over the intermediate piers in
most projects.
203
4.13.2 SUPERSTRUCTURE OVER PIERS
The simplest case is exemplified in Figure 4.69,
where a deck of constant depth rests upon the pier
cap with bearings located under the web of the box
girder. The reaction is transferred directly from
the web to the bearings, and there is need only for
a simple inside diaphragm designed to transfer the
shear stresses, due to possible torsion moments, to
the substructure. A more complicated situation
arises when the bearings are offset with regard to
the webs, Figure 4.70. Reinforcing and possibly
prestressing must be provided in the cross section
immediately above the pier to fullfill the following
functions:
Suspend all shear stresses carried by the web under
point A, w here a 45” line starting at the bearing
edge intersects the web centerline (hatched area in
the shear diagram).
Balance the moment (R * d) induced by the bearing offset.
Looking at other schemes, we find that decks of
variable depth pose several challenging problems.
Figure 4.71 shows an elevation of a box girder
resting on twin bearings designed to improve the
rigidity of the pier-to-deck connection and consequently reduce the bending moments in the
deck, which will be described in greater detail in
Chapter 5.
When the loading arrangement is symmetrical in
the tw o adjacent spans, the transfer of the deck
reaction into the piers through the four bearings is
just as simple as for the case show n in Figure 4.69.
Matters look very difficult for an unsymmetrical
loading condition either in the completed structure, Figure 4.71, or during construction, Figure
4.72. Let us assume that the total deck reaction is
transferred to the pier through one line of bearings
only (for example, R, in Figure 4.71, for an excess
of load in the left span). The compression C, carried by the bottom flange at the right is no longer
balanced by the corresponding reaction R,, and an
abrupt change in the system of internal forces results in a large vertical tensile force T,, w hich has to
be suspended on the total width of the box section
by special reinforcement or prestress. In long-span
structures, these local effects are of no small magnitude. Taking the example of a 40 ft (12 m) wide
box with a 20 ft (6 m) wide bottom flange and a
span of 300 ft (90 m), the load carried by the bottom flange will probably be around 3000 t (2720
mt) and the angle change above the right bearing
zyxwvuts
Design of Segmental Bridges
t
~~~~ t-
b e a r in g s
S E C TIO N
FIGURE
FIGURE
4.70.
4.69.
A-A
S E C TIO N
zy
c-c
Pier segment for deck of constant depth and simple support.
Deck over piers with offset bearings.
about 10 percent. The corresponding unbalanced
load is therefore 300 t (272 mt), and this is more
than enough to split the pier segment along the
section between the web and the bottom flange if
proper consideration has not been given to the
problem with respect to design and detailing.
The situation may be even more critical during
construction, Figure 4.72, if the unbalanced mo-
ment induces uplift in one of the two bearings. The
load of the anchor rods (2) has to be added to the
unbalanced load resulting from the angle change
of the bottom flange.
The diaphragm systems shown in Figures 4.71
and 4.72 are of the A type where both inclined
diaphragm walls intersect at the top flange level.
Any unsymmetrical moment that produces a tension force in the top flange T and a compression
force in the bottom flange may thus be balanced by
normal loads such as F, and C,, Figure 4.7 1, with
no secondary bending. In this respect, then, it is a
satisfactory scheme. Detailing may, however, be
difficult because of the concentration of reinforcement or prestress tendon anchors in the top
flange area, which usually is already overcrowded
with longitudinal tendon ducts. A simple and more
practical design, although less satisfactory from a
theoretical point of view, is to provide vertical
diaphragms above the bearings. This is the logical
choice when the deck is rigidly connected with a
205
Dejections of Cantilever Bridges and Camber Design
4.13.4
EXPANSION JOINT AND HINGE SEGMENT
The expansion joints required at intermediate
points in very long structures need a special segment to transfer the reaction between the two sides
of the deck. When the expansion joint is located
close to the point of contraflexure there is no provision for any uplift force, even with a load factor
on the live loading.
The hinge segment is therefore made up of two
half-segments, as shown in Figure 4.75:
The bearing half (reference A), which is connected
by prestress to the shorter part of the span
The carried half (reference B), connected by prestress to the longer part of the span
Measures are taken to continue cantilever construction through the hinge segment until closure
is achieved at midspan; see Section 4.8.6.
Inclined diaphragms provide an efficient way to
suspend or transfer the reaction through the
bearings into the flanges and webs on both sides of
the box section, Figure 4.75.
One of the largest structures incorporating a hinge
segment of this type is the Saint Cloud Bridge, described in Section 3.12. A typical detail of this segment is show n in Figure 4.76.
Neoprene bearings -
?
zyxwvutsrqponm
I
FIGURE 4.71. Deck of variable depth, permanent
deck-to-pie]-
.
bearing arrangement.
box pier and where the pier walls are continued in
the deck, as shown in Figure 4.73. Here again the
transfer of all symmetrical loads between deck and
pier is simple, and design difficulties arise for unsymmetrical loading. .4t the connecting points A
and R, Figure 4.73, between the top flange and the
vertical diaphragms, the part of the top flange tension load T such as T, induces into the diaphragm
another tension load T,, and both loads result in an
unbalanced diago nal co mpo nent T,, w hich must be
resisted both by the webs and bv special provisions
such as stiffening beams.
4.13.3 E.VD ABUTME,VTS
A special segment will be provided at both ends of
the bridge deck with a solid diaphragm to transfer
torsional stresses to the bearings, as shown in Figure 4.74. The expansion .joint is, therefore, adequately supported by the end diaphragm on one
side and the abutment wall on the other side.
4.14
Dejections of Cantilever
Camber Design
Bridges
and
Each cantilever arm consists of several segments,
fabricated, installed, and loaded at different points
in time. It is important therefore to predict accurately the deflection curves of the various cantilevers so as to provide adequate camber either in the
fabrication plant for precast segmental construction or for adequate adjustment of the form travelers for cast-in-place construction.
When the structure is statically determinate, the
cantilever arm deflections are due to:
The concrete girder weight
The weight of the travelers or the segment placing
equipment
The cantilever prestress
After continuity between individual cantilevers is
achieved, the structure becomes statically indeterminate and continues to undergo additional
deflections for the following reasons:
206
Design of Segmental Bridges
e
FIGURE 4.72. Temporarv pier and deck connection.
Continuity prestress
1. Cantilever arms.
Removal of travelers or segment placing equipment
2. Short-term continuous deck
3. Long-term continuous deck.
Removal of provisional supports and release of
deck to pier connections
Placing of superimposed loads
Subsequent long-term deflections due to concrete creep and prestress losses will also take place.
Compensation for the following three types of
deflections must be provided for by adequate
camber or adjustment:
It has already been mentioned that the concrete
modulus of elasticity varies both with the age at the
time of first loading and with the duration of the
load (see Section 4.8.7). Deflections of types 2 and
3 above are easily accommodated by changing the
theoretical longitudinal profile by the corresponding amount in each section to offset exactly
all future deflections. A more delicate problem is to
Dejections of Cantilever Bridges and Camber Design
FIGURE
4.73.
Pier segment with vertical diaphragms.
I
50
S E C TIO N
2.2 5
207
accurately p red ict and ad eq u ately fo llo w the
deflections of the individual cantilever arms during
construction. It is necessary to analyze each construction stage and to determine the deflection
curve of the successive cantilever arms as construction proceeds, step by step. A simple case with a
five-segment cantilever is shown in Figure 4.77.
The broken line represents the envelope of the
various deflection curves or the space trajectory
followed by the cantilever tip at each construction
stage.
By changing the relative angular positions of the
various segments by small angles, such as -LY,,
-(Y*, and so on, the cantilever should be assembled
to its final length with a satisfactory longitudinal
profile as shown in Figure 4.78, for the simple case
co nsid ered . The p racticalities o f this im p o rtant
problem are covered in Sections 11.4 and 11.6.
t
b
S E C T I O N c-c
AmA
I
,
-?
2.25
S E C TIO N B-B
I
+
FIGURE 4.74. Outline of end segment over abutm e n t.
Design of Segmental Bridges
208
3.43zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
SECTIOS
A-A
S E C TIO N C-C
6 30
bl.CTIOS
B-B
FIGURE
4.75.
COUPE A.A
20.40
FIGURE
A
4.00 r
4.76.
Saint
ELEVATION
J
CIOL~CI
joint.
B
P
&
COUPE B.B
1
zyxwvutsrqponmlkjihg
Hinge segment \vith espmsion
4.00
Br-idge, hinge segment \cith expansion joint
It is interesting to compare the relative importance of deflections and camber for cast-in-place
and precast construction. Figure 4.79 shows values
for an actual structure, where computations have
been made for the two different methods. The cal-
culational assumptions given in Figure 4.79 indicate that in m o st c ases the d if f erenc e would b e
even more significant if a cast-in-place cvcle of less
than one week were emploved and if precast segments w ere stored for more than tw o w eeks. Hove-
zyxwvu
Segments N’
ENVELOPE OF
DE’YECTION
CURVES
FIGURE
4.78.
Choice and control of camber.
FIGURE
4.77.
Deflections of a typical cantilever.
t
Cl
_z
CROWN
45.00
SUPPORT
c
zyxw
I
d+5
d+b
db24
d+24
de7
d+25
i
d*8
dt25
~
d+9
~ de26
/
.
I
1
I
d+lO
d+ll
d+26
d+27
dtl2
‘dc27
I
I
I
I
1
E
2
L
-.-
I /
ASSUMPTIONS
PRECAST
_ __ __ __ __
:
in
:
g
Casting : one segment per day
Placing : two segments par day
Segments at least 2 weeks old for
E
Y
LL
x
3
-
placmg
CAST-IN-PLACE :
, -----_---_______
Casting : one segment per week
P r e s t r e s s i n g : 30 days after casting
- 4
-
-L-
FIGURE
4.79.
Comparison of deflections between precast and cast in place structures.
209
Design of Segmental Bridges
210
ever, one would normally expect a cast-in-place
cantilever arm to resist deflections two or three
times greater than the precast equivalent.
4.15
Fatigue
in
Segmental
Bridges
Basically, prestressed concrete resists dynamic and
cyclic loadings very well. Eugene Freyssinet demonstrated this fact fifty years ago. He tested two
identical telegraph poles under dynamic loading.
One was of reinforced concrete and the other of
prestressed concrete; both were designed for the
same loading conditions. The reinforced concrete
member failed after a few thousand cycles, while
the prestressed concrete member sustained the
dynamic load indefinitely (several million cycles).
Fatigue in concrete itself has never been a problem in any known structure, because a variation of
compressive stress in concrete may be supported
indefinitely. When reference is made to fatigue in
prestressed
concrete, it is alw ay s inferred that
fatigue problems arise in the prestressing steel or
conventional reinforcing steel as a result of cracking due either to bending or to shear. If cracking
could be avoided in prestressed concrete structures, the fatigue problem would be completely
eliminated.
Figure 4.80 shows the resistance to fatigue of
prestressing strands currently used in prestressed
concrete structures. The diagram shows the limit
of stress variation causing fatigue failure versus the
mean stress in the prestressing steel. For convenience, both values are expressed as a ratio with
respect to the ultimate tensile strength. For a steel
stress of 60% of the ultimate the acceptable range
of variation is 28% of the ultimate for a number of
cycles between lo6 and 10’. Using, for example,
270 ksi quality strand, this variation is therefore
222,000 psi or a total range of 44,000 psi.
Because dynamic loading on a bridge is of a
short-term nature, the concrete modulus is high
and the ratio between steel and concrete moduli is
of the order of 5. Consequently, the maximum
concrete stress in an untracked section that would
cause a fatigue failure would be 44,000/ 5 = 8800
psi, a value which is probably ten times the stress
variation under design live loads in highway box
girder bridges. An untracked prestressed concrete
structure is therefore completely safe with respect
to fatigue, regardless of the magnitude of live
loads. A limited amount of cracking, although consid ered unad v isable f-rom a c o rro sio n p o int o f
view, is not critical if kept under control.
Tests and experience show that a grouted prestressing tendon can transfer bond stresses up to
.6 -
4-
Stress variation
causing failure
fs ? Afs
2-
FIGURE
4.80.
Resistance to fatigue of prestressing strands.
Fatigue in Segmental Bridges
500 psi to the surrounding concrete. Taking the
example of a typical (twelve 3 in. diameter strand)
tendon with an outside diameter of 2.5 in. (64
mm), a stress variation of 40,000 psi in the steel
produces a tendon force variation of 73,000 lb (33
mt), and the bond development length across a
crack is then 73,000/ (500
x 2.5 x 7r) = 18 in. (0.46
m), see Figure 4.81. The ,corresponding crack
width l is equal to the elongation of the prestressing steel between points A and B w ith the
triangular stress diagram-that is, 40 ksi over an
average length of 18 in., or
40
' = EL = 26,000
x 18 = 0.028 in. (0.7 mm)
A safe crack width limit of 0.015 in. (0.4 mm) can
be accepted to eliminate the danger of fatigue in
the prestressing steel. In fact, instances of fatigue
in segmental structures are extremely few and far
between.
An isolated case has been reported of a bridge in
Dusseldorf, Germany, where failure occurred as a
result of fatigue of prestressing bars. The cast-inplace structure was prestressed with high-strength
bars coupled at every construction joint. After ten
years of service, a joint opened up to # in. (10 mm)
and caused bar failures at the couplers. An investigation revealed that a bearing had frozen and prevented the structure from following the longitudinal movements due to thermal variations. This
accidental restraint induced high tensile stresses in
the concrete and caused cracking, which first appeared in the construction joints precisely where
bar couplers were located. The live-load stress level
in the prestressing steel increased from 850 psi (6
nw,”
I
MPa) fo r the p rev io usly untracked sec tio n to
14,000 psi (96 MPa) for the cracked section and
induced failure in the bars. A recommendation was
made as a result of this fatigue problem that couplers should be moved at least 16 in. (0.40 m) away
from the construction joints and that reinforcing
steel should be provided through the joints if
practical. Another sensitive factor relating to
fatigue in web reinforcing steel was mentioned in
Section 4.10.2 for reinforced concrete test beams.
No such danger would exist in prestressed concrete if shear and diagonal stresses were kept
within the limits that control web cracking.
In conclusion, fatigue in prestressed concrete is
not a potential danger if design and practical construction take into account a few simple rules:
1.
Avoid bending cracks in girders by allowing no
tension or only a limited amount at either top
or bottom fibers for normal maximum loads,
such as the combination of dead loads, prestressing, and design live loads including moment redistribution and half the temperature
gradient.
2. Avoid web cracking by keeping diagonal tensile stresses within allowable limits by proper
web thickness and possibly vertical prestress.
3. Design and maintain bearings and expansion
joints that allow free volume changes in decks.
Temperature stresses that cannot be controlled can give rise to enormous forces that
may either tear the deck apart or destroy the
piers and abutments. In this respect, elastomeric bearings, which work by distortion and
cannot freeze, are safer than friction bearings,
which are more easily affected by dust and
weathering of the contact surfaces.
Insofar as crack control in segmental structures
is concerned, it is usually felt in Europe that excessive concrete cover over the reinforcing steel and
prestress tendons does not prevent corrosion but
merely increases the crack width.3 For example,
the typical 2 in. (50 mm) cover commonly used in
bridge decks in the United States is considered extreme in Europe. The 4 in. (100 mm) cover for
concrete exposed to sea water would be a complete
surprise to European engineers.
Several examples of common practice in segmental bridges are given as a simple comparative
reference in Table 4.2.
zyxwvutsrqponmlkj
07./d’
fJ
211
FIGURE 4.81. Fatigue in prestressing steel across a
cracked section.
Design of Segmental Bridges
212
T ABLE
4.2. Concret e Cov er t o Reinforcing St eel
and Prest ress T endons in Europe
Co ncrete
(in.)
co ver
zyxwvutsr
Germany
1) to 2
1t
lf
Description
Reinforcing steel
Outside exposure, tendons
Inside exposure, tendons
France
1
1t
the longest box girder bridge in the Americas as of
this writing. Typical dimensions were given in Section 2.14. This section will deal with some design
asp ec ts o f this p restressed c o nc rete seg m ental
bridge.
Transverse reinforcing steel
Longitudinal reinforcing
steel or tendons
(no rmal atmo sp here)
Co rro siv e atmo sp here
(salt water)
2
Netherlands
lb
1;A
2 to 2;R
4.16
Reinforcing steel and tendons
(normal exposure)
Lightweight concrete
Salt water exposure
Provisions for Future Prestressing
4.17.1 LONGITUDINAL BENDING
Each of the four identical cantilever arms is made
up of:
Ten segments 8 ft long (maximum weight 4 15
kips)
Six seg m ents 12 f t lo ng ( m axim u m w eig ht 464
kips)
Thirteen segments 15 ft long
457 kips)
4.17 Design Example
The H o u sto n Ship C hannel Brid g e no w u nd er
construction in Texas, U.S.A., is an outstanding
example of segmental construction and represents
weight
Longitudinal tendons are as follows:
Cantileuer t e n d o n s : 4 2 ( n i n e t e e n 0 . 6 i n . d i a
strands) + 50 (twelve 0.6 in. dia). Twelve additional bars used during construction are incorporated in the permanent prestress system.
Continuity tendons in side spans:
For larger segmental bridges, it may be necessary
to modify the prestress forces after construction.
An example would be a bridge built using cantilever construction where positive-moment (continuity) tendons are added after erection. Or, as
discussed in Section 4.8.6, some tendons may be
released to articulate a joint. In addition to these
adjustments immediately after construction, addi-.
tional prestressing may be required at a later date
to correct for unanticipated creep deflection or for
additional loads such as for a new wearing surface.
In Europe on some bridges spare tendon ducts are
provided for this reason. A reasonable assumption
would be to provide for 5 to 10% of the total prestress force for possible future addition.
Since the tendon anchorages for the spare ducts
are inside the box girder and generally located at
the web-flange fillet, they are readily accessible. If
future prestressing is needed, it is only necessary to
insert the required tendon in the duct, jack it to its
designed load, anchor and grout it. Since all this
work can be done inside a box girder, it is not necessary to interrupt traffic, and the workmen are
fully protected.3
(maximum
20 (twelve 0.6 in.
dia).
Co ntinuity tendo ns in center span:
40 (twelve 0.6 in.
dia).
A typical layout of the cross section was given in
Figure 2.82.
The main loading combinations considered in
the design are summarized in Table 4.3. The lonT ABLE 4.3. Houst on Ship Channel Bridge, Main
Design Load Combinat ions
Loading Case
Description
A llo w ab le
Tension on
Extreme
Fiber, Top
or Bottom
(ksf)
(1)
(G) + (P) + (E)
(2)
CD) + (P) + CL + 1)
(D) + (P) + (L + I) + l(AT) + (T)
(D) + (I’) + t(L + I) + (AT) + (T)
CD) + (f’) + (W
(3)
(4)
(5)
0
2:
25
25
Notations: (C) girder load, (D) total dead load including
superimposed dead load, (L + I) live load plus impact, (P) prestress, (E) construction equipment, (AT) temperature gradient
of 18°F between top and bottom fiber, (T) temperature and volume changes, (W) wind load on structure.
Concrete strength and stresses:rC = 6000 psi = 864 ksf (42.1
MPa).
Basic allowable compressive stress: 0.4fi = 346 ksf (16.8MPa).
zyxwvuts
Design Example
gitudinal bending of the box girder has been
analvzed using the BC program, which considers
the effects of the creep, shrinkage, and relaxation
at each construction phase. Figure 4.82 show s the
diagram of prestress forces due to cantilever and
continuitv
tendons at two different dates:
After completion of the structure and opening to
traffic (780 days after start’of deck casting)
After relaxation and creep have taken place (4000
davs)
Significant values of the prestress forces are given
in ‘Table 4.4. The variation of stresses in the center
and side spans is show n in the follow ing diagrams
for the corresponding loading cases:
Figures 4.83 and 4.84, all dead loads and prestress
at top and bottom fibers
Figures 4.85 and 4.86, live load and temperature
gradient at top and bottom fibers
It is easily shown from these diagrams that all
stresses in the various sections are kept within
the allowable values mentioned in Table 4.3. The
WEISYT
O F O N E TRLwLm
:
130 I(.
(m m t)
4.82. Houston Ship Channel Bridge, typical
segment layout and longitudinal prestress.
FIGURE
213
T ABLE 4.4. Houst on Ship Channel Bridge,
Significant Values of Prest ress Forces
Prestress Force (kips)
Maximum cantilever
prestress in side
span
Maximum cantilever
prestress in center
span
Maximum continuity
prestress in side
span
Maximum continuity
prestress in center
span
Day
780
Day
4000
Percent
Loss
54,710 51,310
6.2
54,390 49,280
9.4
9,540
8,760
8.2
18,130 16,780
7.5
maximum compressive stress at the bottom fiber
level appears in the section located 124 ft from the
pier and is equal to 335 ksf under the combined
effect of all dead and live loads and prestress.
4.17.2 REDISTRIBUTION OF M OM ENTS
The exceptional size of the structure gives rise to a
moment redistribution of particular importance.
The BC program allows a complete analysis of the
behavior of the structure under the separate and
combined effects of loads and prestress; also the
effect of concrete creep and steel relaxation can be
considered separately.
Figure 4.87 shows the variation of stresses at top
and bottom fibers along the center span between
days 780 and 4000, which correspond to bridge
opening date and the time when materials will have
stabilized (concrete creep and shrinkage having
taken place and prestress having reached its final
value). The magnitude of the variation is remarkable, particularly at bottom flange level where it
exceeds IO0 ksf (700 psi or 4.90 MPa).
To isolate the effect of concrete creep on moment and stress redistribution, a section near
midspan may be analyzed where cantilever prestress is neglibile. Results for the section located at
a distance of 352 ft from the pier are summarized
in Figure 4.88. The redistribution moment is equal
to 52,000 ft-kips.
It is interesting to compare this result, obtained
through the elaborate analysis of the BC program,
with the result of the approximate method outlined in Section 4.8.7. Figure 4.89 shows the moments in a typical cantilever under girder load and
final prestress. The prestress moment has been
computed using a reduced eccentricity obtained by
zyxwv
Design of Segmental Bridges
214
!
T OP
fc
TOP FIBER
o(-)
HIDSPAN
4.83. Houston Ship
Ship Channel
Channel Bridge,
Bridge,top
topfiber
fiber prestress
prestress for
for (LX.)
(LX.) tt (P)
(P) at
at time
time
780 days and 4000 days. Stresses at top fiber of the deck. Dead load at time 780 days \vhen
the bridge is just opened to traffic and at time 4000 days.
FIGURE
4.173 STRESSES AT MlDSPAlV
transforming the steel area in the concrete section.
Therefore, the prestress moment is equal to:
Pe( 1 - 7zP)
where e = geometric eccentricity,
n = 10, transformed coefficient,
p = percentage of prestress steel in the section (varying between 0.5 and 0.7%).
The total midspan moment produced in the continuous span with fixed ends under the combined
effect of girder load and final prestress is equal to
84,000 ft-kips. Therefore, the actual redistribution
moment obtained by the BC program is equal to:
52 000
) = 62% of the total moment
84,000 -
The recommendation given in Section 4.8.7 to take
a ratio of 2/ 3 gives a satisfactory approximation.
Because of the moment redistribution the bottom
fiber near midspan is subjected to increasing tensile stresses while the top fiber is always under
compression. It is therefore sufficient to consider
the state of stresses at the bottom fiber after creep
and relaxation.
The results are shown in Table 4.5. It is instructive to compare the relative magnitude of the various factors influencing the stresses at midspan
(stresses in ksf at bottom fiber):
1. Live load
44
Moment redistribution
(difference between 250 for
CL and 159 for prestress)
3. Temperature gradient
4. Temperature fall
91
2.
48
18
Design Example
T ABLE 4.5. Houst on Ship Channel Bridge, St resses
at Midspan
Bottom
Partial
Stresses (ksf)
.Moment
redistribution due
to GL
Slomenr r e d i s t r i b u t i o n d u e
10 presr ress
Uoment redistribution due
to (GL) + (P)
All dead loads and all final
prestress (from BC program including moment
redistribution)
Fiber
Cumulative
are light in comparison with those used in other
countries, particularly in France and Great Britain.
These two factors tend to increase the importance
of moment redistribution in relation to the effect
of loads computed in the conventional manner.
+250
4.17.4 S H E A R
-159
+ 91
Live load + impact
~Teniperarure gradient,
AT = 18°F
Temperature
tall,
-66
-
The variation of shear stresses along the center
span under design loads is given in Figure 4.90 together with the corresponding longitudinal compressive stress at the centroid.
The most critical section is located 187 ft from
the pier centerline. The numerical values in this
section are as follows:
44
1.
48
18
T = -40°F
Loading combination (‘L),”
(D) + (P) + (L + 1)
Loading combination (4),’
-22 Max
+22 ( 2 5 )
(D) + (P) + $(L + I) + AT + T
“See loading combinations in Figure 4.85.
Tombination
differential is
215
of Maximum A T + T (maximum temperature
improbable in winter).
2.
3.
Vertical dead-load shear force: 4350 kips.
Resal effect: the compressive stress at the centerline of the bottom slab is 192 ksf and the
angle with the horizontal is 0.055 radians.
Bottom slab area: 53.5 sq ft.
Resal effect: 192 x 53.5 x 0.055 = 570 kips.
Net dead-load shear: 3780 kips.
Live-load shear force: 430 kips.
Corresponding shear stresses in this section:
I/Q = 14 ft web thickness
b=4ft
The influence of the temperature fall (effect 4) is
imputable to the frame action between deck and
piers and would not appear in a conventional deck
resting on its piers with flexible bearings. Considering only the other three factors combined, as in
loading combination (4) of Table 4.3, the
maximum tensile stress at the bottom fiber of the
midspan section is:
9 1 +44+48= 1 5 9 k s f
2The live-load stress is only 44 ksf or 44/l 59 = 28
percent of the total.
In all good faith, a design engineer would have
completely overlooked effects 2 and 3 only a few
years ago and consequently underdesigned considerably the continuity prestress. The situation
has now completely changed, and the knowledge
of materials together with the powerful tool of the
computer allows segmental structures to be designed safely and realistically.
It is as well to remember that the Houston Ship
Channel Bridge is of exceptional size (which tends
to increase the importance of dead load and moment redistribution) and that American live loads
Total shear stress under design load (no load
factor) :
V = 3780 + 430 = 4210 kips
Shear
stress:
4.
5.
4210
v = ~ = 75.2 ksf
14 x 4
Longitudinal compressive stress:f; = 160 ksf
Vertical prestress. The contract specifications
called for a vertical prestress for the entire
deck giving a minimum compressive stress of:
6.
3q = 232 psi = 33.5 ksf
Verification of allowable shear stress.
Using the formula proposed in Section 4.10.4:
u = 0.05fi + 0.2Of* + 0.40fy
the allowable shear stress is:
Vlll,, = 0.05 x 864 + 0.20 x
160 + 0.40 x 33.5 = 88.6 ksf
while the actual shear stress is only 75.2 ksf
zyxwvutsrq
D esign of Segmental Bridges
216
fc
BOTTOM
B O T T O M Fl0ER
w
FIGURE 4.84.
Houston Ship Channel Bridge, bottom fiber stresses for (DL) + (P) at
time 780 days and 4000 days. Stresses at bottom fiber of the deck. Dead load at time 780
days when the bridge is opened to traffic and at time 4000 days.
7.
Principal stresses at design loads for the state
of stress:
u = 75.2,
fJp =
160,
and
fu =
33.5 ksf
The two principal stresses are 3 (tension) and
195 (compression).
The angle of the principal stress with the horizontal is given by:
Corresponding
shear
stress:
VU = 102 ksf
Principal
stress: - 23 (tension) and 217 (compression).
Direction of the principal stress given by:
tan p = 0.56
tan p = 0.466
If vertical prestresses were not used, the principal stresses would become:
-30 (tension) and 190 (compression)
8.
Principal stresses at ultimate stage.
For the load factors 1.30 + 2.17L, including
the effect of prestress, the ultimate shear force
is:
V, = 5710 kips
Web shear cracking at this level of stress would
be unlikely. Assuming that the concrete carried
none of the ultimate shear across the potential
crack shown in Figure 4.91, the total shear load
should be resisted by the vertical tendons and the
conventional stirrups acting on a length equal to:
‘x
Q
1
-=&=25ft
tan/ 3
.
The unit force per foot of girder is therefore:
2 17
Design Example
TEMPERATURE
t 18-F
L IVE
\
/
MAXI
MINI
//
TOP FIBER
\
GRADIENT/
LOAD
\
/
\
\
L IVE
\
375
FIGURE
4.85.
375 FT
FT
LOAD MAxI
4
Houston Ship Channel Bridge, top fiber stresses for (L + I) and (AT =
18°F).
5710
- = 228 kips/lineal ft
25
shear force per unit length of girder to be carried
across the crack is:
The ultimate capacity of tendons and stirrups is:
Tendons in three webs
Stirrups-O.88 in.Vineal
per web at 60 ksi
ft
220 kips/lineal ft
158 kips/lineal ft
278 kips/lineal ft
1 x 5710
- x 0.5 = 240 kips/lineal ft
0.85 0.14
The corresponding amount of steel (grade 60)
would be for each web:
L,2!&
The condition V,/C#I < V, becomes:
228
- = 268 < 378 kips/lineal ft
0.85
and is easily met.
If no vertical prestress had been used, the slope
of the shear crack would be:
tan /3 = 0.487
Using the limiting value tan /3 = 0.5 instead of the
actual value (as explained in Section 4.10.4), the
3
1.33 in.*/lineal ft
This amount of steel would still be reasonable
(0.7%).
4.17.5
DESIG,V
OF THE CROSS- SECTION FRAM E
Owing to the magnitude of the project, particular
attention was given to this problem. Five finite
element analyses were performed to analyze:
The local effects in the transverse frame,
Design of Segmental Bridges
218
$BOTTOM
L I V E
L O A D [::,:
40 KSF
1
LIVE LOAD
“ Axi 1
I
L I V E L O A D MAxi
H I DSPAN
,/’ \ TEMPERA&7E
\\
\
/I
375
FIGURE
= 18°F).
4.86.
G R A D I E N T (+18-F) ,-j
I
375
FT
FT
\
‘.
- A - - - -
Houston Ship Channel Bridge, bottom fiber stresses for (L + I) and (AT
The possible differential deflections
three webs of the box section,
between
the
The relative behavior of sections close to the piers
or at midspan,
The effect of diaphragm restraint near the pier.
The dimensions of the cross section at midspan are
given in Figure 4.92 with the nine critical sections
where moments and axial loads were computed for
as many as fourteen loading combinations.
A typical set of results is shown in Figure 4.93 for
the midspan section. For the section located 187 ft
from the pier centerline (already considered for
maximum shear stresses), the moments and axial
loads are substantially the same as for the midspan
section. Excluding the vertical prestress, the most
critical loading arrangement gives the following
values at the upper section of the outside web (section e of Figure 4.92).
Moment 1 I .9 kip-ft/ ft
Axial load 5.4 kip/ ft
The steel section required at design stage for grade
60 steel stirrups is 0.34 in.2/ lineal ft. Applying the
recommendations of Section 4.10.4 for the simple
case of a section without web prestress, the req u irem ents f o r steel o n b o th f ac es o f the w eb
would be:
For shear of the longitudinal member:
3
x
1.33 = 0.67 in.2/ lineal f t
For bending of the transverse member:
0.34 in.2/ lineal ft
Quantities of Materials
A
f
219
TOP GIRDER LOAD
STRESS
VARIATION
AT TOP FIBER
Af
D
z
B O T T O M PRESTRESS
STRESS
VARIATION
A T BOTTOM
8
FIBER
c_-----v_
Af
FIGURE 4.87.
relaxation.
BOTTOtl
GIRDER LOAD
Houston Ship Channel Bridge, variation of stresses due to creep and
.The minimum area should thus be the higher of
the f’ollowing values:
0.67 + 1 x 0.34
= 0.84 in.*/ lineal ft
1 x 0.67 + 0.34
= 0.67 in.*/ lineal ft
0.i(0.67 + 0.34)
= 0.71 in.2/ lineal ft
In the actual structure, the stirrups in this section
are #6 bars at 12 in. centers, giving on each face a
steel area of 0.44 in.* together with the minimum
v ertic al p restress o f 44.2 kips/ lineal f t ( av erag e
compressive stress of 230 psi).
‘I‘he ultimate capacity of the section reinforcement is theref-ore:
With vertical prestress: 378/ 3 = 126 kips/ lineal ft
Without vertical prestress:
kips/ lineal ft
2
x
0.84
x
60 = 101
4.18 Quantities of Materials
Before closing this chapter, it is interesting to give
some statistical results concerning the quantities of
m aterials req uired in seg m ental b o x g ird er
bridges. Unit quantities have been computed by
d iv id ing the’ to tal q u antities fo r the b rid g e
superstructure by the deck area, using the total
width of the prestressed concrete structure. The
Design of Segmental Bridges
220
Srresses,
Stresses, ‘Top Fiber (ksf)
Loading Case
Bo tto m Fiber- (list)
780 Days
4000 Dqt
780 Dny.\
4000 lkJ\
Cantilev er Prestress
Girder + superimposed
dead load
- 6.36
- 56.93
130.32
-266.50
-20.20
61.89
- 161.08
293.50
Total
Variation from
780 davs to 4000 davs
-63.29
- 136.18
4 1.69
132.42
+9o.i3
-72.89
No& I: .I‘ensile stresses are positive.
Note
2: This moment is the difference between girder load, 142,000 tt-kips, and cantilever prestress, 90.000 t’t-kips.
f, = (I =
72.89
Corresponding moment variation:
AM = (f, ffd +
4774 FZ4)
= (72.89 + 90.73) F
-in
AM = 52,000 ft-kips
II
4
2
cd
II
c?
A
I+?
fz= ++ 90.73
fz=
FIGURE 4.88. Houston Ship Channel Bridge, analysis of section at 352 ft from pier.
average concrete quantity per span foot varies with
the span length. For each structure considered, the
span length used is the average span of the various
two-arm cantilevers. The longitudinal prestressing
steel is given in pounds per cubic yard of deck concrete versus the same span length. It is assumed
that prestressing tendons are made up of strands
with 270 ksi guaranteed ultimate strength. From
the charts given in Figures 4.94 and 4.95, it may be
seen that the average quantities of materials ma)
be represented by the following approximate formulae:
Concrete (ft3/ ft2)
= 1.0 + (L/ 250)
Longitudinal prestress (lb/ ft”)
spans up to 750 ft)
2- 1.0 + (L/ 60) (for
4.19 Potential Problem Areas
As with any type of construction with any material,
problems arise that require the attention of not
only the designers, but contractors and subcontractors as well. No matter how good the design, if
FIGURE 4.89. Houston Ship Channel Br-idge,
computation of moment redistribution.
rapid
221
Potential Problem Areas
I60
.----
I
fx(I
- . zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
.\ \
‘3
140
60
120
/
IO0
40
zyxwvutsrq
EF
PG
80
60
i
.20
40
20
20
I 24
FIGURE 4.90. Houston Ship Channel Bridge, variation of web shear stress and average compressive stress in center span under design load.
the structure is not properly constructed, there will
be problems. Conversely, no matter how diligent
the contractor, if the design details are poor,
problems will result. Obviously, if the design and
the construction are poor, problems are compounded.
8
d 1zyxwvutsrqponmlkjih
110"
I’
13s’
7
i
I
FIGURE 4.91. Houston Ship Channel Bridge, shear
and principal web stresses in section 187 ft from Pier
(under design loads).
di
FIGURE 4.92. Houston Ship Channel Bridge, design
of transverse frame at midspan.
Design of Segmental Bridges
222
b
Sectio n
M, dead lo ad
M,
-6.29
2c37
d
-6.05
1e22
16.59
13.22
- 0.92
8.01
3.01
0.22
0.08
10.92
-5.24
6.93
-6.68
1.45
4.23
5.88
-2.93
- 1.25
-2.88
~ 1.25
2.14
5.03
1.96
-8.82
4.11
- 1.75
7.98
-9.51
11.87
-4.55
-4.50
0.06
50.75
-0.53
51.06
-0.65
- 0.59
51.26
51.35
4.24
-0.31
6.08
-0.31
0.55
- 0.29
50.8 1
50.53
50.61
50.76
3.93
5.77
50.76
1.10
5.36
5.77
Prestressing
M, DL + PIT
M, live load
f
-5?67
with 1 M 1 maxi
M, DL + PIT + LL
-3.15
h
2.14
i
-5.29
0
0.35
0.06
- 5.23
- 0.78
2.59
-6.25
0.26
-
0.37
-
0.24
0.26
0.37
0.24
-:96
+I
N, dead load
N, transverse
prestressing
N, DL + PIT
N, live load
N,D L +PIT+LL
+I
50.8 1
50.53
50.61
Web vertical prestress is not included.
,Vote:
ab
4:
II
i
-------.
F4 \-=- \
I
\
c
1
II‘ensile c rac ks b ehind tend o n anc ho rag es,
particularly for high-capacity continuity tendons in the bottom flange of box sections.
4.
Transverse cracking or opening of Joints, or
adjacent thereto, due to the combination of
several factors such as:
Compressive axial forces are
positive. Positive bending mom ents c au se tensio n at the
f -\ B--------1
3.
broken
line
a.
face.
I
c . W arp ing o f seg m ents d u e to im p ro p er
curing
procedures.
CONVENTION
FIG URE
ments and
4.93.
H o u sto n
Ship
C hanne l
Brid g e ,
mo-
Sev eral su c h p o ints hav e b een alread y a d dressed in this chapter; others are discussed in
C hap ter 11. Sho u ld the rec o m m end atio ns
given be followed both in design and constructio n m etho d s and in su p erv isio n, no m o re
difficulties of this nature are to be expected.
axial forces in transverse frame at midspan.
Problems are generally associated with quality
control, poor design details, or a la c k of understanding as to how the structure will behave, either
through ignorance or because a particular phenomenon is unknown to the current state of the
art, or a combination of all these factors. The following list of problem areas, as they are known to
the authors, is presented so that those involved in
designing and building segmental bridges may
ta ke adequate measures and precautions to avoid
these problems.
1.
redistribution
b. Thermal gradients in the box section.
9
SIGN
Underestimation of moment
due to concrete creep.
Improper performance of epoxy due to mishandling of mixing and application procedure,
particularly in rain and cold weather. The consequences are largely reduced by the use of
adequate shear keys in webs and in both top
and bottom flanges of the box section.
2. G ro u t leakag e b etw een ad jo ining d u c ts at
joints between segments, particularly in precast segmental construction. Conformity of the
ducts at the joints is a desirable feature if practical. The use of tendons outside the concrete
eliminates this problem.
5.
Laminar cracking in de c k slab or in bottom
flange due to wobble and improper alignment
of ducts at the joints between ad-jacent segments. Such incidents have been experienced
more often in cast-in-place construction than
in precast construction. However, care should
always be taken insofar as deck alignment is
concerned in all segmental projects.
6.
Freezing of water in ducts during construction,
esp ec ially tho se anc ho red in the d ec k slab
(vertical prestressing tendons or draped continuity
tendons).
7.
Exc essiv e fric tio n in d uc ts d ue to w o b b le.
Proper alignment will reduce friction factors in
segmental construction to those currently observ ed in c o nv entio nal c ast-in-p lac e posttensioned
construction.
8.
Im p ro p er su rv ey c o ntro l in seg m ent m anufacture for precast segments as well as in the
field for cast-in-place segments.
i
I
100
200
300
400
500
600
AVERAGZ
FIGURE
4.94.
Average quantities of
de c k
700
600
SPA N L ( ft)
c o nc re te .
zyxwvutsrqponmlkjihgfe
15
I
T
FIGURE
/
4.95.
AVERAG E SPAN
L(ft)
Average quantities of longitudinal prestressing steel.
223
224
Design of Segmental Bridges
,
References
1. F. Leonhardt, “ New Trends in Design and Construction of Long Span Bridges and Viaducts (Skew,
Flat Slabs, Torsion Box),” International Association for Bridge and Structural Engineering,
Eighth Congress, New York, September 9-14,
1968.
2. Jean Muller, “ Ten Years of Experience in Precast
Segmental Construction,” Journul of the Prestressed
Concrete Institute, Vol. 20, No. 1, January-February
1975.
3. C. A. Ballinger, W. Podolny, Jr., and M. J. Abrahams,
“ A Report on the Design and Construction of Segmental Prestressed Concrete Bridges in Western
Europe- 1977,” International Road Federation,
Washington, D.C., June 1978. (Also available from
Federal Highway Administration, Offices of Research and Development, Washington, D.C., Report
No. FHWA-RD-78-44.)
4. “ Effets de I’effort tranchant.” Federation Internationale de la Precontrainte, London, 1978.
Foundations, Piers, and Abutments
5.1
5.2
I N T R O DUC T I O N
LOADS APPLIED TO THE PIERS
5.6.2
5.6.3
5.3
5.2.1
Loads Applied to the Finished Structure
5.2.2 Loads Applied During Construction
S UG G E S T I O N S O N A E S T H E T I C S O F P I E R S A N D
River Piers and Foundations for Choisy-le-Roi,
Courbevoie, and Juvisy Bridges, France
Piers and Foundations of Chillon Viaducts, Switrer-
5.6.4
land
Main Piers and Foundations of the Magnan Viaduct,
5.6.5
France
Main Piers and Foundations for the Dauphin Island
5.6.6
Bridge, U.S.A.
Deformation and Properties of Piers with Flexible
5.6.7
Legs
Elastic
A BUT M E N T S
5.3.1 structure Layout
5.3.2 Aesthetics of Piers
5.4
5.3.3 Aesthetics of Abutments
MOMENT RESISTING PIERS
T I ONS
AND
THEIR
FOUNDA-
5.4.1
Main Piers for the Brotonne Viaduct, France
5.4.2 Piers and Foundations for the Sallingsund Bridge,
5.7
5.7.1
5.7.2
with
THEIR
Flexible
Legs
STABILITY
5.8
Saint Jean Bridge In Bordeaux, France
Review of the Various Methods of Providing Stabili t y Dur i n g Can t i l e v e r Co n st r uc t i o n
A BUT M ENT S
5.8.1 Scope
5.8.2 Combined Abutment/Retaining Wall
Upstream Paris Belt Bridge, France
Properties of Neoprene Bearings
No t at i o n s
5.8.3
Separate End Support and Retaining Wall
5.8.4 Through Fill Abutment
5.8.5 Hollow Box Abutment
5.5.4
Deformations of Neoprene Bearings
Deformation of Piers with a Double Row of Neop
rene Bearings
5.8.6 Abutments Designed for Uplift
5.8.7 Mini-Abutment
5.5.6
Properties of Piers with a Double Row of Neoprene
Be a r i n g s
Influence of Thickness and Arrangement of Neoprene Bearings on the Variation of Force in a
Three-Span Structure
PIERS WIT H T Wl N
FLEXIBLE LEGS
5 .6 .1 Inttoduction
5.1 Introduction
Probably the area most challenging to the civil engineer is that of foundation design and construction, presenting the largest potential dangers but
DURING
Scope
Description of Representative Structures with Tem-
5.5.3
5.5.5
5.6
5.7.3
Scope and General Considerations
5.5.2 Description of Structures
Oberon Viaduct, France
Blois Bridge, France
AND
Piers
po’;uy Supports
Downstream Paris Belt Bridge, France
5.4.4 Main Piers for the Houston Ship Channel Bridge,
U.S.A.
PIERS WITH DOUBLE ELASTOMERIC BEARINGS
5.5.1
PIERS
of
C O N S T R UC T I O N
Denmark
5.4.3 Concept of Precast Bell Pier Foundation for the
I-205 Columbia River Bridge, U.S.A.
5.5
FLEXIBLE
Stability
5.9
EFFECT OF DIFFERENTIAL
TINUOUS DECKS
5.9.1
SETTLEMENTS ON CON-
Effect of an Assumed Pier Settlement on the
Stresses in the Superstructure
5.9.2 Practical Measures for Counteracting Differential
Settlements
REFERENCES
also yielding the most significant savings to proper
design concepts or refined construction methods.
The first industrial application of prestressed concrete was related to solving an insurmountable
problem of foundation underpinning.
225
226
Foundations, Piers, and Abutments
The transatlantic terminal built in Le Havre
Harbor in France on the English Channel was
opened for operation in 1934 to receive the new
generation of fast passenger ships between Europe
and America. Improper foundation of the rear
bays of the new building caused immediate constant settlements at the rate of 1 in. (12.7 mm) per
month with no foreseeable limit, except the total
ruin of the facility, Figure 5.1. Eugene Freyssinet
proposed a unique system of underpinning, which
was immediately accepted and implemented,
whereby prestressed concrete piles were manufactured in the basement of the existing building
in successive increments and progressively driven
by hydraulic ja c k s to reach the stable lower soil
strata, found at a depth of more than 100 ft (30.5
m), Figure 5.2. This example should certainly
make one cautious against excessive optimism in
foundation design; at the same time it exemplifies
the remarkable potential of prestressed concrete in
solving unusual problems.
In concrete bridges, often greater savings may
be expected from optimization of foundation and
pier design than from the superstructure itself.
This chapter will deal with certain specific aspects
of piers, abutments, and foundations for bridges
built in balanced cantilever. Similar concepts may
be extended to cover other construction methods
(span-by-span, incremental launching, and so on).
Piers with many different shapes have been used
in conjunction with cantilever construction. For
example, single piers, double piers, and momentresistant piers have all been used. The cantilever
segmental construction method has an important
influence and bearing on the design concept of the
structure. Resistance and elastic stability of piers
during construction require careful investigation.
Temporary piers or temporary strengthening of
permanent piers or a combination of both have
been used. However, the choice of piers that have
adequate stability without temporary aids is highly
desirable. Piers of a box section, or twin flexible
legs, either vertical or inclined, are equally satisfactory.
The use of full continuity in the superstructure
implies that proper steps have been taken to allow
for volume changes (shrinkage, creep and thermal
expansion) at the supports. Bridges such as the
Choisy-le-Roi (Section 3.2), Courbevoie (Section
3.2), and the Chillon Viaduct (Section 3.6) show
how the use of piers with flexible legs makes it possible to achieve full deck continuity and to build
frame action between d e c k and piers without impairing the free expansion of the structure. The
converging pier legs used at Choisy-le-Roi reduce
and even cancel the amount of bending transferred to the pier foundations. Vertical parallel
legs such as those in the Courbevoie and Chillon
-__zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
iI
FIGURE 5.1. Le Havre transatlantic terminal, typical section.
227
Introduction
Jaws Ior str.efchmngE
Steel Rods.
mj
E n d P l a t e of Mould
zyxwvuts
Horizontal Section
b insu/at/nq
fnveiope,
Sheet Iron
Internal Uould.
internal
Moo/U
FIGURE 5.2. Le Havre transatlantic terminal. (a) Vertical section and plan of
composite foundation girder. (6) Details of pile mold.
structures may be used on multispan structures because their ad d itio nal flexibility acco m m o d ates
larger horizontal displacements. For longer structures, bearings with a variable number of laminated elastomeric pads may be used to provide the
desired horizontal flexibility.
If in the finished structure single slender piers
are designed solely to transfer the deck loads to the
foundations (including horizontal loads), the piers
may be unable to resist the unsymmetrical moments due to the cantilever construction (i.e., with
an unbalance of one segment and the equipment
Foundations, Piers, and Abutments
228
load). Thus, temporary shoring is required, often
at considerable cost. In some cases, the stability of
the cantilever under construction has been provided by the launching gantry used for placing the
segments.
With double piers, two flexible legs (either inc lined o r v ertic al) m ake u p the p ier stru c tu re,
which usually is supported on a single foundation.
Stability during construction is excellent and requires little temporary equipment, except for some
bracing between the slender walls to prevent elastic
instability.
Moment-resistant piers are designed to withstand the unbalanced moments during construction by providing a temporary vertical prestress
between the deck and the pier cap, thus producing
a rigid connection. Flat ja c ks are usually placed
between the pier top and the pier segment soffit to
permit the substitution of temporary bearings for
the p erm anent neo p rene p ad s. W hen the ratio
between span lengths and pier height allows it, the
rigid connection and corresponding frame action
m ay b e m aintained p erm anently b etw een the
superstructure and piers.
I
+
42’
I
Piers do not necessarily have to be a massive solid
cross section; a box section, Figure 5.3, mav be
m o re ef f ec tiv e and m o re ec o no m ic al. In the
United States it was generally felt that a solid pier
was more economical. However, for tall piers the
ec o no m ic s o f p ier c asting sho u ld b e ev alu ated
against the cost of the additional dead load supported by the pier shaft and transferred to the
foundations. It may be desirable to precast the pier
as tubular segments that are prestressed vertically
to each other as well as to the foundation; this concept was use d for the Linn Cove Viaduct in North
Carolina and the Vail Pass structures in Colorado.
In certain cases the tubular section may be replaced by an I section, Figure 5.4. However, the low
resistance to torsion of this section imposes certain
precautions to limit the deformation of the cantilevering superstructure during construction, in
particular with respect to the effect of wind forces.
For the case of a continuous structure on short
stiff piers, the volumetric changes of the concrete
(shrinkage, creep, and thermal expansion) compound the redundant effect of longitudinal prestressing to produce, by virtue of the rigidity of the
l
I
_f/q
IO’
13’
p
IO’
FIGURE 5.3. Code Bridge, box pier.
Introduction
+
29’
+
u-
2.
229
Large flexibility in the horizontal plane (relative to the displacements parallel to the longitudinal axis of the superstructure), perm itting the reso lutio n o f the p ro blem o f
expansion posed by the continuous structure,
3. Stability of the superstructure during construction by a simple temporary bracing.
FIGURE 5.4. Pyle Bridge, I-section pier.
piers, bending forces that must be transmitted to
the foundations, thus condemning the use of a
rigid connection between the superstructure and
its support. This disadvantage then requires the
introduction of a continuous superstructure resting on a number of supports that permit the
longitudinal movement of the superstructure (neoprene pads, teflon, and the like). However, it is necessary to insure the stability of the superstructure
during cantilever construction. This may be accomplished as stated earlier by the use of temporary shoring in the proximity of the pier or by providing a temporary fixity at the pier.
Another solution is the use of piers with twin
slender flexible legs. The transmission of horizontal loads in the direction of the longitudinal axis of
the bridge is accommodated by the legs’ flexibility.
This type of pier offers three advantages:
1. Efficient fixity of the superstructure to the
piers with regard to the vertical loads by the
action of the separate supports,
In the final structure, the leg flexibility is sufficient
to accommodate the longitudinal braking forces.
When the geometry of the structure permits, it is
more economical to incline the walls in order to reduce the bending moment transmitted to the
foundation. If the legs are hinged at the superstructure and if the axes of the two legs converge
near the level of the foundation, the bending
moment is either canceled or minimized and the
distribution into the supporting soil is essentially
uniform, as for a vertical reaction, Figure 5.5. This
type of structure is similar to a frame or an arch.
The thrust produced by the effect of a horizontal
load parallel to the longitudinal axis of the bridge
is translated into a tension force on one leg, which
then acts as a tie beam, and a compressive force in
the other leg, which then acts as a strut. For this
reason it is often necessary to prestress the legs to
accommodate the tension force.
When the legs are vertical, they do not profit appreciably from the frame or arch action, and the
stability is essentially contained in their bending resistance. For the case where the legs are hinged at
both ends, no resistance is offered and it is neces-
FIGURE 5.5. Piers with flexible walls.
Foundations, Piers, and Abutments
230
sary to stiffen a pier to provide a fixed point in the
structure.
Because of pier flexibility a careful analysis is required to assure the elastic stability of the structure. The legs supporting the superstructure are in
effect very slender, and their resistance to buckling
must be carefully examined. This type of pier
structure will be examined in greater detail in the
sections that follow.
Another family of piers that lends itself to cantilever construction is that of moment-resisting
piers with a double row of neoprene bearings between the pier top and the superstructure, such as
to benefit from pier rigidity during construction or
in the finished structure while allowing free expansion of the continuous deck, Figure 5.6. The
proper choice of dimensions for the neoprene
bearings will allow control of the amount of bending transferred to the foundation; in fact, rigid
piers with double neoprene bearings behave in
much the same way as piers with twin flexible legs.
We see, then, that piers and foundations for
cantilever concrete bridges will fall into one of the
four following categories:
1.
Moment-resisting piers either fixed or hinged
to the superstructure.
2.
Moment-resisting piers with double neoprene
bearings.
3.
Piers with twin flexible legs.
4.
Conventional
flexible
piers
properly
strengthened during construction to resist unbalanced loading conditions.
FIGURE 5.6. Piers with twin neoprene bearings.
After reviewing the loads applied to the piers
and considering some suggestions pertaining to
the aesthetics of piers and abutments for concrete
segmental bridges, we shall deal separately with
each of the four pier types. The chapter will conclude with a review of several types of abutments
and the effect of unequal pier settlements on the
stress in the superstructure.
5.2
Loads Applied to the Piers
All loads must be carefully considered in the design of the piers and their foundations, both in the
finished structure and during its construction.
5.2.1
LOADS APPLIED TO THE FIAVISHED
STRUCTURE
In addition to the various loading arrangements
taken into account for conventional structures and
used in combination as set forth in the AASHTO
specifications, for example, it is necessary to include some design aspects particular to segmental
cantilever construction as follows:
1. When a frame action is realized between
superstructure and piers, proper transfer of moments to piers must be considered, particularly
under unsymmetrical live loading. The piers are
thus an integral part of the structural system and
their flexibility must be first evaluated and then incorporated in the overall structural system. Figure
5.7 shows the usual parameters used to define the
flexibility of a pier as the relationship between the
applied loads (M, Q, and N) and the corresponding
components of the deformation at the same point
(0, u, and v). The four flexibility coefficients A, B,
C, and K must include all components of the pier
and its foundation: soil, piles (if used), footing, pier
shaft (or walls), neoprene bearings (if used). Loads
and deformations are taken at the level of the deck
girder neutral axis.
The deck construction scheme usually imposes
special loads to the substructure. Piers adjacent to
an expansion joint located at the point of contraflexure (see discussion of this aspect in Chapter
4) are subjected to appreciable bending moments
due both to the relaxation of the hinge after cantilever construction and to live loading placed on
either side of the hinge. Loads applied to the
structure by the construction equipment result also
in moment transfer in piers connected to the
superstructure. Two typical cases often encountered are:
Loads Applied to the Piers
_
APPLIEO
231
LOADS
M,Q,N,
zyxwvut
cOIRESPOH313G
i
0=
-0
OEFOeUATioN
,Mu,W,
AM + BQ
AA r BM+CO
IJ
FIGURE 5.7.
=
KN
Basic components of pier flexibility.
a. In precast segmental construction with segments placed with a launching gantry, the gantry
leg reactions are applied to a temporary static
scheme and released in another static scheme
(after continuity between two adjacent cantilever
arms is realized).
b. In case-in-place cantilever construction, the
weight of travelers is applied to the free cantilever
arms during construction but it is removed from
the structure after continuity is achieved. On long
spans the effect on the deck is usually beneficial,
but important moments may simultaneously be induced.
2. Volume changes (shrinkage and thermal
variations) and long-term shortening of materials
(concrete creep and steel relaxation) both induce
moments and horizontal loads in the piers, which
must be included in the design.
balance), the application of random loads (difference between actual and computed dead loads
or w ind gusts), or accidental conditions (such as the
fall of a traveler).
Normal Loua The most critical condition appears for one segment out of balance at the outboard end of the cantilever arm. Even in the case
of cast-in-place construction with symmetrical
travelers allowing simultaneous casting of both
corresponding segments, the assumption of the
total segment weight out of balance is a safe one,
because no total guarantee can be given that concrete pouring will proceed simultaneously at either
end of the cantilever. If construction equipment is
designed to be installed on the deck, Figure 5.8, it
5.2.2 LOADS APPLIED DURING CONSTRUCTION
Balanced cantilever construction imposes on the
piers a loading configuration that is globally symmetrical. Unbalanced conditions appear, however,
as a result of intermediate construction stages
(normal loads due to a traveler or a segment out of
FIGURE 5.8. Loading conditions during construction.
.
Foundations, Piers, and Abutments
232
must be accounted for in the design of the pier.
For example, a tower crane is often used on one
side of a cantilever.
Random loads essentially are
s u c h a s t o p r o d u c e systematic geometric difference, although within acceptable tolerances.
With proper workmanship and supervision, it is
reasonable to assume such difference in weight at
22%. It corresponds to a variation of top slab
thickness of 2 in. (9.5 mm) for a 40 ft (12 m) wide
box with a cross-sectional area of 60 ft2 (5.6 m’).
However, it is very unlikely that the maximum
weight decrease in one cantilever arm would appear simultaneously with the maximum weight increase in the other. It is therefore reasonable to
limit the moment transferred to the pier to 2% of
the maximum deck cantilever moment due to the
girder weight. Other random loads related to the
construction are produced by the small equipment,
trucks, storage on the deck of materials such as
post-tensioning tendons, and so on. An equivalent
uniform load of 5 psf (24.4 kg/m*), together with a
moving concentrated load of 20 k (9 mt), should be
a safe allowance to cover these random loads.
Taking as an example the Houston Ship Channel Bridge, which was considered in Section 4.17,
the effect of these three random loads would be:
R ando m
Loads
difference in dead weight,
1,600,OOO ft-kips x 2%
random uniform load,
( 5 x 60)/1000 x 365*/2
random concentrated load,
20 kips x 365 ft
32,000
ft-kips
20,000
ft-kips
7,000
ft-kips
59,000
ft-kips
This moment should be compared to the effect of
one segment out of balance at the far end of a cantilever:
300 kips X 367 ft = 110,000 ft-kips
One last source of random loading is provided
by gusts of wind that apply an uplift pressure or
suction to the box girder intrados during construction. For long spans and construction sites exposed to hurricanes, it is desirable to make special
aerodynamic tests. For an incident angle of 10”
above the horizon, the upward pressure would be 5
psf (0.2394 MPa) during construction. This value
may be substantially increased in exposed sites. For
construction of the Gennevilliers Bridge, a
maximum pressure of 9 psf (0.4309 MPa)
corded in the wind-tunnel tests.
was re-
A ccidental Loads These are the result of a construction incident or of human failure, causing
either the fall of a traveler in cast-in-place construction or of the lifting equipment in the case of
precast construction. Such loads should be multiplied by a factor of 2, representing the impact
coefficient for the case of immediate loading. It is
never envisaged to consider the fall of a cast-inplace segment and traveler after casting, nor the
fall of a precast segment immediately after its
placement in the structure. A very long record of
safety in such construction methods justifies that
approach. However, in the case where the consequences of such major accident would be exceptionally disastrous (where, for example, the work
takes place over a highway or a railway under operation), special provisions should be incorporated
in the design and in construction procedures to
double all safety features at each step of erection.
5.3
Suggestions on Aesthetics of Piers and
Abutments
The problem of aesthetics is subjective and controversial. There is, however, a consensus among
engineers, owners, and users that certain bridge
structures are more pleasing than others. At a time
when so much emphasis is being placed on protection of our environment and of nature from aggressive man-made structures, it may be helpful to
review some ground rules based on experience
that contribute to aesthetics of concrete bridges
with very little added cost.
53.1
STRUCTURE
LAYOUT
Generally speaking, an attempt should be made to
match the structure to the environment and to preserve the existing landscape. Avoid long, high embankments at the ends of the bridge as well as long,
high retaining walls that accentuate the intrusion
of the new structure. Allow the number and shape
of the piers to maintain a maximum of transparence. Cost optimization of superstructure span
lengths will normally help to avoid serious aesthetical mistakes. It is equally disgraceful to see a heavy,
long-span superstructure
rampant
over the
ground as a multitude of closely spaced, high piers
supporting a slender deck floating up in the air.
The true appearance of a structure is usually not
233
Suggestions on Aesthetics of Piers and Abutments
conveyed by the drawings, where often a distorted
scale is used for convenience.
Finally, it is very important to keep the unity of
appearance of a structure crossing different obstacles, in spite of the practical difficulties that may be
entailed when project coordination involves different owners or agencies. When an overpass crosses,
for example, a freeway and a parallel railroad
track, nothing may be worse than to build two
separate structures (probably of different height)
connected by a short embankment contained at‘
both ends by wing walls of variable height, Figure
5.9.
FIGURE 5.10. Piers for the Broronne
duct.
,~pp~o;~ch
via-
5.3.2 AESTHETICS OF PIERS
A significant advantage of segmental construction
is to allow deck continuity, rather than simply supported structures. There is no longer a need for
heavy bents protruding underneath the superstructure soffit. Piers can have simple graceful
lines and be designed to receive directly the box
girders of the superstructure.
Box piers of prismatic section but with curvilinear shapes improve the appearance over the
conventional rectangular section. The approach
piers of the Brotonne Viaduct, Figure 5.10,
utilized that concept and also the piers for the Linn
Cove Viaduct in North Carolina. More refined
shapes may be used, such as for the river piers of
the Blois Bridge, Figure 5.11, where the sculpture
of the faces was designed to recall the appearance
of a pier with twin inclined walls similar to that of
the Juvisy Bridge, Figure 3.25. Architectural
studies may be pursued further and reach beyond
the immediate structural needs of the designer. An
interesting example is afforded by the river piers
FIGURE 5.9. An unacceptable example of’ an overpass built as two separate structures.
of the railroad bridge at Clichy near Paris, Figure
5.12.
A difficulty arises often for skewed bridges when
bents include multiple pier shafts. A satisfactory
solution was developed for the Paris Downstream
Belt Bridge, Figure 5.13. The four columns of a
river pier are given the shape of a lozenge, with
one axis of symmetry matching the alignment of
the superstructure while two of the four faces
exactly align the four columns in the direction of
the river flow.
FIGURE 5.11.
Blois Bridge.
Piers with architectural shapes for
234
Foundations,
Piers, and
Abutments
FIGURE 5.12. Piers for Clicln Railroad Bridge.
FIGURE 5.13. Piers for a skew bridge (Paris Ring
Ro ad ).
When the piers will be seen only from a great
distance, it is usually not worthwhile to call for a
special treatment of the concrete faces. The eye
will judge only the general shape of the structure
and its overall proportions. For urban bridges
the situation is very different and often justifies
some architectural treatment of the piers. The river
piers of the Saint Cloud Bridge were cast with a
system of closely spaced vertical grooves, which
greatly enhance their appearance at very little
added cost, Figure 5.14.
5.3.3
AESTHETICS
OF
ABUTMENTS
At both ends, the structure has to blend with the
existing landscape with a minimum of disturbance.
Between the two systems of wing walls shown in
Figure 5.15, the preference should strongly be
with type (a), which allows a much more gradual
transition between the lines of the superstructure
and those of the approach embankment.
When tapered webs are used in the superstructure box girders, it has been found that the lateral
wing walls in the abutments can be given the same
FIGURE 5.14. Saint Cloud Bridge. (CL)
(b) General v iew .
River piers.
inclination to improve the transition between deck
and abutments, Figure 5.16.
5.4
Moment-Resisting Piers and Their
Foundations
We shall cover this topic by describing salient features of several characteristic structures.
5.4.1
MAIN PIERS FOR THE BROTONNE
VIADUCT, FRANCE
The two main pylon piers for the Brotonne Viaduct rest on 41 ft (12.46 m) diameter cylindrical
235
Moment-Resisting Piers and Their Foundations
Wing walls parallel to bridge Q
-.__.-. -.---.-.-.-.-
-
!I
- _-.-.-
-
__.-.__ -.-.-.-.Wing walls perpendicular to bridge F.
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
(b)
FIGURE 5.15. Wing w alls and abutments.
:i ”
::.*i .:
columns with a maximum wall thickness of 9.3 ft
(2.83 m) and are 115 ft (35 m) below ground level
in a limestone stratum overlain by alluvium, silt,
and gravel beds. The maximum reaction at footing
level is 19,000 tons. Typical dimensions of a main
foundation syst.em are show n in Figure 5.17.
It was decided to select the theoretical foundation level at 115 ft (35 m) below the original
ground level, where the limestone bed had the
following minimum characteristics determined
from laboratory soil tests and in situ tests: angle of
internal friction 20”, cohesion 5 tons/ ft2, and a
pressure limit (on triaxial tests) of 45 tons/ ft2. The
foundation system had to resist very large loads
(both vertical and horizontal) together with important overturning moments.
The main foundation column embedded in the
soil and resting on the lower limestone stratum was
analyzed as a rigid body subjected to the applied
loads (M, V, and H) shown in Figure 5.18 and receiving from the soil lateral reactions along the
shaft and vertical reactions under the base. Values
of lateral and vertical reactions were ascertained
for the various soil strata and the equilibrium was
determined by considering the total body to be
subjected to an angle of rotation cy around the in-
.*\q*>
-.
.,I
\. ,
. >..:~ ..:.\ ‘,;,.,
,.j \‘;:i\ ;\;cA. :&
\
. .>;.y
FIGURE 5.16. Inclined wing walls in end abutment
(Bordeaux St. Jean Bridge).
stantaneous center of rotation C. The coordinates
of point C are the following:
Vertically, it represents the level where lateral
reactions from the soil change sign (change from
direct passive pressure on the front face to counterreaction at the back face).
Horizontally, it is the position of the neutral axis
for the stress under the base.
The maximum loading configuration is represented numerically in Figure 5.18 along with the
diagrams for:
Lateral reactions on the column
Bending moments along the column
Bearing stress under the base
If there were no lateral support, the bending moment at the base would have been 370,000 ft-kips.
In fact, the actual moment is only 130,000 ft-kips,
236
Foundations, Piers, and Abutments
FIGURE 5.17. Brotonne Viaduct, pylon foundations.
which explains why the extreme fiber stress is no
more than 24 tonsIft while the average bearing
pressure is 14.25 tons/ ft2.
The actual safety factor for the foundation
against soil failure is betw een 3 and 4, depending
on the assumptions of soil characteristics.
Insofar as the construction method is concerned,
each main foundation column was built in the dry
inside a cofferdam made up of a continuous slurry
trenched concrete wall excavated down to the
limestone stratum, Figure 5.19. Grouting of the
base allowed dewatering of the site after excavation
to inspect the foundation material and confirmation of the actual soil characteristics by in situ soil
tests. Following this inspection, the cofferdam was
flooded and a tremie seal was placed at the base to
prevent any risk of washing out of the footing concrete due to water seepage; the water head was
above 100 ft (30 m). The reinforced concrete
footing was cast in the dry above the seal and the
foundation shaft was then slip-formed inside the
cofferdam. The pier shaft was given the shape of
an octagon with curvilinear sides for aesthetic reasons. The general dimensions of the foundation
shaft and of the pier shaft allowed a very natural
and direct transfer of loads at ground level with no
need for a heavily reinforced footing. The construction of both foundations went very satisfactorily. The only incident was created by the fact
that one panel of the cofferdam in the south pier
was excavated out of plumb at its lower end. Consequently, the continuity of the horizontal ring to
resist the hydrostatic pressure was not realized at
the lower part of the cofferdam. Grouting of the
surrounding soil w as achieved in this area and an
additional reinforced concrete ring was cast inside
before the completion of excavation and final dewatering.
Regular survey measurements at the site have
shown that settlements of both pier foundations
have been very minimal and are now stabilized.
5.4.2
PIERS AND FOUNDATIONS FOR THE
SALLINGSUND BRIDGE, DENMARK
The substructure and piers of this structure present an interesting construction methodology and
use of materials, Figures 3.89 and 5.20. The piles
are steel tubes, which are concreted after driving.
Their length is about 98 ft (30 m), the diameter is
Moment-Resisting Piers and Their Foundations
n. 230.050
FT. k.
v, 19.000
237
t
zyxwv
TSF
zyxwvu
1
“““‘J
FIGURE
of main
,.--Jre
5.18. Brotonne Viaduct, loads and soil reactions on column
foundations.
274 in. (700 mm) and the wall thickness is about 0.4
in. (10 mm). Each pier has 24 piles. The first piles
driven are tested in compression and tension be-
FIGURE 5.19. Brotonne Viaduct, view of pier excavation.
fore the remaining piles are driven. When the
driving is accomplished, the template trough is
filled with tremie concrete around the pile tops up
to the upper edge of the template.
The template is precast at a plant located in the
harbor. It is shaped like a circular slab surrounded
by an annular trough, in which there are holes for
the piles. The template is transported to the pier
locations by the floating crane and lowered down
to rest on three temporary vertical piles. The bottom is about 52.5 ft (16 m) below the water level.
For an exact positioning in its submerged position,
it is provided with an alignment tower, the top of
which is always above water, Figure 5.21.
The pier box, shaped like a truncated cone approximately 39.3 ft (12 m) high, is precast in three
lifts at the precasting plant in the harbor. First its
lower part is cast on staging above water. During
the following lifts it is progressively sunk. Since
after the third stage it is too heavy to be lifted by the
floating crane, it is provided with a lid, and com-
Foundations, Piers, and Abutments
238
Assembled
ure 5.23. The finished bridge is shown in Figure
5.24.
pier
5.4.3zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQP
CO,VCEPT
OF PRECAST BELL PIER
FOC,VDATION
FOR THE I- 205 COLL’,WBI,4 RIVER
BRIDGE, C’.S.A.
/-y-7
Gmcmting
r
T e mp l a t e
Il.3
Concreting
of piles
concrete
FIGURE 5.20. Sallingsund
A somewhat comparable system to that used for
the Sallingsund Bridge was contemplated for approach spans 15 through 26 of the I-205 Columbia
River Bridge in the State of Oregon, as shown
schematically in Figures 5.25 and 5.26. Steel H
piles of 200 ton capacity were to be driven through
a template box, allowing tremie concrete to be
placed inside the trough. The precast segments
were designed to be stacked upon one another
above the template to make up the pier shaft and
transfer the superstructure load to the piles.
This scheme was not actually used, as the contractor decided on a more conventional method of
construction. However, the scheme of precast bell
pier foundations was used on the Richmond-San
Rafael Bridge and the San Mateo-Hayward
Bridge, both in San Francisco Bay, and the Columbia River Bridge at Astoria, Oregon. A comprehensive discussion of these structures is presented by Gerwick in reference 3.
plug
Bridge. schematic of sub-
StrllCtlll-e.
pressed air is pumped into the cavity. The floating
crane then transports the pier box to the pier location and lowers it down to rest on the template. A
reinforced concrete ring structure is made by connecting the pile tops to the pier box by reinforcing
and concreting the space between them, Figure
5.21.
The icebreaker’s shell is a reinforced concrete
box, precast at the harbor site, Figure 5.22, transported to the pier location by means of the floating
crane and placed on top of the pier box. Its top is
then 8.2 ft (2.5 m) above and its bottom 8.2 ft (2.5
m) below the water level. When the box is in place,
the water in the cavity of the pier box and the icebreaker box is pumped out. Next, the piles are filled
with concrete and the pile tops and the lower part
of the pier box are cast together. Finally the cavity of
the icebreaker is filled with concrete. A schematic
sequence of operations in constructing the substructure is shown in Figure 5.21.
Piers are cast in place in lifts 10 ft (3 m) high by
means of climbing forms and are hexagonal, Fig-
5.4.4
M AI,: PIERS FOR THE HOUSTOS SHIP
CHA,V,VEL
BRIDGE, U.S.A.
Each main channel pier, Figure 5.27, is made up
of the following:
A rectangular shaft 161 ft (49 m) high with a cross
section varying in dimensions from 20 X 38 ft (6.10
x 11.60 m) at the base to 20 x 28 ft (6.10 X 8.50 m)
at the top. The section is a single-cell box with wall
thicknesses of 2 ft (0.61 m).
A reinforced concrete footing 75 X 81 X 15 ft
(22.90 x 24.70 x 4.60 m).
A group of two hundred and twenty-five 24 in.
(0.61 m) diameter steel pipe piles having a wall
thickness of 4 in. (12.7 mm).
The superstructure is completely integral with the
two main channel piers to form a rigid frame, both
during construction and in the finished structure,
Figures 1.67 and 2.80.
Stresses in the concrete and reinforcing steel
were analyzed in both stages with the service-load
design approach, a n d u l t i m a t e s t r e n g t h w a s
verified by the load-factor method. The analysis is
Moment-Resisting Piers and Their Foundations
239
PLACING OF TEMPLATE
dteel pik
TEST LOADING
PILE DRIVING
PLACING OF PIER BOX
PLACING OF ICE BREAKER BOX
CONCRETING OF STEEL
PILES AND FOUNDATION
FIGURE
tio ns.
5.21.
Sallingsund Bridge, schematic of substructure opera-
rather strenuous, because in the completed structure only there were 19 unit loads combined into
37 load combinations for service-load design and
into 42 loading combinations for load-factor design.
The concrete cross-sectional area together with
the corresponding reinforcing steel area is as follows:
top:
A , = 176 ft2,
A , = 200 no. 11 bars = 297 in.*,
p = 1.17%
bottom: A, = 216 ft2,
A, = 264 no. 11 bars = 392 in.*,
fi = 1.26%
Under service load the average concrete stress of
the cross section is as follow s:
top:
bottom:
3 1,700 kips + 176 ft* = 180 kips/ ft*
36,600 kips + 2 16 ft2 = 170 kips/ ft*
In large structures, such as the Houston Ship
Channel Bridge, the average concrete stress in the
pier shafts usually varies betw een 160 and 200
kips/ ft*. The use of a varying-width pier in the
transverse direction allows the maximum stress
and the required amount of reinforcing steel to increase at a slow rate with the pier height, while a
prismatic pier shaft will be subjected to a very critical stress at the base.
FIGURE 5.22. Sallingsund Bridge, aerial view of precast yard and harbor for substructure construction.
CLIMBING
FORM
Tower crane
Ice-breaker-
:’
.
Pierbox
CROSS SECTION
0
04Om
3 50m
55Om
FIGURE
5.23. Sallingsund Bridge, schematic of pier construc-
tio n
FIGURE 5.24. Sallingsund Bridge, view of finished
brid ge.
241
Piers with Double Elastomeric Bearings
TYPICAL PIER
5810’
,
TYPICAL
I_ S E G M E N T 5 _
I
I
I
MER
PROFILE
116’-2”
CAST-IN-PLACE
CONCRETE
DESIGN HIGH WATER
ELEV.
28.0’
-
SEGMENT
4
SEGMENT
3
//
\\
I
r
SEGM
SEGM
FIGURE 5.25. I-205 Columbia River Bridge, main
piers and foundations.
P I E R ELEUTION
PLA.h S E C T I O N B.B
PLAN SECTION 4.A
PRECAST BELL PIERS
SEGMENT
4
FIGURE 5.27. Houston Ship Channel Bridge, main
river piers.
5.5
5.5.1
SEGMENT
SEGMENT
SEGMENT
FIGURE 5.26. I-205 Columbia River Bridge, schematic of construction of precast bell piers.
Piers
with
Double
Elastomeric
SCOPE AND GEAVERAL
Bearings
CONSIDERATIO,VS
Recognizing the inherent advantages of a rigid
connection between piers and superstructure (stability during construction and increased superstructure stiffness reducing the effect of live load),
the designer is rapidly limited in its use in long
bridges because of unacceptable effects of volume changes. This situation allowed the birth of
a new type of structure developed to maintain the
two desirable features that were previously contradictory: flexural rigidity on one hand and horizontal flexibility on the other. The concept of the
double row of elastomeric bearings was first developed for the Oleron Viaduct and used thereafter on a great many bridges.
242
Foundations, Piers, and Abutments
With piers of this type, two observations are required concerning the transfer of forces between
the superstructure and pier. The first observation
concerns the transfer of service loads, Figure
5.28~. Under the effect of unsymmetrical loads,
the upper and lower flanges of the superstructure
are respectively subjected to unequal tension forces
TL and T, and compressive forces CL and CR. If a
vertical diaphragm is positioned over each of the
two rows of bearings, the center portion of the top
flange of the pier segment, to be in equilibrium,
must accept the tension force TL - TR. This is not a
satisfactory disposition, as the thickness of the
flange and amount of reinforcing have to be increased between the two rows of bearings, and
there is the risk of cracking.
However, if the two diaphragm: are inclined and
converge at the level of the top flange, the differential in tension, T, - T,, is divided into two components of force, C (compression) and T (tension),
directed into the plane of the diaphragm, while the
tension force may be accommodated by prestressing the diagonal bracings.
Another important aspect of the pier segment
design relates to the imbalanced
loading condition
resulting at the bottom flange from the unequal
reactions R1 and R, of the bearings, which calls for
careful analysis of the stress developing in the
diagonal bracings in all loading stages of the
structure.
The second observation concerns the superstructure-pier connection during the temporary
phase of constructing the superstructure in cantilever, Figure 5.286. To accommodate a moment unbalance resulting from the construction
procedure, the pier segment is supported on four
temporary bearings of steel or concrete, 0, and
temporarily fixed by prestressing to the top of the
pier, 0. After closure at midspan occurs, producing
a continuous span, the joint is “ unlocked” by releasing the prestressing. Flat jacks, 0, are then activated so as to substitute permanent bearings for
the temporary bearings.
5.5.2 DESCRIPTION OF STRUCTURES
Many structures have been designed and built
utilizing the system of piers incorporating a double
row of neoprene bearings. This section will describe the salient features of three particular
bridges as exemplifying the advantages of this system as used in connection with a variety of foundation schemes.
TL
-
Neoprene
bearings
CO++
i‘3
Flatjacks
Steel banded
concrete block
fb)
FIGURE 5.28. Connection of superstructure and
pier. (a) In service. (6) In temporary construction phase.
Piers with Double Elastomeric Bearings
Oleron Viaduct, France
Of the 45 piers, only the 27 piers supporting the
center portion of the viaduct with span lengths of
260 ft (79 m) are designed with a double row of
bearings. In this portion of the viaduct there is an
expansion joint every fourth span, and the elastomeric bearings had to accommodate the volume
changes of the deck in a maximum distance of
three spans (i.e., 780 ft or 237 m). Out of these 27
piers equipped with a double row of bearings, 12
are founded on spread footings constructed directly on limestone rock inside a temporary sheet
pile cofferdam, Figure 5.29. The other 15 piers are
supported by a system of pipe piles driven to the
limestone, which in this area is at a depth of 75 ft
(23 m) below mean water level, Figure 5.30.
The 12 piles in each pier consist of four vertical
piles, one at each corner, and eight battered piles,
so inclined as to resist the horizontal loads (longitudinal and transverse) applied to the structure.
For the most critical loading combination (comparable to the AASHTO requirements) the maximum load in a pile is 330 t (300 mt), which
should be reduced to compare to American prac-
FIGURE 5.29.
243
tice by a factor of 1.33. The comparable design
load would then be 250 t (230 mt) for a pipe pile 20
in. (500 mm) in diameter with a thickness of 3 in.
(12.7 mm) driven to refusal in the rock and filled
with concrete after driving. The corresponding
steel stress of the pipe alone would be 16 ksi (110
MPa), a somewhat higher value than normally used
in similar circumstances. When considering the
global section of concrete and steel, the stress in the
concrete is only 800 psi (5.5 MPa)-a very reasonable value, confirmed by the fact that none of the
15 piers showed any sign of settlement during
the fifteen years of operation of this viaduct. The
pipe piles were driven open-ended and excavated
inside by a homemade airlift system conceived
by the driving subcontractor. It took only a few
minutes to perform this operation on each pile.
For the piers on piles, a tremie seal was used inside the cofferdam to allow dewatering and construction of the reinforced concrete footing
poured in the dry.
All box pier shafts were slip-formed to a
maximum height of 82 ft (25 m) at the rate of 15 to
20 ft (4.5 to 6 m) a day, and the construction of a
shaft took approximately one week, Figure 5.31.
Oleron Viaduct, piers on spread footings.
Vertical
0.30
prestresslng
1.30
.
.-.
.
.
..’
5
.
6
0
-
.,
p-i-
+ zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
Treme concrete
,‘..
I,-*
I.I’ :
9’
c
r
FIGURE 5.30. Oleron Viaduct, pier-s on piles.
244
7.30
l
The special feature of this project is that a very
comprehensive optimization study of the substructure system with a double row of bearings allowed the use of only half as many piles as the basic
scheme with single bearings, without increasing the
unit bearing capacity of the piles.
Upstream Paris Belt Bridge, France
FIGURE
5.31.
Oleron Viaduct, aerial view of founda-
tions.
Blois Bridge, France
The Blois Bridge crossing the Loire River is a fivespan, prestressed, precast concrete segmental
superstructure consisting of twin box girders with
the following span dimensions: 202, three at 300,
202 ft (61, three at 91, 61 m). It is supported by
f o u r r i v e r p i e r s elastically restrained at the
superstructure with a double row of bearings. Dimensions of a typical pier are given in Figure 5.32
and a view of a finished pier in Figure 5.11.
65'
t
I
I
t
I
This important bridge was built over the Seine
River to carry Europe’s most heavily traveled
urban freeway, the Paris Beltway. As shown in a
longitudinal section, Figure 3.22, it has two major
river piers resting on a unique foundation system,
while land piers and abutments are conventionally
founded on piles.
A typical transverse section of the bridge shows
the orientation of the piers, Figure 5.33, and various cross sections through the piers is shown in
Figure 5.34. Each of the twin bridges carries four
lanes of traffic on two box girders, which are supported on two separate pier shafts connected below
water by a single footing. Two lower foundation
shafts extend under this footing to a maximum
depth of 70 ft (21 m) to carry the bridge loads to
the supporting soil strata through a series of
heterogeneous seams of silt, fine sand, and clay.
Each of these lower shafts (there are eight such
shafts for the two river piers) w; s built inside a rectangular steel sheet pile cofferdam, driven as low
as possible before excavation. The shafts were extended below the tip of the sheet piles to reach the
load-bearing soil by incremental stages of excavation and continuous concrete lining, Figure 5.35.
Cement grouting and temporary lowering of the
aquifer by pumping allowed this work to be performed in the dry. Except for the minor blowout in
one of the eight shafts, which called for special
grouting work, the foundation project was performed safely and successfully. Figure 5.36 shows
one of the river piers completed and receiving the
precast pier segment of the superstructure.
5.5.3 PROPERTIES OF NEOPRENE BEARINGS
Notation
A neoprene bearing may be designated by the following physical parameters, Figure 5.37a:
a and b
12
t
2e
FIGURE 5.32.
Blois Bridge, dimensions of river piers.
= plan dimensions of bearing (a < 6)
= number of elastomer sheets
= thickness of one elastomer sheet
= thickness of the internal steel sheet
(twice the external sheet)
Ab = a * b = area of bearing
Foundations, Piers, and Abutments
246
EL
C c-
124
zyx
D t-
zyxwvutsrqpon
34’
3 4’
1
1zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
1
? 4’
1
T
FIGURE 5.33.
Upstream Paris Belt Bridge, typical elevation of river piers.
An example, with dimensions in millimeters, is as
follows:
a x 6 x n(t + 2e)
300 x 400 x 2(10 + 2)
Where differing thickness of steel plates are used,
the successive thicknesses of steel and elastomer
are given:
aXbXn(
)
300 x 400 x 2(5 + 8 + 2 + 8 + 1)
The relationship between Young’s modulus (E)
and the shear modulus (G) is presented in Table
5.1. The shear modulus, G, of neoprene varies not
TABLE 5.1. Elastic Constants
45
50
55
60
65
Vertical Defo rmatio n (Co mpressio n) Under a normal force V every lamination is subjected to a vertical shortening, v, Figure 5.376, such that:
v=C
Deformation of Neoprene Bearings
Hardness
(IRHD ?4)
only with the material hardness, as indicated in
Table 5.1, but also with the rate of loading. Tabulated values are for the case of slow loading; for an
instantaneous loading the value of G is doubled.
Young’s
Modulus E
(N/mm*)
Shear
Modulus G
(N/mm*)
1.80
2.20
3.25
4.45
5.85
0.54
0.64
0.81
1.06
1.37
t3
V
z&-z
C is a coefficient that depends on the plan dimensions of the bearing and that expresses the restraint effect on the lamination by the steel plate;
refer to Table 5.2.
For a bearing consisting of n stacks or laminations, the value of the shortening is equal to:
(5-l)
Ro tatio nal Defo rmatio n U n d e r a b e n d i n g m o ment M the upper face of each lamination undergoes a rotation 8 relative to the lower face:
tI= C’ &M
b
Piers with Double Elastomeric Bearings
247
5ECllDN
: D.-D
I
I
FIGURE
5.34.
Upstream Paris Belt Bridge, typical horizontal sections of river piers.
C’ is a coefficient that depends on the plan dimensions of the bearing and that expresses the restraint effect on the laminations by the steel plate;
TABLE 5.2.
bla
C
0.5
5.83
0.6
4.44
0.7
3.59
0.75
3.28
0.8
3.03
0.9
2.65
TABLE 5.3.
bla
0.5
C’
136.7
0.6
116.7
0.7
104.4
0.75
0.8
100.0
0.9
96.2
90.4
refer to Table 5.3. The value a is the dimension in
plan of the bearing measured perpendicular to the
axis of rotation, Figure 5.376.
Values of the Coefficient C
1.0
2.37
1.2
2.01
1.4
1.78
1.5
1.70
2
1.46
3
1.27
4
1.18
5
1.15
10
1.07
30
1
30
Values of the Coefficient C’
1.0
86.2
1.2
80.4
1.4
76.7
1.5
2
3
4
5
10
75.3
70.8
66.8
64.9
63.9
61.9
60
Foundations, Piers, and Abutments
FIGURE 5.35.
Upstream Paris Belt Bridge, detail of concrete lining of lower shafts.
For a bearing consisting of n sta c ks or laminations, the value of the rotation is equal to:
Horizontal Deformation (Distortion) Under a horizontal force, Q, the upper face of each lamination,
relative to the lower face, undergoes a horizontal
displacement u :
with a corresponding distortion u/t.
For a bearing consisting of n sta c ks o r laminations, the value of the horizontal displacement is
equal to:
5.5.4
FIGURE 5.36.
finished pier.
Upstream Paris Belt Bridge, view of a
DEFORMATION OF PIERS WITH A DOUBLE
ROW OF NEOPRENE BEARINGS
In structures where deck and piers are rigidly
fixed, it is necessary to analyze accurately the deformation of ihe various piers to incorporate their
proper stiffness into the model of the total structure. This is particularly important for unsymnietrical live loading applied to one pier and for the
effect of volume changes. There is a relationship
between the loads applied at the top of one pier
(usually at the level of the neutral axis of the deck
over the pier) and the corresponding displacements at the same point that depends solely upon
the mechanical properties of the pier and its foun-
Piers with Double Elastomeric Bearings
249
s ~(arbj
(a)
(b)
rdhep-cp
C” * hBp-‘p
(4
Cc)
‘n: -
cf)
FIGURE 5.37. Piers with double row neoprene bearings (Oleron Viaduct). The most
lcidely used polychloroprene is Neoprene (trademark of Du Pont de Nemours).
dation, Figure 5.7. The elasticity coefficients A, B,
C, and K may be computed from the material
properties and dimensions of the pier.
For example, a pier with constant section and the
following properties:
Height h, area of cross section A,
Moment of inertia I
Modulus of elasticity E
duces a partial fixity of the superstructure on the
piers. The neoprene bearings intervene in the defo rm atio n o f the p ier by their no rm al fo rce
( 2 M lpd) produced by the moment M, Figure
5.37~. The rotational stiffness of the neoprene
bearings may be neglected.
The moment M applied at the top of the pier
may be divided into componentsf and m in the
bearings, Figure 5.37d, such that:
M =fd + 2m,
assumed to be fixed at the base onto a totally rigid
foundation, has the following elasticity coefficients:
A=&,
h*
B = 2EI’
e
= 2vld
w ith:
h3
C=m,
B*
h
K=A-C=4EI
from which:
In structures where neoprene bearings are
placed between piers and deck, the corresponding
change in elasticity of the system must be taken
into account. In fact, the presence of two rows of
neoprene bearings, spaced at a distance d, pro-
In the majority of cases the quantity 2a*/ C’ is small
relative to d */2C.
250
Foundations, Piers, and Abutments
Example
Dimensions of the neoprene bearing: 600 X 600
mm. Spacing between the axes of the neoprene
bearings: d = 2.4 m.
-=
b 1
a
’
C = 2.37,
C’ = 86.2
= + $ (1.215 + O.OOS)e
In neglecting the second term in the parenthesis,
in other words the rotational stiffness of the neoprene, it can be seen that the error is slight, of the
order of 1%. Therefore:
2nCt3
’ = pGAgzd2 M
increase of the moment in the bearings, Figure
5.37e.
During construction of the superstructure by
cantilevering, stability in the temporary construction phase may be provided by the substitution of
concrete pads for the neoprene bearings and the
use of a temporary vertical prestressing.
By a judicious choice of neoprene thickness, it is
possible to reduce the bending moments applied to
the foundation. Consider a pier with a double row
of neoprene bearings supporting a continuous
superstructure. For a bending moment M at the
top of the pier, under the effect of a loading in the
superstructure with no horizontal displacement,
the bending moment transmitted to the base of the
pier is (Figure 5.37f):
M’=M
where h represents the height of the pier. Because
u = 0, one may write:
Accordingly, the flexibility coefficients of the
neoprene bearings may be written as:
A,=(+?
+Qh
BM+CQ=O
from which:
t3
pd2 m
B, = 0
(5-4)
Cn=2LL
2P GAb
wherep represents the number of neoprene bearing s p er ro w . Therefo re, if the flexibility
coefficients of the pier shaft are denoted by A,, B,,
C,, and K,, the total flexibility coefficient may be
defined as:
A =Ap+A,
B =B,
c = c, + c,
K = K, + K,
5.5.5
PROPERTIES OF PIERS WITH A DOUBLE ROW
OF NEOPRENE BEARINGS
Piers with a double row of neoprene bearings have
properties similar to those of piers with flexible
legs, by insuring an effective fixity for loads while
allowing the free expansion of the superstructure.
This fixity presents the advantage of reducing
the bending moments in the spans without much
an d
M’ = (1 - +)M =
(1 - c,B$cn)M
=
4M
The value of the coefficient 4 varies w ith the
thickness of neoprene pads. If it is desired to transfer no moment to the foundation at the level of the
pier base, M’ = 0, the transfer coefficient 4 must
be equal to 0, from which:
C, = hB, - C,
(5-5)
On the other hand if the neoprene thickness
becomes very large, the value of 4 tends to the
limiting value of 1 and the bending moment remains constant in the pier; that is, M’ = M, Figure
5.37f.
As an example, consider a pier w ith a constant
moment of inertia, fixed at its base, with a double
row of neoprene bearings and supporting a
maximum reaction of 1000 tons.
Pier characteristics: Assume a box section with external dimensions of 5.0 x 3.0 m and a w all thickness of 0.30 m, h = 33 m, I = 7 m4:
E/ &,=+=
4.71
Piers with Double Elastomeric Bearings
251
= 2 x 860 x 2 x 160 x 0.24 = o 034 m
3.9 x 106
EB, = & = 77.7
nt = 34 mm
EC, = & = 1715
Four neoprene bearings are arranged in two rows
at a spacing of 2.4 m in the longitudinal direction
of the bridge. Dimensions of each bearing are 600
X 400 X 3(12 + 2) (see Section 5.5.3).
Flexibility of the neoprene bearing: a = 0.40 m, bla =
1.5, C = 1.7, Ah = 0.24 m*, n = 3,p = 2, t = 1.2 X
10m2 m, G = 160 t/ m’, E, = 3.9 X lo6 t/ m’:
A comparison of the constants A, B, C, and K with
the nu m b er o f neo p rene lam inatio ns ( f o r this
example) is presented in Table 5.4. If the height of
pier Were changed from 33 m to 20 m, the total
neoprene thickness would correspondingly change
from 34 mm to 8 mm.
5.5.6 INFLUE,VCE
OF THICKNESS AAiD
ARRA,VGEME,vT OF ,YEOPRE,\‘E
BEARI,XIGS
ON THE
VARIATIOX OF FORCE IS A THREE-SPAN
STRUCTURE
EB, = 0
nt
EC,=E-zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
= 915
In order to better understand the influence of the
2PG A h
Totaljlexibilit?
qf
the pier:
EA = 4.71 + 0.97 = 5.68
EB = 77.7
EC = 1717 + 915 = 2630
Elasticit?
of the pier in the structure:
EK = E (.4 - g) = 5 . 6 8 - ‘ :;;;’ = 3 . 3 8
Elasticity qf’ the pier zuithout neoprene:
Ek’ = E [;4 - $1 = 0.25+ = 1.18
Coe@cieut
of momerlt
transmission in the pier:
4 = 1 - B; - 1 - 77;76;033
= +0.03
The bending moment M’ transmitted to the base
of the pier is very small (3% ofM). For the moment
M’ to be theoretically equal to zero:
thic kness o f neo p rene p ad s, stu d ies hav e b een
conducted to determine the variation of the bending moment in a three-span continuous structure
when only the number of- neoprene laminations at
the top of the intermediate piers is modified.
The stru c tu re c o nsid ered is a sy m m etric
su p erstru c tu re o f three c o ntinu o u s sp ans su p ported on two identical piers; it consists of a box
girder with a variable moment of inertia, whose
spans are 44 m, 70 m, and 44 m.
Bend ing m o m ents in the su p erstru c tu re and
piers are calculated under the following assumptions:
Superstructure fixed at the pier
Sup erstructure p artially fixed elastically at the
piers with neoprene bearings with the varying
lamina of 1, 2, 3, 6, or 9 (thickness 12 mm)
Superstructure supported on the piers by simple
supports
Assumptions used in the conduct of the study are:
EC,, = 9 = 860
and the corresponding
then:
EC,=n
t E
2p G A ,
or
nt = ‘L(EC.)PGA,
E
thickness of neoprene is
Superimposed dead load represented by a uniform
load, q = 1.9 t/ m
Expansion of the deck at a rate of 2 X 10e4, corresponding to an increase in temperature of 20°C.
Shrinkage of the deck at a rate of 4 x 10P4, corresponding to a decrease of temperature of 20°C
combined with the effect of shortening and timedependent deformations (creep) resulting from
prestressing (2 X 10m4).
Foundations, Piers, and Abutments
252
TABLE 5.4.
Number of Neoprene Lamina
Coefficient
EA
EB
EC
5.03
77.7
2020
5.36
77.7
2325
5.68
77.7
2630
6.00
77.7
2935
6.33
77.7
3240
2.03
2.76
3.38
3.93
4.46
1.18
-0.27
Diagram of
bending moment
in the pier
(h = 33 m)
Diagram of
bending moment
in the pier
(h = 20 m)
+ 0.64
Applied load Sz = 4.5 t/ m in the center span
Applied load S, = 6.8 t/ m in the end spans
Braking force F = 15 t on the superstructure, corresponding to approximately one-twentieth of the
structure dead load
The bending moments in the superstructure as a
result of the above loads are tabulated in Tables
5.5a through 5.5~:
Table 5.5a: bending moment at the top of the pier
Table 5.56: bending moment at the base of the pier
Table 5.5~: maximum bending moments in the
superstructure
This study leads us to the following observations:
1. Regarding the superstructure, the maximum
moments vary little with the number of neo-
prene laminations. When the number of laminatio ns inc reases f ro m o ne to six, the
maximum bending moment at the support decreases by 4% and the maximum positive moment in the center span increases by 10%. The
extreme case of nine lamina is to b e avoided
because of risk of instability presented by the
tall sta c k of neoprene (alnt < 5). Compared
with a simple bearing support, the double row
of bearings provides an important decrease in
moment in the spans for a relatively smaller increase of moment at the pier support.
2. Regarding the pier, there exists an optimum
thickness of neoprene allowing a minimal
transfer of moment to the level of the foundations. In the example considered this thickness
is equal to three lamina of 12 mm, which corresponds closely to the value determined in Section 5.5.4 for the case of a structure restrained
ho rizo ntally.
Pi er s wi t h T wi n Fl exi bl e Legs
T A BL E
5.5~.
Bending Moment at t he T op of t he Pier as Funct ion of t he Bearing Thickness0
Number
of’
zyxw
253
Neoprene
Lamina
0
(Fixed
Lo ad ing
Pier)
Su p erstru c tu re D.L.,
q = 1.9
Deck
1
2
3
6
9
+ 124
+
106
+
93
+
84
+
68
+ 58
+
t/111
92
+
68
+
53
+
43
+
27
+
+ 2 x 10-4
Deck
shrinkage,
expansion,
- 184
-
36
-
106
-
86
-
54
- 38
- 4 x 10-4
I: m o m ents I + ,\I
( n o L.L.) -‘VI
+ 216
60
+
174
+
146
+ 127
+
95
+
77
-
30
-
13
-
+
6
+
20
+ 1700
- 1420
+ 1440
+ 1270
+ 1150
+ 930
+ 790
- 1240
- 1120
- 1030
- 850
- 740
? 101
+
2
k
2
k
+2017
- 1581
+1711
+ 1059
+ 1367
+ 1105
+941
- 1367
- 1226
- 1122
- 924
- 795
2
19
L.L in center span,
si = 4.5 t /m
L.L. in end spans,
S, = 6.8 t/ m
Braking force, F = 15 t
Maxi 1llu111
+‘ M
m o m e n t s I - ‘\/ I
“ Values have heen
97
93
90
calculated at the intersection of the axis of the pier with the center of gravitv
T A B L E 5.5b.
80
74
of. the super-structure.
Bending Moment at t he Base of t he Pier as Funct ion of t he Bearing T hickness
N u m b er o f N eo p rene Lam ina
Sim p le
0
(Fixed
Pier)
Lo ad ing
Support,
1
2
Superstructure
D.L.,
q = 1.9 t/ m
-
62
- 31
-
Deck
3
15
-
6
4
+
9
t = 24 mm
13
+ 20
0
expansion
- 202
-157
-129
- 111
- 77
- 6 0
-130
+2 x 1o-4
Deck shrinkage
- 4 x 10-4
+ 404
+314
+258
+222
+154
+120
+260
C m o m ents
+M
( 1 1 0 L . L . )I -M
+ 342
- 264
4283
-188
+243
-144
+218
-115
+167
- 6 4
+ 140
- 4 0
+260
-130
L.L. in center span,
- 820
-435
-198
- 47
+176
+ 265
0
S, = 4.5 t/ m
L.L. in end spans
+ 197
- 7 4
-207
-265
-380
-400
0
S, = 6.8 t/ m
Braking force, F = 15 t
+ 159
2163
*167
-e170
+180
?I86
(+520)
+ 698
- 1243
+609
-786
+577
-518
+558
-550
+527
-624
+591
-626
+780
-650
M ax im u m
m o m e n t sI
5.4
+ M
-M
Piers with Twin Flexible Legs
5.6.1 IN TRO D U CTIO N
The concept of piers with twin flexible legs was first
used with the first match-cast segmental bridge of
Choisy-le-Roi. It was further used on several other
p re c ast se g m e ntal b ridges either in France or
Europe and more recently in the United States.
Several examples of such structures will be described below with particular emphasis on the design and construction methods of the foundation
system.
Foundations, Piers, and Abutments
254
T A BL E
5.5~.
Maximum Bending Moment s in t he Superst ruct ure as Funct ion of t he Bearing T hickness
Number of Neoprene Lamina
Loading
Moments 1 Center span
at
support
Side span
Moments
in
span
Center span,
(0.5 &)
Side span,
(0.4 11)
0
(Fixed
Pier)
1
2
3
6
9
Simple
Supporl
-3125
-3060
-3020
-2985
-2925
-2895
-2660
-3105
-2960
-2845
-2770
-2635
-2545
-2055
+ 910
+ 960
+ 990
+1015
+ 1060
+ 1090
+1270
+ 890
+ 935
+ 965
+ 980
+ 1020
+ 1040
+ 1200
5.6.2 RIVER PIERS AND FOUNDATIONS FOR
CHOISY-LE-ROI, COURBEVOIE, AND JUVISY
BRIDGES, FRANCE
These structures were described in Chapter 3.
Cho isy - le- Ro i Bridge o v er the Seine
This structure is composed of two parallel twin
bridges, Figure 3.3 and 5.38. Each structure has a
continuous three-span su p erstru c tu re in prestressed concrete with spans of 123 ft (37.50 m),
180.4 (55 m), and 123 ft (37.50 m), fixed at the
center piers and forming a symmetric frame.
Piers are supported on a system of steel pipe
piles driven to refusal in rock. The superstructure
is supported on two slender inclined legs having a
thickness of 16 in. (0.40 m) and inclined to the vertical axis at 0.065. Dimensions of the substructure
are shown in Figure 3.3. The precast legs with an
approximate weight of 27.5 ft (25 mt) have their
centerlines converging to a point approximately at
the level of the foundations so as to reduce the
bending moments to .the foundation. The legs are
joined to the body of the pier at one end and to the
superstructure at the other end by prestressing
tendons. Before construction of the superstructure by the balanced cantilever .method, the legs
are temporarily stiffened by a triangular steel
framework in the space between them. The construction stages are described graphically in Figure
cal spans of 131 ft (40 m), 197 ft (60 m), and 131 ft
(40 m).
Each river pier consists of two half-structures
whose foundations are fixed in dense rock, Figure
3.9. The top portion of each half-pier consists of
two vertical slender legs, oriented, in plan, perpendicular to the longitudinal axis of the bridge,
and in a transverse section of the bridge, disposed
in the shape of a V. These legs, w hich have a
parallelogram form, are spaced in a longitudinal
direction at 6 ft 9 in. (2.05 m) on center with a constant wall thickness of 18 in. (0.45 m). The legs
were precast and joined to the superstructure and
the lower portion of the pier by prestressing tendons.
The Juvisy Bridge consists of six prestressed
concrete continuous spans with a total length of
700 ft (213.5 m). Spans are successively from the
left bank 62 ft (18.8 m), 62 ft (18.8 m), 137 ft (41.8
m), 218 ft (66.6 m), 137 ft (41.8 m), and 84 ft (25.7
ml.
The two piers located in the Seine are split piers
resting on a common foundation, Figure 3.26. The
foundations were constructed inside a sheet pile
cofferdam, which permitted the flexible legs to
be fixed at the bottom and hinged at the top. The
thickness of the legs varied from 24 in. (0.60 m) at
their base to 16 in. (0.40 m) at the top. They were
symmetrically inclined at 0.0805 to the vertical and
were cast in place and prestressed.
5.6.3
5.38.
PIERS AND FOUNDATIONS OF CHILLON
VIADUCTS, SWITZERLAND
Courbevoie and Juvisy Bridges over the Seine
The Courbevoie Bridge is very similar in concept
to the Choisy-le-Roi Bridge. It consists of a continuous three-span superstructure with symmetri-
This structure, 1.24 miles (2 km) in length, is a twin
parallel viaduct overlooking Lake Leman and follow ing a sinuous route corresponding to the contour
of the hillside on which it is located, Figure 5.39. It
5
FIGURE
5.38.
Choisy-le-Roi Bridge, construction stages of foundations and piers.
Foundations, Piers, and Abutments
256
1 (y
16-6
FIGURE 5.39. Chillon
Viaduct, general view.
consists of 23 continuous spans of prestressed concrete, span lengths being 301.8 (92 m), 321.5 (98
m), or 341.2 ft (104 m). Four expansion joints divide each viaduct into sections with a maximum
length of 1890 ft (576 m). The longitudinal stability of each section is provided either through a
fixed bearing over the end abutment or by special
fixed piers designed to withstand the horizontal
reactions of the superstructure.
The piers, Figure 5.40, consist of two slender
vertical legs with a constant thickness of 2 ft 8 in.
(0.80 m). Height of pier varies in increments of 26
ft (8 m) with a maximum height of 118 ft (36 m).
Legs less than 72 ft (22 m) in height are hinged at
the top and bottom. Legs over 72 ft (22 m) in
height are fixed at the base and hinged to the
superstructure.
Because of the leg spacing there is no tension
generated in the legs, so no vertical prestressing is
required. During construction of the superstructure the stability of the pier is increased by temporary steel bracing anchored into the legs.
5.6.4
MAIN PIERS AND FOUNDATIONS OF THE
MAGNAN VIADUCT. FRANCE
The Magnan Viaduct consists of four continuous
spans; span lengths are 413 ft (126 m), tw o at 433 ft
(132 m), and 249 ft (76 m), Figure 2.98. The piers
are constructed of twin H-shaped shafts 40 ft (12
m) on center and with a maximum height of 3 18 ft
(95 m) above the valley floor, Figures 5.41~ and
5.41b. These piers are similar to slender vertical
legs of variable cross section fixed at the base. Because this structure is located in an area of seismic
activity, the superstructure is fixed at the west
abutment and restrained transversly at the piers
and the other abutment.
l-
Ia56
-1t
FIGURE 5.40. Chillon
5.6.5
Viaduct, pier section.
MAIN PIERS AND FOUNDATIONS FOR THE
DAUPHIN ISLAND BRIDGE. U.S.A.
The Dauphin Island Bridge is an 18,000 ft (5.5
km) long structure over Mobile Bay connecting
Dauphin Island to the mainland of Alabama. In
order to permit ship traffic, the central portion of
the structure was designed with a three-span continuous unit of 2 11, 400, and 211 ft (64, 122, and
64 m). This provided a clear shipping channel of
350 ft (107 m) horizontally and 85 ft (26 m) vertically. This project is currently (1980) under construction and is anticipated to be completed by late
1981.
Each main pier of this three-span structure consists o f tw in, I-shaped w alls spaced lo ngitudinally at
21.5 ft (6.6 m) on center, Figure 5.42. An individual w all is 24 ft 7 in. (7.5 m) w ide and is
moment-connected to the single cell box girder
superstructure as well as to the footing.
Piers with Twin Flexible Legs
The foundation is to be made with circular,
standard sheet pile construction. Alternate pilings
were detailed on the plans to be either 30 in. (0.76
m) square precast, pretensioned concrete or
54 in. (1.37 m) hollow, cylindrical, precast, posttensioned concrete. Piling will be driven to a
capacity of 450 kips (204 mt) for the 30 in. (0.76 m)
square pile or 550 kips (249 mt) for the 54 in. (1.37
m) cylindrical pile. A dewatering seal will be
poured under water after the piles have been driven. This seal will be located 25 ft (7.6 m) below the
water surface and have a thickness of 5 ft (1.5 m).
After dewatering, a circular footing with a diameter of 44 ft (13.4 m) and a thickness of 10 ft (3.05
m) will be poured. The twin wall piers will be constructed from a point 10 ft (3.05 m) below the
water level and reach a total height of approximately 93 ft (28 m).
The design included checking of AASHTO
loads and combinations, including a stream flow of
3.5 fps (1 mps). Additionally, the structure was
checked at an ultimate condition for a storm wind
of 200 mph (322 km/h). The load factor for this
condition was taken as 1.0.
EL630
21:4*
2f.4"
257zyxwvutsrqponmlkj
i
EL.315
v
, *
EL.266
f
*
‘5’
*
*.
i.1 J
65.60'
)
5.66
D EFO R M A TI O N A N D P R O P ER TI ES O F P I ER S
W ITH FLEXIBLE LEGS
The following notation is used (Figure 5.43):
M, Q, W components of external load acting at
point 0,
m, t, n = components of load acting at the top of
the leg of the pier, oriented to the axis of
the leg,
8, U, z, = displacements corresponding to M, Q, N
at point 0,
W, (Y, /3 = displacements corresponding to m, t, n at
the top of the leg,
E = modulus of elasticity of the concrete leg,
1= length of the leg between points A and B,
2d = spacing of the legs at the top between
points A and A’,
a = cross sectional area of leg,
i = moment of inertia of a leg,
p0 = ad V2i dimensionless coefficient,
4 = angle of inclination of the legs with the
vertical.
Identical and symmetrical legs, of length 1, are
inclined to the vertical by the angle 4. The crosssectional area and moment of inertia of each leg at
FIGURE 5.41. zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
Magnan Viaduct. (n) Pier section. (h)
a distance x from the top, A or A ‘, are respectively
Completed pier.
a(x) and i(x).
258
Foundations, Piers, and Abutments
6
BRIDGE
tf
HER
BRIDGE
SEGMENT
r-h-
24’7’
STEEL SHEET PILING
EL 0 0 0
L SEAL CONCRETE
SECTION
SECTION
The symbol u is designated as an equivalent area
of the leg such that:
24’. 7 ”
I
1
l’ u!x
-=1 s0 a(x)
u
and U, V, and W the characteristic integrals as:
f PIER
UT
’dx
0 i(x)’
v=
s
w =
I--
E BRfDGt-
‘xdx
s0 i(x) ’
‘x2d.x
sO;(x)’
At the level of’the superstructure, AA ‘, the combined area and moment of inertia of the two legs,
designated by A and I respectively, is represented
bv:
A = 2a and I = 2i + 2ad2
PLAN VIEW
FIGURE 5.42. Dauphin Island Bridge, dimensions of
main piers and foundations.
with 2d being the distance between the two legs at
the top.
Setting p. = ad2/2i, the combined moment of inertia of the two legs becomes I = 2i( 1 + 2p,).
Piers with Twin Flexible Legs
*-A---+ mA*
FIGURE 5.43. Piers with flexible legs, notations.
The positive directions of forces and displacements are indicated by the arrows in Figure 5.43.
The deformations of the pier are given by linear
equations that relate the displacements of the top
of the pier (0, U, v) to the applied forces (M, Q, N).
Legs AB and A ‘B ’ are assumed to be connected at
their ends by two rigid and indeformable sections
AA’ and BB ‘. Section BB’ is assumed fixed (no
translation), and the deformation equations are
given by:
8=AM+BQ
u=BM+CQ
v=KN
w here A, B, C, and K rep resent d efo rm atio n
coefficients of the legs.
Force components M, Q, N acting at point 0
(center of AA’) are the resultant of the external
forces applied to the pier, and 8, U, u are the corresponding components of displacement of the section AA’ at point 0 (Figure 5.436). To determine
the forces m, t, n and m’, t’, n’ in the legs atA and A’
requires the fo rm ulatio n o f the eq uatio ns o f
equilibrium, deformation, and compatibility.
1. Equilibrium equutions:
The equilibrium of the
system about point 0 is given by
M = m + m’ + d sin 4(t + t’)
- d cos +(n - n’)
(5-6)
Q = (t + t’) cos 4 + (n + n’ ) sin 4
N = - (t - t’) sin 4 + (n + n’) cos 4
2.
zy
Deformation equations: Displacement o, (Y, P
and w’, cr’, p’ at pointsA and A’ (with respect to
the axis of the legs) are given by:
w=w,+
a=o,,l+
s
‘m + t x
mU
oTdx=o,+E +tvE
mV
tw
‘m + t x
-xdx=cq,l+so Ei
E + E
p=pY=lr
EU
0 a
(5-7)
where w. is the rotation of the leg AB at B, and
E is the modulus of elasticity of the concrete.
Corresponding equations give the displacements o’ , (Y’, / 3’ at point A’.
Displacements of points A and A’ with respect
to the axis of the pier, 8, A, p and 0’, A’, p’ are
determined as
8=6J
A = a cos 4 + p sin 4
p = Q sin 4 + p cos
4
zyxwvu
Foundations, Piers, and Abutments
260
Legs hinged at both ends
A’ = CY’ cos 4 - p’ sin
4
4
For any of these four cases the legs may be of constant or variable cross section, either inclined or
vertical. A comprehensive study was made of this
3. Compatibility equations: The c o nd itio ns o f
problem by J. Mathivat and reported in references
compatibility between the displacements of
1 and 2, with several complete derivations of forpoint A, A ‘, and 0 require that
mulas applying to each particular case.
exoEw ’
(if there are no hinges
An important practical application is that of twin
atA andA’)
vertical walls with constant cross section, for which
eq u atio ns b ec o m e v ery sim p le. Tab le 5.6 su m (5-9)zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
m ariz es the v alu e o f the g lo b al eq u iv alent
coefficients of elasticity of the pier. In this case p,, =
nd2/2i, which becomes p0 = 6(dlh)2 with 2d the disp) = 7l + de
tance on centers of both legs and h the wall thick‘The foregoing equations are sufficient to calness. Usually p0 varies between 30 and 80.
culate 8, U, and 11 as fSunctions
of the applied
It is evident, in fact, that a pier made up of twin
loads represented by ,M, Q, and ,V.
legs behaves much in the same way as a conventional pier with a cross-sectional area A and a moFour practical cases need to be considered:
ment of inertia I insofar as the effect of vertical
loads and moments on vertical displacements and
Legs fixed at both ends
rotation is concerned.
Legs fixed at the superstructure and hinged at the
The behavior is completely different when conbase
sidering the horizontal displacement due to the
application of a horizontal load (braking force or
Legs hinged at the superstructure and fixed at the
thermal expansion). The conventional value of the
base
p’ = CY’ sin
TABLE
5.6
4
+ j3’ cos
Fle xi bi li t y Co e f f i c i e nt s o f a Pi e r wi t h T wi n Ve rt i c al Walls o f Co nst ant Cro ss Section”
End Conditions for Legs
Flexibilit\
Co efficient
E.wct Fonttui~i,s
.4
5lultiplier
Co efficient
1
El
.4pproxitttntr Forttt~clnsh
‘4
1
Fixed .I‘op
Hinged Bottom
Hinged .I‘op
Fixed Bottom
I+1
2P,t
1+1
1
B
C
Fixed stop
and Botrom
13
3EI
1+-p
zyx
Hinged ‘I‘op
rind Bottom
1+1zyxwvutsrqpon
2 PO
2P”
0
0
1 + 2Po
x
(1 + $)(3 + 2/J,,)
I
El
“Notation: I = 2i(l + 2p,), equivalent global inertia of twin walls. p,, = nci2/2i
thickness.
*When l/p, is negligible with regard to 1.
= C(~/IZ)~,
with 2~f distance between walls, h wall
zyxwvutsrqponmlkjihgfe
Piers
1:’
with
Twin
elasticity coefficient C = - IS multiplied by the
3EI
dimensionless factor 1 + po/ 2 in the case of vertical walls fixed top and bottom or by (1 + 2 pO) for
walls hinged at one end.
The elasticity coefficient becomes infinitely large
for double-hinged vertical walls, which proves simply that stability toward horizontal loads must be
o b tained thro u g h so m e o ther restraint in the
structure such as fixed connections or elastomeric
bearings over the abutments.
A detailed study of several typical cases was conducted for the Choisy-le-Roi Bridge, considering
in particular:
The position of the point of contraflexure in the
pier varies very little when the pier is subjected to a
moment only; it is considerably more sensitive to
the effect of a horizontal load.
The horizontal rigidity of the pier varies appreciably with ttie degree of fixity of the legs.
5.6.7 ELASTIC STABILITY OF PIERS WITH
FLEXIBLE LEGS
It has been shown that the use of twin Hexible legs
(whether vertical or inclined) provides an economic solution to the dilemma between rigidity for
bending versus rotation and flexibility for horizontal load versus displacement. In this respect the
elastic stability of the system is the limiting factor,
because there must always be an ample margin
against buckling.
A ssu m e the b rid g e su p erstru c tu re to b e d isplaced horizontally by 11 under a random horizontal load. The resistance against such displacement
is offered by the pier rigidity, including the bending resistance of the legs if they are at least partially
fixed at the top or bottom and possibly including
the horizontal rigiditv of the bearings over the
abutments.
The minimum value of the vertical reaction in
the pier (or the normal force in the legs), for which
the imposed displacement does not have a tendency to spontaneously diminish until the cause
provoking the displacement vanishes, represents
the critical buckling load of the pier. This critical
load is generally smaller than that where the legs
are considered Isolated and subjected to the same
load conditions.
Legs hinged at both ends
Legs fixed on top and hinged at the base
Tab le 5.7 p resents the essential resu lts o f this
studv, which also includes consideration of the
flexibility of the body of the pier to the base of the
foundation, where:
MO = bending moment in the superstructure
at the pier section (side of the center
span),
M, = bending moment in pier (top section),
Q = horizontal reaction in the pier.
The following conclusions may be drawn from
the study:
The superstructure is very efficiently fixed over
the river piers by the twin inclined wall system. The
end moment for the center span totally fixed at
both ends would be 255. The actual end moment
varies between 230 a nd 232 (i.e., 90% of the fixed
end moment).
5.7.
Choisy -le-Roi Bridge: behav ior of Riv er Piers under Horiz ont al and Vert ical Load@
Unit Vertical
Load in
Center Span
Flexibilit)
Coefficients
Type of
Legs
Fixed
Fixed/hinged
Hinged
261
Legs
The elasticity of the pier depends very little upon
the conditions of fixity of the walls at the top and
bottom (0.92 to 1.03).
Legs fixed at both ends
T ABLE
Flexible
Unit Horizontal
Load Applied
to Deck
Unit
Volume
Change
A
B
c
Elasticity
El,
M,,
iv,
Mll
iM1
iv,,
4.06
12.7
54.6
234
-
973
4670
-
0.92
0.98
1.03
-232
-231
-230
-157
-154
-150
+3.4
+5.1
+6.3
+5.7
+a.7
+ 10.7
+7.4
+6.4
+5.9
iv,
Q
+24.7
2.4
+21.5
1.3
+ 19.7 0.9
“Notation: A, B, C = flexibility coefficients of pier. E L = global elasticity of pier. M, = end moment of center span (in tm). ,M, =
bending moment at pier top (m tm). Q = horizontal reaction in pier.
*Units: All coefficients in metric system. A uniform vertical load of 1 t/m is applied over the center span. A unit horizontal load of I t
is applied at deck level. A unit shortening of the deck is applied such that EA = lo?
Foundations, Piers, and Abutments
262
The deformations (8, U) produce internal forces
(m, t, n and m’, t’, n’) in the top of the legs, which
require the following conditions:
t, = t;,
m , = mi,
72, = -n;
If R. represents the rigidity of the superstructure against rotation and R, toward longitudinal
displacements, and if M and Q represent the moment and horizontal force that the superstructure
transmits to the pier, we have:
M = - RoO f h, b, n,)
These equations may be transformed to substitute the deformations of the superstructure (0, U)
for those of the legs:
with aw = (Y sin 4 + /3 cos 4 and /3 = (IIEc)n,.
The condition of initial load of the leg (expressed by no) is modified from the case of the displacement imposed to the structure and becomes:
Normal force:
no + nl
B e n d i n g m o m e n t : m,
t,
Transverse force:
The additional forces m, and t, may be expressed
as a function of the displacement of the legs (w, (Y)
and of the initial force rzo. By substituting these
forces, as functions of (Y and o, into equations 5- 11,
we obtain a system of linear equations in three unknowns, n, a, w.
When we assume that the displacements (a, w)
are different from zero when the cause inducing
the displacement vanishes, the determinate form
N umber
1
4
5
6
7
ncr
=
T2Ei
x2
with h equal to the effective buckling length. Thus
the equivalent buckling length of one leg as part of
the total pier system will be:
*A&
r
The example of the Choisy-le-Roi Bridge will
again be considered. Seven typical cases were investigated with either vertical or inclined legs and
different leg end restraints. Also the horizontal restraint of the bridge over the abutment was varied.
Table 5.8 summarizes the results for the following
numerical values:
Wall length 1 = 8.50 m, on center spacing 2 d =
2.00 m
Area a = 6.40 m2, moment of inertia i = 0.085 m4
Neoprene pads over the abutments: area A b = 1.28
m2, E/G = 20,000
The first six cases are hypothetical assumptions
used for comparison. Case 7 is the actual case of
the Choisy-le-Roi Bridge with the legs hinged at
the base and fixed to the superstructure.
Choisy-le-Roi Bridge: Elastic Stability of Twin-Flexible-Legged Pier for
Various Support Conditions
Conditions of Legs
at River Piers
Case
2
s I
5.8.
where r is a dimensionless coefficient which may be
related to the usual Euler formula for buckling:
(5-l 1)
R,(a cos 4 - P sin 4) g’(n,, tJ
TABLE
Ei
ncr = r2l2
(5-10)
Q = - R,u gh nd
Ref(m,, t,, n,)
of the three equations is nil, which allows us to obtain the value of critical load nIc.
The critical buckling force of one pier leg may be
expressed as:
Hinged vertical legs
Vertical legs hinged at the base and fixed at the
top
Vertical legs fixed top and bottom
Legs inclined 6.5%, hinged at base, fixed at top
Legs inclined 6.5%, hinged at base, fixed at top
(actual case of Choisy-le-Roi)
Support Condition
at A butments
Rigidity neglected
Rigidity neglected
Five neo prene pads
Three neoprene pads
Rigidity neglected
Rigidity neglected
Three neoprene pads
A
c
Factor of
Safety
0
2.G
1.20
1.00
1.00
0.88
0.97
1.1
2.8
4.0
4.0
5.2
4.8
Flexible Piers and Their Stability During Construction
The designer should be aware that the following
three factors play an essential role in the elastic stability of the structure:
Inclination of the legs to the vertical
Horizontal rigidity of the neoprene bearings at the
abutments
Fixity conditions of the ends of the legs in the piers
The fundamental difference between cases 2 and 6
(Table 5.8) indicated by the considerable increase
in the factor of safety (1.1 to 5.2) is due to the introduction in case 6 of the arch effect of the inclined legs. Horizontal displacements of the
superstructure cannot occur without mobilizing
the bending stiffness of the pier assembly. For case
2 the elastic stability relies solely on the bending
stiffness of the legs, and the critical buckling force
is the same as for a beam fixed at one end and free
at the other.
5.7
Flexible Piers and Their Stability
During Construction
5.7.1
SCOPE
In the preceding paragraphs we considered piers
having a bending capacity allow ing the deck cantilever construction to proceed with no further
strengthening. Such moment-resisting piers are
usually joined to the superstructure to benefit from
the frame action, both to reduce the cost of foundations and minimize the effect of live loading in the
superstructure.
Another type of substructure remains to be considered here, one more conventional in design and
where the piers receive the vertical reaction of the
superstructure through a single row of bearings.
Such piers are usually flexible, and the stability
during cantilever construction requires that temporary supports be added to the self-bending
strength of the pier shaft.
5.7.2 DESCRIPTION OF REPRESENTATIVE
STRUCTURES W ITH TEM PORARY SUPPORTS
Dow nstream Paris Belt Bridge, France
The four river pier shafts previously described and
illustrated in Section 5.3.2 rest on a reinforced
concrete substructure built inside a cofferdam
sealed with tremie concrete. Dimensions are shown
in Figure 5.44.
263
Because of the limited dimensions of the pier
shafts and their consequent marginal bending
capacity, a temporary support was used during
construction for stability of the superstructure before deck continuity was achieved. Only one support was used for each pier, Figure 5.45, on one
side of the concrete shaft within the space available
inside the temporary cofferdam. Consequently the
lever arm between the pier and support centerlines
was only 8.5 ft (2.40 m), so that a heavy reaction
was imposed on the temporary support.
The maximum reaction computed for the case
of one precast segment out of balance, including
the lifting equipment, was 1170 tons (1060 mt). Including provisions for random loads and the
added reaction of the temporary prestressing tendons, the maximum design reaction in the support
was 2030 tons (1840 mt). Each temporary support
consisted of:
A 40 in. (1 m) steel pipe filled with concrete, Figure
5.46, resting on the spread footing of the ‘permanent pier
A V-shaped concrete frame placed upon the pipe
and allowing the deck reaction to be transferred
directly from the box section webs to the pipe
Vertical prestressing tendons were also anchored in the pier footing and stressed from deck
level to prevent accidental overturning of the cantilever, although limitations were imposed during
construction to always start segment placement on
the side of the temporary support.
Temporary connection between the pier segment and the concrete pier shaft included one
looped tendon and four high-strength bars. An
immediate consequence of the high vertical reaction imposed upon the deck by the temporary support in case of unbalanced loading was a reversal of
shear stresses between the temporary and the
permanent supports. This situation was even more
critical because of the permanent draped tendons,
shown in the detail of Figure 5.47, located in that
zone together with the Resal effect produced by
the inclined bottom flange. The corresponding
shear stress in the webs reached a maximum of 680
psi. Two special tendons (twelve 3 in. diameter
strands) were placed on either side of each web of
the box girder to reduce the shear stresses to allowable values. In fact, these four tendons worked
as a tension tie between the top and bottom flanges
of the box girder across the distance between the
permanent and temporary support.
SECTION
TRANSVERSAL
l l ln
KOLW
----PLE UP -
cl+44
/'
d-P-Y E
-Tif i
f
PLAN VIEW
FIGURE
HOFUONTAL SECTION
5.44.
Downstream Paris Belt Bridge, dimensions of river piers.
-SEGMENT
-MAX.
WEIGHTS: 60 to
STATICAL
VERTICAL
SUPPORT
IN
40
t
SUPPORT
U360 t
42.40
In-.-____c
.-
PROVISIONAL
__~ -.~
REACTION
PRESTRESSING
.-
\PRESTREZfSING
RODS
FIGURE 5.45. Downstream Paris Belt Bridge, schematic of temporary support and
stability of river pier during construction.
264
:
265
Flexible Piers and Their Stability During Construction
STEEL
P/T
CAP
N
M
ANCHORS
FLANG
JOINT
Saint Jean Bridge In Bordeaux, France
FIGURE 5.46. Downstl c;m Paris Belt Bridge, details
of temporary support. (a) Dimensions of support. (b)
View of support.
This problem has been described at some length
to show that a single temporary support subjected
to high loads may call for a rather complex arrangement to satisfy all requirements of stability
and resistance of all parts of the structure at each
construction stage.
For aesthetic reasons the river piers were designed
as rather slender shafts, which had to accommodate an important variation of the waterline due to
tidal effects in the mouth of the Garonne River.
The bridge was relatively low above the water,
particularly at high tide.
E a c h p i e r s h a f t w a s f o u n d e d o n a n opendredged concrete caisson anchored in a bed of
sand and gravel of good quality, overlying a deep
formation of marl and clay.
Dimensions of the piers and foundations are
shown in Figure 5.48. The caisson had a cuttingedge diameter of 18 ft 4 in. (5.60 m) and the
maximum average bearing pressure on the sand
and gravel bed was 8.1 t/ft2 at the time of first
loading; the foundation settlement was a maximum of 1.1 in. (28 mm) and the long-term additional settlement was negligible, 0.16 in. (4 mm).
Construction of the piers called for the use of an
auxiliary floating platform that could be raised on
eight temporary pipe piles, comparable in principle to the legged jacking platforms used on
offshore work, Figures 5.49 and 5.50. The reinforced concrete caisson was floated into place, suspended from the platform resting on its legs, and
incorporated into the permanent structure. As excavation proceeded inside the caisson to lower it to
its final elevation, precast segments were added to
increase the height of the caisson wall as required.
zyxwvutsrqp
SHEAR AT SECTION OF
TEMPORARY SUPPORT
v, 1ooot
t
TRANSFORMED
/
TEMPORARY
CONCRE
TRANSFORMED
SECTION
:
BEARING PADS
zyxwvutsrqponmlkjihgfed
L%m
FIGURE 5.47. Downstream Paris Belt Brid ge,
detail of loads on cantilever and temporary
suppo rt.
FIGURE 5.49. copposite). St. Jean Bridge
in Bordeaux, schematic
of construction of river
p iers.
,
266
RA IN
C O NC R~
TREMIE
CONCRETE
FIGURE 5.48.
St.
Jean Brid g e, in Bo rdeaux, dimensions of
river piers.
TING
-
C
- L A -M
FLEXI
c
R / C CAl%ON
BORDEAUX
-
FLOAT5
-
PLACING
RIVER
CAlSON
_
ELEVATION
low- 72.
FLOATING CLAM
5HELL C R A N E
V_ERTICAL
F’PE
PIL
267
Foundations, Piers, and Abutments
268
P
5.7.3 REVIEW OF THE VARIOUS METHODS OF
PROVIDING STABILITY DURING CANTILEVER
CONSTRUCTION
A situation is considered here where the permanent pier cannot provide adequate stability during
cantilever construction. Several methods may be
used, either separately or in combination, to provide the required stability under the loading combinations briefly reviewed in Section 5.2.
In the general case
Temporary Eccentric Prestress
where the construction procedure allows the unbalanced segment in a typical cantilever to be
placed always on the same side of the pier, the unbalanced moment varies between 0 and Wd (segment weight W at a distance d from the pier centerline as shown in Figure 5.52).
Assume a temporary vertical tendon, anchored
in the pier foundation or in a separate dead-man,
to be stressed for this unbalanced loading configuration to a load P such that
FIGURE 5.50. St. Jean Bridge at Bo rdeaux, platfo rm
on legs used for river pier caissons. (a) Platform in floating stage. (6) Platform on legs and caisson during excavation.
and the unbalanced moment in the pier now becomes
+w d
- 2
Match casting was used for making the various
segments, and it proved very efficient and very
simp le.
The cofferdam required to build the pier shaft
in the dry was made up of temporary additional
caisson ring segments stacked upon the permanent
caisson and bolted together. This cofferdam was
used during construction of the deck to make a
moment-resisting pier shaft as a substitute to the
flexible permanent pier. The deck was therefore
resting only upon the cofferdam and the lower
caisson through two temporary caps, offering a
stable base for unbalanced loading, Figure 5.51~.
After cantilever construction was finished and
continuity achieved in the deck, flat jacks were
used to transfer the total reaction of the box girder
from the temporary caps and cofferdam onto the
permanent concrete piers. All the temporary ring
segments above low water were further removed.
This example shows how the foundations and even
part of the substructure can be used to minimize
the cost of temporary supports required for cantilever construction.
and the actual bending capacity of the pier is
theoretically doubled. The true gain is somewhat
lower, because it is not practical to change the tendon load at each stage of segment placing. A
proper temporary connection with high-strength
rods between pier and deck must always be provided.
Unsymmetrical Distribution of Segments with Regard
to the Pier If the pier segment is eccentrically
placed with regard to the pier shaft centerline,
Figure 5.53, a permanent moment is applied to the
pier when an even number of segments is incorporated in the deck. Dimensions may be such that the
maximum unbalanced moment due to one segment’s being placed on the proper side of the pier
will result in applying only half to the pier. This
approach results in significant complications in the
layout of the prestress tendons in the deck. Both
methods described thus far have one disadvantage,
in that the deck cantilever is never in balance over
the pier and so it is more complicated to following
up the geometry of the deck during construction.
FIGURE 5.51. St. Jean Bridge at Bordeaux, temporary arrangement of piers for deck cantilever construction. (a) Schematic of temporary cofferdam and deck
support. (b) View of the pier segment and travelers.
269
270
Foundations, Piers, and Abutments
zy
///I/Py”Hf
d
FIGURE 5.52. Temporary stability of deck and pier during construction by prestressing tendon.
Overhead truss in cast-in-place construction, Siegtal Bridge or Pine Valley Creek Bridge
Launching gantry in precast construction, Rio
Niteroi Bridge and the B-3 South Viaducts
Overhead beam in precast construction, B-3 South
Viaducts; a similar scheme is being contemplated
for several contemporary projects in the United
States.
FIGURE 5.53. Unsymmetrical pier segment.
Stability of the Concrete Cantilever Provided by the
Deck Construction Equipment Figure 5.54 outlines a
few typical schemes developed for either cast-inplace or precast construction where the stability
during cantilever placing is achieved by the construction equipment itself, such as an overhead
truss or launching gantry. Several such examples
were previously described in Chapters 2 and 3:
Temporary
Su@orts (Fig ure 5.55) If a sing le
temporary support is used on one side of the pier
at a distance a, the reactions are as follows:
pM
a
temporary support:
c
M=W.d
IV
Q
zyxwvutsrqponmlkjihgfedcbaZY
Ties for stability
Overhead truss
+M
a
---_
__- -
----
M=Wd
QV
Winch
Overhead beam
FIGURE 5.54. Cantilever stability by deck construction equipment.
FIGURE 5.55. Cantilever stability by temporary support(s).
271
Abutments
If two symmetrical temporary supports are used,
the system is statically indeterminate and the actual
distribution of reactions depends upon the respective flexibilities of the pier and of the supports.
The load distribution is as follow s:
Tempo rary
Support, T,
Pier, P
Tempo rary
Support, T,
PV
(1 - 2P)V
PV
Effect of vertical load Y
Pv-g
If it is desired that the temporary supports never
be subject to an uplift force, to resist which requires anchors and adequate foundations, the
stiffness of the support must be such that a
sufficient proportion of the vertical load compensates the effect of the moment. The minimum
value of p must be such that:
M
2a
0
2a
Total
p V--20 o
M
M
Effect of moment M
r
M
2a
(1 -
+2a
2P)VzyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFED
PV + $
the same loading configuration. The double support system is therefore exactly twice as expensive
as the single support system. The only advantage is
to allow the construction of the deck to proceed
indifferently from either side of the pier or to
maintain an equal safety of the system should a
mistake be made in the required sequence of operations for the case of a single support.
p2---
Consequently the maximum reaction at support T2
becomes at least equal to M /a, w hich is precisely
the value of the reaction for a single support with
Tempo rary Stay s In a limited number of structures, stability during construction was provided by
temporary vertical or inclined stays anchored in
special foundation blocks or in the permanent
footing of the pier, Figure 5.56.
When feasible, this last system is particularly
simple, because the temporary stays are usually
made of simple prestressing tendons and are far
less expensive than rigid temporary supports. Such
a system must be used in conjunction with a strong
temporary connection between pier and deck to
reach an adequate level of safety.
5.8
Abutments
5.8.1
SCOPE
Although the abutments provided at both ends of
the bridge are not necessarily of special design
when associated with cantilever and segmental
construction, it may be of interest to review briefly
several types of structures actually used in completed projects.
The abutments serve a twofold purpose:
FIGURE 5.56.
Cantilever stability by temporary stays.
They provide the first and last support to the
bridge superstructure, allowing a smooth transition of the roadway surface from the deck to the
272
Foundations, Piers, and Abutments
approaches while allowing free expansion with an
adequate roadway and sidewalk joint,
They make the retaining wall contain the fill of the
approach embankment where geometric conditions require it. Design and construction methods
of the abutments depend greatly upon the soil
conditions and the level of the water table when
present.
Basically, the two functions outlined above ma!
either be integrated into a single structure or filled
by two separate structures. On the other hand, the
function of a retaining wall may be greatly minimized by allowing the approach fill to take a slope
of repose under the structure.
By variously combining these characteristics,
twelve different sketches were prepared in Figures
5.57 through 5.68 as an outline of typical structures encountered in practice. For convenience,
these designs have been grouped into six different
categories as described in the following paragraphs.
5.8.2 COMBINED ABUTlMESTIRETAI.VI.1’G
WALL
Type IA (Figure 5.57) A simple retaining wall
perpendicular to the bridge centerline and anchored to a conventional spread footing both contains the approach fill and provides the deck end
bearing. The back wall receives a transition slab to
avoid the roadway profile discontinuity so frequent
in earlier bridges between the rigid deck and the
flexible pavement over the approach embankment.
Two side walls of triangular shape contain the fill
inside the abutment.
Type ZB (Figure 5.58) The retaining wall is
made of a vertical wall and a lower slab properly
strengthened by longitudinal buttresses. The entire system is founded on piles.
Type IC (Figure 5.59) Where the poor quality of
the soil makes it difficult to resist the horizontal
loads due to earth pressure combined with braking
and thermal reactions, the previous system may be
founded on a system of vertical piles, while the
FIGURE 5.57. Abutment type
IA.
1
If/
1
FIGURE 5.58. Abutment rype
IB.
t-‘--t
FIGURE 5.59. Abutment type IC.
273
274
Foundations, Piers, and Abutments
horizontal loads are resisted by embedded prestressed concrete ties anchored in the back into a
continuous dead-man.
5.83
SEPARATE END SUPPORT AND
RETAINI,VG WALL
Type ZZ (Figure 5.60) The two functions of deck
support and retaining wall are entrusted to two
separate structures. Shown in this figure is a front
vertical column, resting on spread footings or piles,
which provides the deck end bearing. Behind this
column and separate thereto, a reinforced earth
retaining w all contains the approach fill.
5.8.4
THROUGH
FILL
ABUTMENT
The fill extends under the bridge deck with a stable
slope (3 : 2 to 2 : 1) to reduce as much as possible the
amount of earth pressure applied to the abutment.
Type ZZZA (Figure 5.61) Vertical longitudinal
walls connect the lower spread footing to the
abutment superstructure. It is important to avoid
horizontal cross bracings at intermediate levels
embedded in the fill, because settlements may
cause significant overloads in such members such
as to cause failure.
-
t-4.G
-
-
& ~~~FOU R-4 ’@m
PI LES-_
_
(b)
FIGURE 5.60. Abutment type II with reinforced earth. (a) Cross section. (b) Elevation
and longitudinal section.
Abutments
275
FIGURE 5.61. Abutment type IIIA.
Type ZZZB (Figure 5.62) The same system may
be adapted to the case where a high water table and
poor soil conditions call for pile foundation built in
a cofferdam.
5.8.5 HOLLOW BOX ABUTMENT
Trpe WA (Figure 5.63) Another way to avoid
high earth-pressure loads on the abutment, where
it is not possible or desired to extend the approach
fill under the deck, is to build the abutment as a
box with a front wall providing the deck end support and the cover slab carrying the roadway between the bridge deck and the approach fill. Such a
structure may be founded on spread footing or on
piles (as shown in the sketch).
Type ZVB (Figure 5.64) The same structure may
rest both at the front and at the rear on opendredged caissons excavated under water to the
lo ad -bearing so il.
5.8.6 ABUTMENTS DESIGNED FOR UPLIFT
The principle has been described previously in
Chapter 4 (design) and for actual structures in
Chapters 2 and 3 (cast-in-place or precast cantilever bridges).
Type VA (Figure 5.65) A large caisson is opendredged and filled after completion of the excavation to the required foundation level with tremie
concrete so as to obtain a sufficient weight to resist
the uplift reaction from the deck.
Type VB (Figure 5.66) Another variation of the
same concept was developed for the Saint Jean
Bridge at Bordeaux to combine into a single abutment a front downward bearing and a rear uplifting
bearing to fix the last span of the bridge while retaining its free expansion.
276
Foundations, Piers, and Abutments
FIGURE 5.62. Abutment type IIIB.
5.8.7
MINI-ABUTMENT
For decks of small height, when prevailing conditions allow the fill to be placed around the deck, the
abutment reduces to a very simple inexpensive
structure shown as types VIA and VIB in Figures
5.67 and 5.68.
5.9
Effects of Differential Settlements on
Continuous Decks
The question has often been raised as to the adequacy of allowing continuous decks to rest on piers
subjected to possible differential settlements. The
authors are aware of a few cases where differential
277
Effects o f Differential Settlements o n Co ntinuo us Decks
d-t
6”
I
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
i'
/
FIGURE 5.63. Abutment type IVA.
settlements were responsible for problems pertaining to the integrity of the superstructure (such
as opening of joints between successive segments).
Differential settlements, however, are very seldom
critical in most soil conditions. In the isolated cases
where they may be critical, precautions can be
taken to counteract their eventual effects upon the
structure.
5.9.1 EFFECTS OF AN ASSUM ED PIER
SETTLEMENT ON THE STRESSES IN THE
SUPERSTRUCTURE
Starting with the simple case shown in Figure 5.69,
where a continuous beam of constant depth with a
large number of identical spans is subjected to the
settlement of one pier by a given amount, one may
easily derive the effect in terms of moments and
stresses in the superstructure. Taking the fixed
end moment p = 6 EZu/12, the moments over the
piers and at midspan are:
Over the pier subjected
to settlement
Over the adjacent piers
+0.732p
Midspan
+0.134/ L
moment
Quarter-span moment
-0.464~
+0.433p
The stress produced in the superstructure is f =
MC/I, where c is the distance between the centroid
278
Foundations, Piers, and Abutments
FIGURE 5.64. Abutment type IVB.
and upper or lower flange. If the moment is expressed as A4 = Ap, the stress becomes:
6Ecuzyxwvutsrqponmlkjihgfed
f=+r
which can be rewritten as follows:
The value of clh varies between 0.4 and 0.6 and
that of hll between & aud &.
Considering the quarter-span .point close to the
pier where settlement occurred, the stress in the
superstructure will be, with k = 0.433 and E =
300,000 kips/ fP (for long-term loading):
f =
FIGURE 5.65. Abutment type VA.
23,400;
For a settlement u = r$6a the stress is equal to 23
kips/ ft2 at the bottom fiber, a very nominal value.
For a 100 ft span, the corresponding settlement is
u = 0.1 ft = 1.2 inches.
The amount of settlement to be considered is
only that part taking place after continuity is
achieved in the deck and so after most of the load
has been applied to the structure.
Effects of Differential Settlements on Continuous Decks
awble
279
1
FIGURE 5.66. Abutment type VB.
5.9.2 PRACTICAL M EASURES FOR
COUNTERACTING DIFFERENTIAL SETTLEMENTS
In most cases, the foreseeable differential settlements may be absorbed by the structure without
any corrective measures and no special provisions
need be taken in that respect.
For some structures the situation may call for
special consideration. Such was the case, for example, with the Houston Ship Channel Bridge, where
large long-term settlements could be anticipated at
the time of design. In such instances, provisions for
eventual realignment of the deck profile must be
incorporated into the design.
Foundations, Piers, and Abutments
280
FIGURE 5.67. Abutment type
VIA.
FIGURE 5.68. Abutment tvpe VIB
Inertia : I
Modulus : z
FIGURE 5.69. Effect of differential settlement on a continuous beam with equal spans
and constant depth.
References
1. J. Mathivat, “Reconstruction du pont de Choisy-leRoi,” Trauaux, Janvier 1966, No. 372.
2. J. Mathivat, “Structures de piles adaptees a la construction par encorbellement,” Problems speciaux
d’etude et d’execution des overages, Journees
A.F.P.C., Avril 22-23, 1974.
3. Gerwick, Ben C. Jr., “Bell-Pier Construction, Recent
Developments and Trends,” Journal of the American
Concrete Institute, Proc. V. 62, No. 10, October 1965.
zyx
6
Progressive and Span-by-Span
Construction of Segmental Bridges
J
6.1
zyxwvu
6.4.5 Guadiana Viaduct, Portugal
6.4.6 Loisach Bridge, Germany
6.4.7 Rheinbriicke Dusseldorf-Flehe,
I N T R O DUC T I O N
6.1.1 Progressive Placement Method
6.2
6.1.2 Span-by-Span Method
PROGRESSIVE CAST-IN-PLACE
BRIDGES
6.2.1
Approach Spans to the Bendorf Bridge, Germany
6.2.2 Ounasjoki Bridge, Finland
6.3
6.2.3 Vail Pass Bridges, U.S.A.
PROGRESSIVE PRECAST BRIDGES
6.3.1 Rombas Viaduct, France
6.3.2 Linn Cove Viaduct, U.S.A.
6.4 SPAN-BY-SPAN CAST-IN-PLACE
BRIDGES
6.4.1 Kettiger Hang, Germany
6.4.2 Krahnenberg Bridge, Germany
6.4.3 Pleichach Viaduct, Germany
6.4.4 Elztalbticke,
Germany
6.1
Introduction
The concepts of the progressive placement and
span-bv-span methods of segmental construction
were introduced in Sections 1.9.4 and 1.9.3, respectivelv.
.fhis chapter will explore these concepts
in greater detail. These two methods have not
made the conventional cast-in-place on falsework
method obsolete; the conventional method is still
applicable and economical where site, environmental, ecological, and economic considerations
permit. What these two methods do is to open up a
held where prestressed concrete structures were
hitherto not practical and where they now can economically compete with structural steel.
.The progressive placement and span-by-span
methods are similar in that construction of the
superstructure starts at one end and proceeds continuously to the other, as opposed to the balanced
cantilever method where superstructure is constructed as counterbalancing half-span cantilevers
6.5
Germany
6.4.8 Denny Creek Bridge, U.S.A.
SPAN-BY-SPAN PRECAST BRIDGES
6.5.1 Long Key Bridge, U.S.A.
6.5.2 Seven Mile Bridge, U.S.A.
6.6
DESIGN ASPECTS
C O N S T R UC T I O N
6.6.1
6.6.2
6.6.3
OF
SEGMENTAL
PROGRESSIVE
General
Reactions on Piers During Construction
Tensions in Stays and Deflection Control During
Construction
6.6.4
Iayout of Tendons for Progressive Construction
REFERENCES
on each side of the various piers. Also, both
methods are adaptable to either cast-in-place or
precast construction.
6.1.1
PROGRESSIVE
PLACEMENT
METHOD
This method was developed to obviate the construction interruption manifested in the balanced
cantilever method, where construction must proceed symmetrically on each side of the various
piers. In progressive placement, the construction
proceeds from one end of the project in continuous increments to the other end; segments are
placed in successive cantilevers from the same side
of the various piers. When the superstructure
reaches a pier, permanent bearings are placed and
the superstructure is continued in the direction of
construction.
The first implementation of this method, which
used cast-in-place segments, was on the Ounasjoki
Bridge near the Arctic Circle in Finland. It was
281
282
Progressive and Span-by-Span Construction of Segmental Bridges
later extended to the first use of precast segments
in the Rombas Viaduct in eastern France.
The essential advantages of this method are as
follows:
1.
The operations are continuous and are carried
o u t f ro m that p art o f the stru c tu re alread,
constructed. Access for personnel and materials is conveniently accomplished over the surface of the structure already completed (free
of the existing terrain). This may be of importance with regard to urban viaducts cantilevering over numerous obstacles.
2.
Reactions to the piers are vertical and not subject to any unsymmetrical bending moments,
thus avoiding the need for temporary bracing
during
construction.
3. The method is adaptable to curved structure
geometry.
The following are the disadvantages:
1.
It is difficult, if not impossible, to utilize this
method in the construction of the first span.
Usu ally the first sp an m u st b e erec ted o n
falsework. In some rare instances it may be
possible to cantilever the first span from the
abutment.
2. Forces imposed upon the superstructure, depending on the method of construction, are
c o m p letely d if f erent ( in sig n and o rd er o f
magnitude) from those present in the structure under service load. Consequently, a temporary external support system is required
during construction in order to maintain the
stresses within reasonable limits and minimize
the c o st o f u np ro d u c tiv e tem p o rary prestressing. Falsework bents may be used (as in
the Linn Cove Viaduct), but the more usual
solution is that of a mobile temporary mast
and cable-stay system (Figure 1.57). For the
progressive placement method the mast and
cable-stay system is relocated progressively
over the piers as construction advances.
3.
In this system the piers are subjected to a reaction from the self-weight of the superstructure
approximately twice that in the final static arrangement of the structure. However, this is
generally not critical to the design of the piers
and foundations, as the effect of the dead load
is rarely larger than half the total load including horizontal forces.
When cast-in-place segments are used in conjunction with the progressive placement method,
the rate of construction is less than that t’or the balanced cantilever method, in that there is onlv one
location of construction activitv. That is, onlv one
segment can be cast (at the end of the completed
portion of the structure) rather than two (one at
each end of the balanced cantilevers). ~fhis slowness may be minimized by the use of longer segments, but this solution is limited bv the low resistance of the young concrete. On the other hand,
the u se o f ep o xy - jo ined p rec ast seg m ents ma!
permit an average rapidity of construction comparable to that of balanced cantilever with a launching girder.
A s ind ic ated in Chap ter 1, the sp an-b y -sp an
method was developed to meet the need for constructing long viaducts with relatively short spans
such as to incorporate the advantages of balanced
cantilever
construction.
From a competitive point of view, the capital investment in the equipment for this type of construction is considerable. It has been suggested’
that one-third of the cost of the equipment be depreciated for a given site and that at least four uses
would be required to achieve full depreciation, including interest on the capital investment. However, costly modifications that may be required
because of changes in bridge widths or span limitations are not considered in the above write-off
policy. It would, therefore, be advisable for a contractor investing in this tvpe of equipment to consid er so m e ty p e o f m o d u lar p lanning so that
modification for future projects might be kept to a
minimum. It might be possible to have a basic piece
o f eq u ip m ent w ith interc hang eab le elem ents.
There is, of course, the potential of leasing this
equipment to others as a means of retiring the
capital investment.
Wittfoht1s2 has categorized stepping segmental
construction intb four subgroups:
1.
With-on-the
ground
nontraveling
formwork.
2. W ith trav eling formw ork o r o n- the- g ro u nd
stepping formwork.
3.
With
off-the-ground
stepping
formwork.
4.
In opposite directions starting from a pier.
The first category is generally used where there
are a large number of approximately equal spans
Progressive Cast-in-Place Bridges
283
of a low height above existing terrain. It is generi-2zyxwvutsrqponmlkjihgfedcbaZYXWVUTSR
rf
ally limited to structure lengths of approximately
1000 ft (300 m) and to nonuniform span lengths
that prohibit a forming system of uniform size.
N o rm ally in sp an- b y - sp an c o nstru c tio n the
superstructure is of constant cross section (at least
insofar as external dimensions are concerned), and
Scaffolding at concreting position
the w o rk p ro c eed s f ro m o ne ab u tm ent to the
other. If a large center span exists, it will be formed
Construction direction
first, possibly to an inflection point in the adjacent
spans. The formw ork is allocated such that it is
used to cast the spans in the approaches proceeding from the center, in both directions, toward the
‘Under-carriage
abutments. Forms and scaffolding are disassemAdvancement of Scaffolding
bled and reerected in an alternating sequence and
in elements that can be conveniently handled by a
crane.
In the second category of span-by-span conffold
struction, for economical justification of equipm
ment, the total length of structure must be at least
re
1000 ft (300 m), the overall cross section constant,
_.,.._
Hinged bottom plate
the stru c tu re o f lo w heig ht, and the terrain
Section 2-2
Section l-l
along the longitudinal axis approximately level.
Maximum span for this category is approximately
FIGURE 6.1. Schematic of procedure for movable
165 ft (50 m), and a large number of equal spans
scaffolding, from reference 3 (courtesy of Zement und
are required to achieve repetitiveness and thus
Beton).
econon1v.3
The falsework and forms are generally a span
struction indicated by the fourth category may be
length (either the dimension from pier to pier or
considered. This system uses a gantry rig that has a
from inflection point to inflection point), Figure
length one and one-half times that of the span. In
6.1 .3 The formw ork is fixed to the scaffolding and
this method segments are cast in each direction
travels with it. The bottom of- the formw ork is defrom a pier, as in the balanced cantilever method,
signed with a hinge or continuous trap-door device
except that the form traveler and segment being
such that the scaffolding and forms can travel past
cast are supported by the gantry. This method is
and clear the piers. The scaffolding is moved foractually a balanced cantilever method and not a
ward on rails. If a foundation for the scaffolding,
span-by-span method of construction as defined
forms, and weight of superstructure is found to be
here.
too costlv or unsafe, a scheme may be used where
The advantages of the span-by-span method of
the rails ‘carry only the load of the scaffolding and
construction, besides those associated with segfo rm w o rk. O nc e in p o sitio n, the sc affo ld ing is
mental construction in general, pertain to the presupported at the piers, or at the forward pier, and
stressing steel requirements. Since the segments
the completed structure at the rear by auxiliary
are supported by the form travelers, there are no
brackets; thus construction loads are transmitted to
cantilever stresses during construction, and prethe pier foundations.
stress requirements are akin to those of convenWhere conditions exist as in the previous catetional construction on falsework or those for the
gory, but the structure is high with reference to the
final condition of the structure.
terrain or crosses over difficult terrain or water,
the third category may be used, whereby during
6.2 Progressive Cast-in-Place Bridges
the stepping and casting operations the equipment
is supported by the piers or by a pier and the pre6.2.X APPROACH SPANS TO THE BENDORF
viously completed portion of the structure.
BRIDGE, GERMANY
Where consecutive spans in the range of 160 to
As discussed in Section 2.2, the Bendorf Bridge
500 ft (50 to 150 m) are contemplated and the facwas constructed in two parts. The western portion
tors mentioned above prevail, the type of con-zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
Progressive and Span-by-Span Construction of Segmental Bridges
284
Main river
Flood
Construction in
free cantilever
216.50 m
Construction on
I-.___--.-__i
falsework
288.50
m
‘Phase 5 by progressive placing, segment length 4.00 m.
FIGURE 6.2. Bendorf Bridge, Part Two (East), construction procedure, from reference 1 (courtesy of Beton- and Stahlbetonbau). Phase 5 by progressive placing, segment
length 4.00 m.
(part one), Figure 2.9, consists of a symmetrical
seven-span continuous girder constructed by the
cast-in-place balanced cantilever method. The
eastern portion (part two), Figure 2.10, consists of
a nine-span continuous approach structure having
an overall length of 1657 ft (505 m) with spans
ranging from 134.5 ft (41 m) to 308 ft (94 m).
In the construction of the approach spans, Figure 6.2, the five spans from the east abutment were
built in a routine manner with the assistance of
falsework bents. The four spans over water were
constructed by the progressive placement method,
using cast-in-place segments and a temporary
cable-stay arrangement to reduce the cantilever
stresses. The temporary stay system consisted of a
structural steel pylon approximately 65 ft (20 m)
high and stays composed of Dywidag bars.
6.2.2 OUNASJOKI BRIDGE, FINLAND
This structure is near the city of Rovaniemi, Finland, and crosses the Ounas River just above its
junction with the River Kemi near the Arctic Circle. The structural arrangement consists of two 230
ft (70 m) interior spans and end spans of 164 ft (50
m), prestressed longitudinally and transversely.
The first end span and 75 ft (22.75 m) of the
second span were cast-in-place in a conventional
manner on falsework inside a temporary windshielded protective cover, Figure 6.3. Outside
temperature during this operation ranged from
- 2 0 t o -30°C. Subsequent progressive cantilever
construction was performed-with the aid of a temporary pylon and stays, Figure 6.4. The same
stages were repeated in the remaining spans. The
superstructure was cast-in-place with the assistance of one form traveler, Figure 6.5. During
these stages of construction, for protection against
low temperatures, form traveler and form were
fully enclosed, Figure 6.5. This enclosure was insulated with 4 in. (100 mm) of fiberglass.
Hardening of the concrete took an average of 76
hours. Temperature of the concrete was maintained between 35 and 45°C at mixing and between
20 and 25°C during casting. Curing inside the
form traveler enclosure was assisted by warm-air
blowers. Concrete strength was 5000 psi (34.5
MPa). Segment length was 11.5 ft (3.5 m), and it
was possible to reach a casting rate of two segments
a week.
Construction started in 1966 and was completed
in 1967. Table 6.1 lists the temperatures recorded
during seven months of the construction period.
The progressive placement method proved effective and work progressed throughout the year
even during arctic conditions.
FIGURE 6.3. Ounasjoki Bridge, temporary protective
structure (courtesy of Dyckerhoff & Widmann).
285
Progressive Cast-in-Place Bridges
TABLE 6.1. Ounasjoki Bridge, Temperature Variations
Month
Temperature
March
April
May
June
July
August
September
Average “C
Maximum “ C
Minimum “C:
-2.5
+5.8
-26.4
-0.4
+9.9
- 16.8
+5.6
+24.6
- 12.2
+11.7
+24.9
+0.1
+ 14.3
+25.7
+3.0
+ 14.8
+28.5
+5.8
+8.7
+ 19.3
-4.7
62.3 VAIL PASS BRIDGES, U.S.A.
FIGURE 6.4. Oulla+ki Bridge, winterproof
ing form (courtesy of Dyckerhof’f & Widmann).
FIGURE 6.5.
travel-
The Vail Pass structures are part of Interstate I-70
near Vail, Colorado, in an environmentally sensitive area. Of the 21 bridge structures in this project, seventeen were designed and bid on the basis
of alternate designs (Chapter 12). In the segmental
alternative the contractor was allowed the option
to construct as cast-in-place segmental. A group of
four bridges approximately 7 miles (11.3 km)
southeast of Vail were successfully bid as cast-inplace segmental and used the concept of progressive placement.
Two of these structures are contained in a fourspan dual structure over Black Gore Creek, Figure
6.6. The other two structures are a three-span
Ounasjoki Bridge, progressive placing scheme.
286
Progressive and Span-by-Span Construction of Segmental Bridges
Existing grcamdlim
TYPICAL ELEVATION
MID-SPAN
NEAR Q PIER
TYPICAL
SECTION
FIGURE 6.6. Vail Pass B I-‘cl
I ges, Black Gore Creek Bridge, typical elevation and section
eastb o u nd b rid g e and a f o u r- sp an w estb o u nd
bridge, both crossing Miller Creek, Figure 6.7.
Because the structures are relatively short and
the spans small, they were constructed by the
progressive placement method with temporary
falsework bents. ‘The work and time required to
transport and reassemble the form travelers (as in
the b alanc ed c antilev er m etho d ) w as thereb v
minimized. Construction started from both abu;ments and proceeded progressively toward the
center of each bridge.”
Fo r eac h o f the tw o stru c tu res in the M iller
Creek Bridge, form travelers were assembled atop
30 ft (9.1 m) long segments at the abutments. As
segment casting began, the side spans were supported at every second segment by a temporary
bent. After reaching the first pier, segment con-
struction proceeded in normal fashion to midspan
o f the eastb o u nd stru c tu re. In the w estb o u nd
structure, when midspan of both interior spans was
reached, temporarv bents were again used to conlplete the remaining half-spans to the center pier.
After reaching the center of the bridge, one form
traveler of each bridge was dismantled, and the
remaining form traveler was used to cast the closure pour. In this manner the form travelers for
each bridge were assembled and dismantled only
once, as opposed to the method of assembling t\vo
forms at each pier and dismantling upon completion of two half-span cantilevers about each pier.
For the Black Gore creek structures, to save critic al c o nstru c tio n tim e, b o th end sp ans o f o ne
structure and one end span of the other structure
were built on falsework, while the form travelers
Progressive Cast-in-Place
Bridges
455’- 3 ”
E. 8 ELEVATION
518’-3”
Bridge abut. 2
E Dridge
abut 1
W. 8. ELEVATION
42'-0"
e Elri ge
,_
:
r2"
1zyxwvutsrqponmlkjihgfedcbaZYX
Aspha lt
lo’-o-
9’-
33-e.
I'-8f'
TYPICAL SECTION
FIGURE
6.7.
Vail Pass Bridges, Miller Creek Bridge, typical elevation and section.
were occupied at the Miller Creek Bridges. Upon
completion of their work at Miller Creek, the form
travelers were transported over the completed end
spans of the Black Gore Creek Bridges and con-
struction continued in the progressive placement
manner, Figure 6.8.
Because of the limited construction time a
three-day cycle was required for segment casting.
288
Progressive and Span-by-Span Construction of Segmental Bridges
Construction specifications required a concrete
strength of 3500 psi (24 MPa) at the time of posttensioning and 5500 psi (38 MPa) at 28 days. Since
the time required for f-orming and placing of rebar
and tendons is somewhat fixed, the only operation
that could be adjusted was the concrete curing
time. This was accomplished by using a special
water-reducing agent that allowed the development of 3500 psi (24 MPa) concrete in 18 hours.
Because of lack of experience with the specific
water reducer, honeycombing was experienced in
the early stages of construction. Eventually a 24 da)
cvcle was achieved.
FIGURE 6.8. Vail Pass Bridges, Black Gore Creek
Bridge, under construction (courtesy of Dr. Man-Chung
Tang, DRC Consultants, Inc.).
ROM BAS
- P L A N VlEL/(a)
FIGURE 6.9.
Rombas Viaduct, plan and sections. (a) Plan. (6) Typical bridge sections.
(c) Typical segment section.
289
Progressive Precast Bridges
coupe c
Coupe A
Coupe D
Coupe El
/ Variable
I
Fig ure 6 . 9 .
Var 680 760
I-
-l
(C,‘o~rtitr~rd)
6.3 Progressive Precast Bridges
The Rombas Viaduct is a constant-depth superstructure, supported on neoprene bearings, with
nine co ntinuo us sp ans rang ing f ro m 75 f t ( 23
m) to 14X ft (45 m). This structure is curved in plan
with a minimum radius of 900 ft (275 m) and of a
variable width, owing to the presence of an exit
ramp, Figure 6.9. Total length is 1073 ft (327 m),
and the viaduct has two parallel single-cell boxes.
In cross section each single-cell box is 8.2 ft (2.5 m)
deep and has a width of 36 ft (11 .O m). A construction view of the end of a segment is presented in
Figure 6.10.
C o nstru c tio n o f this stru c tu re em p lo y ed the
progressive placing of the precast segments. Temporary stability was provided by a cable-stay system, Figures 1.57 and 6.11, which advanced from
pier to pier as the construction progressed. Segments were progressively placed, starting from one
Progressive and Span-by-Span Construction of Segmental Bridges
290
FIGURE
6.10.
Rombas Viaduct, end view of segment.
FIGURE
6.12.
Rombas Viaduct, view ot swivel crane.
FIGURE 6.13. Linn Cove Viaduct, photomontage.
FIGURE 6.11. Rombas Viaduct, view of cable stays
and mast.
abutment, by means of a swiveling hoist, Figure
6.12, advancing along the deck.
6.32
LJNN COVE VIADUCT, U.S.A.
A progressive placement scheme is being used for
the Linn Cove Viaduct on the Blue Ridge Parkway
in North Carolina, Figures 6.13 and 6.14. It is a
FIGURE
6.14.
Linn Cove l’iaduct,
artist’s
rendering.
291
Progressive Precast Bridges
1243 ft (378.84 m) eight-span continuous structure
with spans of 98.5, 163, 4 at 180, 163, and 98.5 ft
(30.02, 49.68, four at 54.86, 49.68, and 30.02 m)
and sharp-radius curves, Figure 6.15. In cross section it is a single-cell box girder with the dimensions indicated in Figure 6.16.
Because of the environmental sensitivity of the
area, access to some of the piers is not available.
Therefore, the piers will be constructed from the
tip of a cantilever span, with men and equipment
being lowered down to construct the foundation
and piers. The piers are precast segments sta c ke d
vertically and post-tensioned to the foundation,
Figure 6.17.
The extreme curvature of the alignment makes
the use of temporary cable stays impractical. Temporary bents at midspan will be used to reduce
cantilever and torsional stresses during construction to acceptable levels. The temporary bents are
erected in the same manner as the permanent
+ Pier 3
i
$ Pier 4
izyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONM
FIGURE 6.15. Linn Cove Viaduct, plan.
H A L F S E C T I O~~~~
N AT POST- TENSIONING BLOCK
_---
FIGURE 6.16.
TYPICAL HALF SECTION
THRU
I,inn Cove Viaduct, typical segment cross section.
SEGMENT
Span-by-Span Cast-in-Place Bridges
293
4.4 Span-by-Span Cast-in-Place Bridges
6.4.1 KETTIGER H,4,VG,
Construction
bar tendons
through
segments
not shown
The first application of’ the ot‘t-ground tnethodology (category 3), Section 6.1.2. was in 1955 on the
Kettiger Hang structure neat- Andernach (Federal
Highway 9), Figure 6.19.3 This system consists of
f’our scaffblding trusses of’ slightly more than a
span length and two cantilever girders of’ about a
two-span length. The scat‘folding trusses support
the entire c o nc rete w eig ht d u ring c asting . The
cantilever girders serve to transfer or advance the
scaf’folding trusses to the next span to be cast. The
concrete fortn or mold rides with the scat‘folding
trusses and is thus repeatedly reused.
h.-l.2
FIGURE 6.17. Lint1 Cole Viatiuct, segtnental
pier.
piers, using a stiff-leg derrick at the end of’ the
completed cantilevered portions of’ the structure,
Figure 6.18. When the temporary bents are no
longer required, they are dismantled and removed
bv equipment located on the completed portion of
the bridge deck.
GER.LC4.Yk
KRz4H.\‘E.\‘BERG
BRIDGE, GERA11.4.Y).
A variation of the of‘f-the-ground system was used
on the Krahnenbergbrticke
near Andernach constructed from 1961 to 1964, Figure 6.20.‘*3 This
structure has a length of’ 3609 f’t (1100 m), a constant depth of 6.56 fi (2.0 m), a width of’ 60.i f’t
(18.5), and spans of’ 105 f’t (32 m). The site is on a
slide-susceptible hillside, requiring difficult foundations, and its curved alignment follo\vs the topography, all of which economically favored the
span-by-span technique.
STIFF LEG DERRICK
LACING PRECAST SEGMENTS
FIGURE 6.18.
Linn Cove Viaduct, erection scheme for progressive placement.
39,26-39, M
Section l-l
Scaffolding truss at concreting position
Cantilever
%i%f
~39,20~3920~3Q,20
39,20-t----3Q,M-39,20
3%
Advancement of the Scaffolding truss including forms
Forward
Rear
slope
Section
2-2
0
n;
I
39.20-39.20
39,20--c-------39>20
39.20
,
39.20
Advancement of the cantilever beams
FIGURE 6.19. Kettiger Hang, schematic of’ the construction procedure, from reterence 3 (courtesy of’ Zement und Beton).
Exterior
scaffold
girder
I’I
‘\
Interior
scaffold
girder
m
t
t
(b)
294
Span-by-Span Cast-in-Place Bridges
In this project four fbrmwork supporting girders were used. Two interior girders were rigidly
connected together by transverse horizontal bracing. The formw ork was arranged so that the forms
hinged at the bottom and folded down to allow
passage, during advancement, past the piers, Figure 6.200. ‘The four girders were supported on the
hexagonal piers by transverse support beams attached to the pier. In this manner the four long itud inal formw ork su p p o rt g ird ers w ere su p ported on two piers, while an additional set of
transverse support beams were attached to the
forward pier. Figure 6.206.
Latticework cantilever extensions at both ends of
the lo ng itud inal formw ork su p p o rt g ird ers extended their length to twice the span length, so that
a stable support was provided by the transverse
support girders during advancement. The outside
girders had joints or links at the connection with
the cantilever latticework so that the curvature of
the structure could be accommodated during their
advancement. The elevation of the outside girders
was adjusted by hydraulic ja c ks to accommodate
superelevation. During the advancement operation the outside girders were advanced first and
then the center two girders, Figure 6.20~. When
the forward end of the interior girders reached the
transverse supporting beams, the rear transverse
beams of the previously cast span were no longer
required. They were dismantled from the pier.
These transverse beams were erected on the next
forward pier by a crane, Figure 6.20b.
The exterior formw ork of’ the two-cell box girder was attached to the longitudinal support girders and only required adjustment for curvature.
The interior forms of- the cells were dismantled
and reassembled on the next span after reinforcement was placed in the bottom flange and webs.
FIGURE 6.20. (Opposite). Krahnenberg Bridge, schematic of construction, from reference 1 (courtesy of the
American Concrete Institute). (a) Cross section. (6)
Formwork equipment in working position. (c) I: Working position: reinforcing, and concreting on formwork
equipment; installing the supporting construction on the
next following pier by means of derrick and straight-line
trolley. II: After concreting and prestressing: lowering
of equipment; opening of formwork flaps; shifting forward of outer girders; dismantling of the first rear supporting girder by straight-line trolley; intermediate storage at center pier. III: Partial pony-roughing of center
girder; dismantling and placing in intermediate for storage of second rear girder. IV: Final shifting forward of
center girders; jacking up of equipment; closing of
formwork flaps; new working position.
295
Average casting rate was 706 ft3 per hr (20 m3).
Fourteen days was required for construction of a
span.
6.43 PLEICHACH VMDUCT, GERMANY
In 1963 construction started on the 1148 ft (350 m)
long Pleichach Viaduct1a3
carrying a federal highway between Wurzburg and Fulda; it was the first
u se o f the sp an- b y - sp an tec hniq u e f o r a d u al
structure, Figure 6.21. Span length is 119 ft (36.25
Rear crane truck
rk
Forward crane truck
! ‘“__ -.-_-----~~-c-~~--~R
Fiv
1’; ” ,I ,A +;
I
//
__---
/
I
I
Scaffolding girder at concreting position
36.25+-x,25-
--i -36.25
Advancement of the scaffolding girder including forms
Construction joint
Advancement of the scaffolding and cantilever girders
R-Scaffolding
girder and forms
W-Scaffolding
and
cantilever girder
I, i I
I
I !
i
Cross section
FIGURE 6.21. Pleichach Viaduct, schematic of the
construction procedure, from reference 3 (courtesy of
Zement und Beton).
296
Progressive
and
Span-by-Span Construction of
Segmental Bridges
m), with each two-cell box girder having a width of
47.2 ft (14.4 m) and a depth of 7.2 ft (2.2 m). The
equipment
w as
superstructure
construction
erected behind an abutment in a position to construct one superstructure. Upon reaching the opposite abutment, the equipment was shifted laterally for the return trip to construct the other
superstructure. Because of the narrowness, only
one longitudinal support girder was required, as
o p p o sed to the tw o g ird ers req uired fo r the
Krahnenberg Bridge. This girder is slightly longer
than twice the span length. The two outside girders
are approximately one span length.
The outside girders were advanced simultaneously by a carrier traveling at the front of the central girder and at the rear by carriers running on
the deck of the previously completed section.
During concreting, the two outside girders are
supported on brackets at the forward pier and susp end ed fro m the co m p leted p o rtio n o f the
superstructure. The center girder, relieved of the
load of the two outside girders, is then advanced
one span and again connected to the outside girders by the hinged bottom formwork, thus functioning as an auxiliary support girder. This sequence of operations is commonly referred to as
the “ slide-rule principle.”
The piers have a width of 16.4 ft (5 m) and have
an opening at the top to allow passage of the central support girder, Figure 6.21. The width of the
pier is determined by the need for sufficient bearing area for the bearings and clearance for the
central support girder. Whether the central opening at the top of the pier should be concreted in is
one of aesthetics.
64.4 ELZTALBRUCKE,GER~~A~
Figure 6.22, was constructed
The Elztalbrticke,5,6
in 1965 at Eifel, West Germany, approximately
18.6 miles (30 km) west of Koblenz. It crosses the
deep valley of the Elz River with a total structure
length of 1244 ft (379.3 m), Figure 6.23. The
superstructure has a width of 98.4 ft (30 m) and is
supported on a single row of octagonal piers up to
328 ft (100 m) in height, Figure 6.24. Owing to the
height of the valley, conventional construction on
falsework would have been economically prohibitive. Therefore, a span-by-span system of selfsupporting traveling scaffolding was used, Figure
1.53.
The Autobahn between Montabauer and Trier,
which had been in planning before World War II,
FIGURE 6.22. El,t,dtmde,
\ ie\\
of
c0111plrteti
structure (courtesy of Dipl. Ing. Manfred Bockel).
had to cross two large natural obstacles, the Rhine
River north of Koblenz (see the Bendorf Bridge,
Section 2.2) and the Elz Valley. In 1962 tenders
were called for on the Elz Valley structure. Bidders
were provided with the grade requirements, dimensions for a single or a dual structure, the location of the abutments, and the foundation conditions.
A consortium of Dyckerhoff SC Widmann AG,
W a y s s SC Freytag K G , and Siemens-Bauunion
GmbH investigated four possible prestressed concrete construction possibilities?
1. A three-span variable depth structure similar
to the Bendorf Bridge
2. A six-span constant-depth structure
3. A frame bridge
4. A nine-span “ mushroom” construction w ith a
center row of piers
These four schemes were proposed, as were a large
number of different ones in both steel and concrete by other firms. The successful low bid was for
scheme 4 above. The nine-span “ mushroom” construction w as approximately 4% less costly than an
orthotropic-deck, three-span continuous steel
girder and 7% less costly than a prestressed concrete girder bridge of six spans.6
The Elztalbrticke, extending the methodology
used earlier for primarily low-level urban viaducts,
was the first application of the “ mushroom” cross
section for a high-level structure crossing a deep
valley. Previously, this type of construction, because of its short, stiff piers, required a number of
expansion joints in the deck to accommodate
thermal forces, elastic shortening, creep, and
shrinkage. In this structure, owing to the flexibility
of the tall piers, only one expansion joint was used,
Koblenz
east abutment
3m5-4 A
31aYlo
c
319,941
D
zyxwvutsrqponmlk
xu,514
E
F
6
321,128
321,779
322 4m
H
D&W rock anchors ; ; z
Longitudinal cross section
(a)
Total length = 379.30 m
~~~
~
Plan
(b)
FIGURE 6.23. Elztalbticke,
longitudinal cross section and plan, from reference 5
(courtesy of Der Bauingenieur). (a) Longitudinal cross section. (h) Plan.
I
Trier
west
abutment
298
Progressive and Span-by-Span Construction of Segmental Bridges
28
.
in the center span. This joint is located 38 ft (11.6
m) from pier E. The superstructure is monolithically connected at all piers and the abutments.
At the center of each span is a 43 ft (13.1 m)
long, massive flat plate, which in cross section has a
thickness varying from the centerline (crown ot
roadway) of 2% in. (650 mm) to 17% in. (450 mm) at
the outside edges. The “ mushroom” portion ot the
span varies in thickness, transversely and longitudinally, to 8 ft (2.45 m) at the pier. The
superstructure is prestressed longitudinally and
transv ersely.
The octagonal piers have, in cross section, external dimensions 01 15.75 by 19 ft (4.8 by 5.8 m) with
a w all thickness of 11% to 1% in. (300 to 350 mm).
Any given pier has a constant cross section for its
entire height. The percentage of vertical reinforcement, with a concrete cover on the outer and
interior faces of 1.5 in. (40 mm), varies from 0.8 to
1.2% of the gross concrete area. Piers were constructed by slip-forming. The eight pier shafts
were constructed in seven months. The tallest pier,
3 11.6 ft (95 m) in height, was slip-formed and cast
at a rate of about 26 ft (8 m) per day and thus required 12 days to construct. The top 4 ft (1.2 m)
portion of the pier was cast with the superstructure
by the traveling scaffolding. On the top of the slipformed pier four 7.2 ft (2.2 m) high pedestals were
cast to provide the support for the cantilever girder from the traveling scaffolding, Figure 6.25.’
The traveling scaffolding was assembled at
abutment A after completion of the abutment and
the half-mushroom projecting therefrom. This
form traveler, Figure 6.26, accommodates a fullwidth span-length segment of 123 ft (37.5 m).
After the first span, two weeks were required to
complete a superstructure span. The first opera-
zyxwvutsrqp
435
FIGURE 6.24. Elztalbticke,
cross section at pier E,
from reference 5 (courtesy of Der Bauingenieur).
FIGURE 6.25. Elztalbticke,
construction view (courtesy of Dipl. Ing. Manfred Bockel).
Side longitudinal girder
&Q*Llo
i-9000--+/
I
L+P-.--
dp
Center support bearing i
Center
girder
Catwalk
Upper
longitudinal
catwalk
Travel direction 37500
Longitudinal cross section
Center
girder
longitudinal
Side longitudinal
girder
Concreting
, + Hydraulic
sequence
II)
a Formwork
at
concreting position
I
Scaffolding after
advancement
concreting
traveling position
position
Section A-B
jack
III)
Forms in stripped
position
zyxwvutsrqp
Section CD
Cd)
Cc)
FIGURE 6.26. Elztalbticke,
form traveler, from reference 5 (courtesy of Der
Bauingenieur). (a) Longitudinal cross section. (b) Plan. (c) Section A-B. (d) Section C-D.
299
300
Progressive and Span-by-Span Construction of Segmental Bridges
tion was to cast a 42.65 ft (13 m) wide center portio n o f the b rid g e. A fter hard ening and initial
stressing, the two outside edges, each 27 ft (8.25 m)
wide, were cast. Subsequently the form traveler
was advanced to cast the next span.5
As mentioned previously, an expansion joint is
located in the center span. During construction this
joint was “ locked” until construction reached pier
G; then the joint was released.5
During concreting the forms are suspended by
steel bars, and during advancement the forms are
carried by the bottom arm of the transverse cantilevered steel members. The form traveler, Figure
6.26, essentially consists of two approximately 141
ft (43 m) long longitudinal girders and eight transverse frames in a “ C” configuration which surro u nd s the d ec k c o nstru c tio n. The transv erse
frames may be provided with a covering to protect
the w o rkm en and the c o nstru c tio n f ro m the
weather. At the forward end an approximately 72
ft (22 m ) long cantilever beam, located on the centerline, is projected to the next pier for support.
Fo rwa rd
This structure is lo cated o n natio nal ro u te 260
c ro ssing the G u ad iana Riv er b etw een Beja a n d
Serpa, Portugal. The viaduct has a total length of’
11 15 ft (340 m ) and consists of 197 ft (60 m ) spans
except for the river spans, which are 164 ft (50 nl).
Transversely, the superstructure is 53.8 ft (16.4 m)
in width composed of two single-cell box girders.
Each box girder is 19.35 ft (5.9 111) wide, with the
depth varying from 6.5 ft (2.0 nl) at midspan to 9.8
ft (3.0 m) at the piers. After construction of the box
girders, a longitudinal centerline closure is poured
and cantilevered sidewalks are constructed.
The superstructure is constructed by the spanby-span method, from inflection point to inflection
point, by an overhead self-launching f’orm carrier,
Figure 6.27. The fo rm carrier consists of 279 ft (X5
m ) long trusses of a depth varying
f‘ro nl 9.8 ft (3.0
m ) lo 16.4 f t ( 5.0 m ) . Forms fo l- c o nc reting the
superstructure are supported bv two series of suspenders. One set pierces the concrete flanges and
End tra ve le r suppo rt
sup p o rt ,
I
Ele va tio n
(a)
Typic a l c ro ss se c tio n
Se c tio n a t fo rwa rd suppo rt- fo rm s o pe n
(b)
Guadiana Viaduct, elevation and sections of form carrier. (a) Elevation.
(b) Section at forward support-forms open. (c) Typical cross section.
FIGURE 6.27.
Span-by-Span Cast-in-Place Bridges
is located inside the box cell. The other set is arranged outside the box and supports the forms
when stripped and traveling past the piers in an
open position, Figure 6.27.
D u ring c o nc reting o f the su p erstru c tu re the
f’orm carrier is supported on the forward pier b\
an arran g e m e n t o f a telescoping tubular cross
frame, at the rear: it is supported on the
superstructure at a location 26 ft (8.0 m) forw ard
of’ the rear pier. When the form carrier is being
launched forward, it moves over a support at the
tip o f rhe c o m p leted su p erstru c tu re c antilev er
(near the inflection point), and its rear support
rides on the surface of the completed superstructure. ‘The form carrier (including all equipment)
weighs 209 tons (190 mt).
6.4.6 LOISACH
BRIDGE, C;ER,\lA,\‘k
‘l-he federal autobahn between Munich and Lindau has an alignment that transverses the Murnauer swamp area near Ohlstadt and thus crosses
the Loisach River and the old federal highway B-2
(Olympiastrasse),
Figure 6.28. Because of flooding
and poor soil conditions an embankment was not
possible, and a decision was made requiring a dual
viaduct bridge structure with a total length of 43 14
f’t (1315 Ill).’
.I‘he 232.8 ft (70.96 m) main span crossing the
Loisach River is a variable-depth single-cell box
girder constructed by the free cantilever method.
Depth of’the box girder varies from 9.84 ft (3.0 m)
to 5.58 ft (1.7 m), Figure 6.29. The approach spans
are of a T-beam cross section, Figure 6.29, constructed by the span-by-span method with the form
carriers running below the superstructure. Figure
6.30 is a longitudinal section of the bridge within
the area of the approach spans, showing the form
carrier running below the level of the top slab. Figure 6.31 shows the form traveler in action.
Box girder
301
T-beam
FIGURE 6.29. Loisachbriicke, cross sections, from
reference 8 (courtesy of Dyckerhoff & Widmann).
The dual structure has a total width of 100 ft
(30.5 m), Figure 6.29, and each half is supported
on two circular piers, excepting the Loisach span
w hich is sup p o rted o n w all p iers. In the to tal
length, the dual structures are subdivided into
three sections by two transverse joints, Figure 6.28.
In plan the structure has a radius of 4265 ft (1300
m) at the Munich end, and the curvature reverses
at the Loisach with a radius of 6562 ft (2000 m).”
The completed structure is shown in Figure 6.32.
The circular piers are 4 ft (1.2 m) in diameter
and are supported on 20 in. (500 mm) driven piles
with an allowable load capacity of 176 tons (160
mt). Pile depths vary from 42 to 72 ft (13 to 22 m).
A total of 1182 piles were driven for a total length
of piling of 63,650 ft (19,400 m), with an average
length of pile of 53.8 ft (16.4 m). Load capacity of
the piles was determined from eleven load tests
taken to 265 tons (240 mt).
Bec au se o f the p o o r so il c o nd itio ns and
ground-water pressure, the substructure was con-
FIGURE 6.28. Loisachbriicke, layout and underside view of bridge, from reference 8
(courtesy of Dyckerhoff & Widmann).
302
Progressive and Span-by-Span Construction of Segmental Bridges
FIGURE 6.30. Loisachbriicke, longitudinal and cross section showing form traveler
(courtesy of Dipl. Ing. Manfred Bockel).
FIGURE 6.31. Loisachbriicke, view of form traveler
in action (courtesy of Dipl. Ing. Manfred Bockel).
quired at midspan. The radius and superelevation
in a support length were held constant. Superelevation varies from +5.5 to -4%. For a normal span
8830 ft3 (250 m3) of concrete were placed in nine
hours.s
Because of the tight time schedule, work was
continued through the winter months in defiance
of the extreme harsh weather conditions in the
Loisach Valley. A weather enclosure was mounted
on the form traveler and heated by warm-air blowers. In this enclosure the reinforcement and preheated concrete was placed. In addition, the fresh
concrete was protected by heat mats. In this manner the work could proceed up to an outside temperature of 5°F (- 15°C). Construction cycle per
span was gradually reduced, after familiarization,
from an original 14 days to seven days. Following
completion of the western roadway up to the
Loisach the form traveler was transferred to the
eastern roadway for the return trip to the Munich
abutment. All 38 spans on the Munich side were
completed by the end of February 1972, saving
nine weeks in the construction schedule.
On the Garmisch side of the Loisach the movable
scaffold system consisted of four principal girders
292 ft (89 m) in length and 9.8 ft (3.0 m) deep,
Figure 6.33. Superelevation varies from +4 t o
-5.5%.
FIGURE 6.32. Loisachbticke,
view of’ completed
structure (courtesy of Dipl. Ing. Manfred Bockel).
strutted in pits enclosed by sheet piling. The round
piers vary in height from 9.8 to 23 ft (3 to 7 m).
Because of the delay in pile driving, resulting from
the soil conditions, the foundation completion was
delayed from October 1970 to April 1971.
The 73 T-beam spans were constructed with two
span-by-span form travelers whose operations
were synchronized. On the Munich side of the
Loisach four 223 ft (68 m) long and 4.26 ft (1.30 m)
high principal form support girders are supported
in the 100 ft (3 1 m) spans on cross beams at each
pier, which in turn are supported off the pile caps.
For the longer spans an auxiliary support was re-
Because of the delay in the pile driving, the first
span was started in December 1970 with a 12-week
delay. The last approach span on the left of the
Garmisch side was completed in August of 1971.
The traveler was then transferred to the other
roadway for the return trip and all 35 bridge spans
were completed by March 1972. By a gradual reduction of thk work cycle from 14 days to seven
days, nine weeks were saved in the construction
schedule. Not only was the loss of time resulting
from the foundation work made up, but a time advantage was attained.
The four box girder spans (two in each dual
structure) on either side of the principal span over
the Loisach were cast on stationary falsework. Auxiliary cross beams to support the falsework girder
were supported on driven piles. The two main
Span-by-Span Cast-in-Place Bridges
303
ers were transferred to the opposite pier for the
remaining seven segments.”
After a construction time of approximately 30
months the bridge was completed in 1972, shortly
before beginning of the Olympic Games.
6.4.7 RHEINBRikKE DUSSELDORF-FLEHE,
G ER M A N Y
FIGURE 6.33. Loisachbrucke, cross section of movable scaffold system, from reference 8 (courtesy of
Dyckerhoff & Widmann).
spans of 232.8 ft (70.96 m) were constructed by the
free cantilever method. Thirteen segments of 16.4
ft (5 m) were required; six segments were cast
from one pier and then the cantilever form travel-
FIGURE 6.34. Rheinbticke
erhoff & Widmann).
This is an asymmetric cable-stayed bridge with an
inverted concrete Y-pylon, Figures 6.34 through
6.37. The overall length from abutment to abutment is 3764 ft (1147.25 m). The Rhine River span
is 1205 ft (367.25 m) long and is a rectangular
three-cell steel box girder with outriggers to support a 135 ft (41 m) wide orthotropic deck, Figures
6.36 and 6.37. At the pylon there is a transition
from the steel box girder to prestressed concrete
box girders, which are used for the thirteen 197 ft
(60 m) spans in the approach viaduct. The structure is continuous throughout its entire length,
having expansion joints only at the abutments.
The approach viaduct has from pier 9 up to pier
13, Figure 6.37, a five-cell box girder cross section
with a width of 96.8 ft (29.5 m) and a depth of 12.5
ft (3.8 m). This heavy cross section, Figure 6.36,
resists the anchorage forces from the cable stays.
For the balance of the viaduct length from abutment to pier 9 the cross section consists of two
single-cell boxes, a continuation of the exterior
cells of the five-cell box girder cross section. However, the interior webs of each box are of less
Dusseldorf-Flehe, artist’s rendering (courtesy of Dyck-
304
Progressive and Span-by-Span Construction of Segmental Bridges
:
FIGURE 6.35. Rheinbticke Dusseldorf-Flehe, view
from construction end of approach viaduct looking toward the pylon under construction.
thickness than that of the five-cell cross section.
The width of each box then becomes a constant 23
ft (7.0 m) outside-to-outside of webs. A diaphragm
occurs at each pier.
The approach spans were constructed segmentally by the span-by-span method with construction
joints at approximately the one-fifth point of the
span. As described in Section 6.1.2, the method
used here employed movable falsework, Figures
1.54 and 6.38, supported from the ground. The
197 ft (60 m) spans were poured in place in one
unit from construction joint to construction joint.
This required continuous placement of as much as
3200 cubic yards (2500 m3) of concrete. After each
section was cast in place and reached sufficient
strength, the prestress tendons were stressed and
the falsework was moved forward to repeat the
cycle.
6.4.8 DENNY CREEK BRIDGE, U.S.A.
The Denny Creek Bridge is the first implementation of the span-by-span method of construction in
the United States. It is located a few miles west of
Snoqualmie Pass in the state of Washington and
will carry the I-90 westbound traffic down off the
pass. It is a three-lane, 20-span, prestressed concrete box girder design with a total length of 3620
ft (1103 m) on a 6% grade, Figure 6.39. The contractor, Hensel Phelps Construction Company,
elected a construction method similar to those used
in many German and Swiss designs where the area
is environmentally sensitive.
Because of the ecological and environmental
sensitivity of the project site, construction of the
piers was carried out under extreme space restrictions. The contractor was allowed a narrow access
road for the full length of the project and additional work and storage area at each pier.”
The 19 pier shafts have a hollow rectangular
cross section with exterior dimensions of 16 by 10
ft (4.88 by 3.05 m), a wall thickness of 2 ft (0.61 m),
and heights ranging from 35 to 160 ft (10.7 to 48.8
m), Figure 6.40. Twelve piers are supported on rectangular footings. The other seven piers are supported on pier shafts sunk through talus and till
and keyed into solid bed rock, Figure 6.41. Piershaft diameter is 12 ft (3.66 m) with a maximum
depth of shaft below the terrain of 80 ft (24.38 m).
The superstructure was constructed in three
stages, Figure 6.42. In the first stage, bottom flange
and webs were constructed from a 330 ft (100 m)
long movable launching truss, Figure 6.43. The
two trusses used for constructing the “U” portion
of the box section rested on landing wings at the
piers, Figures 6.44 and 6.45, as the launching truss
moved up the valley, sliding from pier top to pier
top. The construction schedule called for one span
every two weeks. The entire scaffold system was
supported on six jacks to adjust for proper alignment, two ja c ks at the rear of the span or initial
pier and four ja c ks at the advance section or next
pier.
The launching truss was designed to support the
outside steel forms of the box section, Figure 6.46,
and to facilitate removal of the inside forms,9 Figure 6.42. Track-mounted cranes installed at the
top of the truss frame lifted and moved the inside
forms from the web, hanging them on the truss so
that they were moved forward with the advancement of the launching truss. Figure 6.47 is an
interior view of the working area between trusses.
Visible are the overhead track for the 15 ton ( 13.6
mt) cranes located near each web. Also visible are
the cable hangars from the roof frame for the
bottom slab support during casting.
The steel trapezoidal box form used for con-
Steel Superstructure
Reinforced Concrete Superstructure
Heavy Section
.*I
t.*LL-
Reinforced Concrete Superstructure
Normal Section
FIGURE 6.36.
Rheinbticke
Dusseldorf-Flehe, elevation of pylon and cross sections.
13. 60.0
* 760.0
367.25
SD
11‘7.26
Bearingcordnion6
+
+
zyxwvu
ffl
+
+
+
+
+
+
+
+
+
+
+
G4
FIGURE 6.37. Rheinbticke Dusseldorf-Flehe, plan and elevation.
+ unreai3aimd
=windbeE&lg
+ u-nad
Span-by-Span Cast-in-Place Bridges
FIGURE 6.38. Rheinbticke Dusseldorf-Flehe, end
view of girder.
struction was insulated w ith Sty ro fo am , Fig ure
6.48, and had heat cables installed (actuated if
need be) to help maintain the temperature and
rate of cure. Also, heat blankets were available to
go over the section to reduce heat loss and maintain a constant temperature in cold weather.
Concrete was batched from a plant erected neal
the west abutment using the highway right-of-way.
‘[‘he contractor used three 8 cu vd (6.1 m3) ready-
FIG U RE 6.39.
307
mix trucks for mixing the concrete, which was then
pumped to the proper location. Superstructure
pours were about 300 cu yd (229.4 m3) and took
about nine hours, using two concrete pumps and
the track-mounted cranes installed in the truss
frame. Concrete strength required was 5000 psi
(34.47 MPa). The contractor obtained 3500 psi
(24.13 MPa) in three days using $ in. (19 mm)
aggregate. The 28-day strength ranged fi-om 6100
to 6600 psi (42.06 to 45.51 MPa).
In stage two the top flange between the webs was
placed. Metal f’orms, Figure 6.49, were supported
from the bottom flange and webs, Figure 6.42.’
In stage three the two top flange cantilevers were
placed, Figure 6.42, by a movable carriage that
rode on top of’ the box cast in stage two, Figure
6.42. Upon completion of’ stage three, the transverse prestressing of’ the top flange was accomplished. The completed section is 52 f’t (15.08 m)
wide, providing three traflic lanes.
The Washington DOT sponsored the design.
Three alternatives were prepared f’or bidding
purposes. One was an in-house state design; the
other two were prepared by outside consultants.
‘l‘he Dyckerhof‘t‘ & Widmann design proved to be
Denny Creek Bridge, perspective sketch.
Progressive and Span-by-Span Construction of Segmental Bridges
FIGURE 6.40. Denny Creek Bridge, view of piers
under construction (courtesy of J. L. Vatshell, Washington DOT).
the most economical. VSL Corporation was the
subcontractor providing the prestressing expertise.
6.5 Span-by-Span Precast Bridges
6.5.1
LONG KEY BRIDGE, U.S.A.
Long Key Bridge in the Florida Keys carries U.S.
Highway 1 across Long Key south to Conch Key.
The existing bridge consists of 2 15 reinforced concrete arch spans, ranging in length from 43 to 59 ft
(13.1 to 18 m) for a total bridge length of 11,960 ft
(3645 m).
The new bridge, presently under construction, is
50 ft (15.2 m) between centerlines and just north
and parallel to the existing structure. It is a precast
segmental box girder constructed by the span-byspan method and consisting of 101 spans of 118 ft
(36 m) and end spans of 113 ft (34.4 m) for a total
length of 12,144 ft (3701 m). The roadway width
between barrier curbs is 36 ft (11 m), Figure 6.50,
to accommodate a 12 ft (3.66 m) roadway and a 6 ft
(1.83 m) shoulder in each direction. Figure 6.51 is
FIGURE 6.4 . Denny Creek Bridge, substructure
types.
an artist’s rendering showing the precast V-piers
with the 7 ft (2m) deep box girder segments.
In the preliminary design stage three methods of
segmental construction were considered: balanced
cantilever, span-by-span, and progressive placement. The progressive placement method was discarded because it was felt (at the time) to be too
new for acceptance in U.S. practice. It was later
introduced on the Linn Cove Viaduct in North
Carolina (see Section 6.3.2).
This is the first use of a precast span-by-span
method in the United States. The segments are
transported from the casting yard to their location
in the structure by barge. The segments are then
placed with a barge crane on an erection truss,
which is supported by a steel grillage at the V-piers.
Each span has a 6 in. closure pour after all the
segments have been placed on the erection truss
and properly aligned. The essential operations are
indicated in Figure 6.52.
Segment weight is approximately 65 tons (59
mt). Each segment is placed on the erection truss
on a three-point support and brought into its final
position. It takes approximately four to six hours
Schematic of movable scaffolding
Stripped position
Staga 2
7
7
50
I
Overhead
dollies
Stage
two
T
Jacks for grade,
superelevation and camber
Stage one
LJacks
Rollers and jacks --
Stage three
FIGURE
6.42. Denny Creek Bridge, schematic of construction stages, from reference 9
(courtesy of the Portland Cement Association).
FIGURE 6.44. Denny Creek Bridge, view of landing
FIGURE 6.43. Denny Creek Bridge, view of launching truss.
wings at piers (courtesy of J. L. Vatshell, Washington
DO-I-).
309
FIGURE 6.46. Denny Creek Bridge, view of outside
steel forms (courtesy of J. L. Vatshell, Washington
DOT).
FIGURE 6.45. Denny Creek Bridge, close-up view of
landing wing (courtesy of J. L. Vatshell, Washington
D O T ) .
to place the segments required for one span. The
contractor has placed as many as three spans per
week for a total of 354 ft (108 m) of completed
superstructure per week and has averaged 2.25
spans per week.
Another major deviation from United States
practice in this project was the use of external prestressing tendons (located inside the box girder
cell). This requires that the tendons be considered
as unbonded for ultimate-strength analysis. Placing the tendons inside the box girder void allows
the w eb thickness to be m inim iz ed . Tend o n
geometry is controlled by deviation blocks cast
monolithically with the segments at the proper location in the span, Figure 6.53. These blocks perform the same function as hold-down devices in a
pretensioning bed. The tendon ducts between deviation blocks or anchorage locations or both are
composed of polyethylene pipe, which is then
grout-injected upon completion of stressing
operations- a corrosion protection system similar
to that used for the cable stays on some cable-stay
bridges. l”,ll
FIGURE 6.47. Denny Creek Bridge, view of interior
working area between trusses (courtesy of Herb &hell,
FHWA Region 10).
FIGURE 6.48. Denny Creek Bridge, insulation on
exterior steel forms with installed heat cables (courtesy
of Herb Schell, FHWA Region 10).
Span-by-Span Precast Bridges
Creek Bridge, vie\v 01‘ tnc.tal
form used for stage-two construction (courtesy of J. L.
Vatshell, Washington DOT).
FIGURE 6.49. l)c~lny
Section
FIGURE
6.50.
at
pier
311
The external tendons overlap at the pier segment to develop continuity. The bridge is continuous between expansion joints for eight spans, 944
ft (288 m). After the closure pour reaches the required strength, the post-tensioning is accomplished and the span is complete. A 30 in. (‘760
mm) diameter waterline is installed inside the void
of the box girder. The erection truss is then lowered and moved away from the completed span.
The erection truss is handled at a one-point
pick-up location by a C-shaped lifting hook, Figure
6.52. The truss is supported against the barge
crane and moved parallel to the new bridge until it
Section at midspan
Long Key Bridge, typical cross section of superstructure.
Progressive and Span-by-Span Construction of Segmental Bridges
312
The span by span erection concept utilizes a temporary steel assembly truss
In conjunction with a barge mounted crane as shown. The steel truss
3 between the piers is equipped
with post-tensiontng
tendons along
)m chord to facilitate adjustments for deflections and kwenng the
LIUw ,,on completion of the span.
PREVIOUSLY A55EMBLfP
5PAN 3
zyxwvu
,
\
i A55E’40~Y
#.
1
TRUSS
FIGURE 6.52. Long Key Bridge, span-by-span erection scheme.
reaches the position for a new span, and the cycle is
repeated.
6.5.2 SEVEN MILE BRIDGE, U.S.A.
The Seven Mile Bridge, Figure 6.54, in the Florida
Keys carries U.S. Highway 1 across Seven Mile
Channel and Moser Channel from Knights Key
west and southwest across Pigeon Key to Little
Duck Key.
The existing structure consists of 209 masonry
arch spans, 300 spans of steel girders resting on
_ _
masonry piers, a n d a s w i n g s p a n o v e r M o s e r
Channel. The spans range in length from 42 ft 7t
in. (13 m) to 47 ft 4$ in. (14.4 m) for the masonry
arches and from 59 ft 9 in. (18.2 m) to 80 ft (24.4
m) for the steel girders resting on masonry piers,
which along with the 256 ft 10 in. (78.3 m) swing
span, produce a total bridge length of 35,716 ft 3
in. (10,SSS m).
Span-by-Span Precast Bridges
313
PERSPECTIVE VIEW
DETAIL 2
ELEVATION
FIGURE 6.53. Long Key Bridge, typical tendon lay-
FIGURE 6.54. Seven Mile Bridge, artist’s rendering.
The new bridge, presently under construction, is
located to the south of the existing bridge. It is a
precast segmental box girder constructed by the
span-by-span method w ith 264 spans at 135 ft
(41.15 m), a west-end span of 81 ft 7$ in. (24.88 m),
and an east-end span of 141 ft 9 in. (43.2 m) for a
total length of 35,863 ft 44 in. (10,931 m). The
roadway requirements are the same as for the
Long Key Bridge and the cross section is almost
identical, Figure 6.50. Seven Mile Bridge crosses
the Intracoastal Waterway with 65 ft (19.8 m) vertical clearance, and its alignment has both vertical
and horizontal curvature.
The consultants, Figg and Muller Engineers,
Inc., used the same concepts as had been used for
the Long Key Bridge, except they omitted the
V-pier alternative in favor of a rectangular hollow
box-pier scheme that is precast in segments and
post-tensioned vertically to the foundation system.
As mentioned in Section 1.9.3, the contractor
elected to alter the construction scheme in this
bridge from that of the Long Key Bridge by suspending the segments from an overhead truss
rather than placing them on an underslung truss.
The essential operations for construction of a typical span are as follow s:
1.
Transportation of all segments by barge to the
erection site.
2.
Assembly of all segments in a span (with the
exception of the pier segment) on a structural
steel frame supported by a barge.
3.
Placing the pier segment on the pier adjacent
to the previously completed portion of the
deck with the overhead truss working in cantilever.
4.
Launching the overhead truss onto this newly
placed pier segment.
314
Progressive
and
Span-by-Span
Construction
5.
Lifting in place the entire assembly of typical
segments with four winches supported by the
truss.
6. Post-tensioning the entire span after the closure joint has been poured between the
finished span and the new span.
7. Launching the overhead truss to repeat a new
cycle of operations.
After a period of adjustment, the method has
allowed a speed of construction equal to that for
the assembly truss scheme used for the Long Key
Bridge. One complete span may be constructed in
one day, and as many as six 135 ft spans have been
placed in a single week. Figure 6.55 shows the assembly of segments being erected in a typical span.
6.6
Design Aspects of Segmental Progressive
Construction
6.6.1 GENERAL
.,
The use of temporary stays to carry the weight of
segments during construction induces only a normal compression load in the deck and a very limited amount of bending. Consequently, the static
scheme of the structure during construction is very
close to that of the finished structure. This is a
significant advantage over the conventional cantilever construction scheme, where continuity of
the successive cantilever arms results in two static
schemes significantly different between construction and service.
Because of this similarity of static scheme
throughout erection and service, it is expected that
the layout of prestress tendons found in cast-inplace structures or in span-by-span construction
FIGURE
span.
6.55.
Seven Mile Bridge, erection of a typical
of
Segmental
Bridges
should be applicable to progressive construction,
with the added advantage that the tendons can be
regularly stressed and anchored at the successive
joints between segments in a simple manner.
On the other hand, progressive construction differs in several aspects such as pier design and
deflection control during construction, calling for a
more detailed examination.
6.62
REACTIONS ON PIERS DURING
CONSTRUCTION
Construction of a typical span proceeds in two
stages, as shown in Figure 6.56: (1) pure cantilever
erection, of a length a from the pier, and (2) construction with temporary stays on the remaining
length (L - a). Length a should be selected (within
the nearest number of segments being placed) such
as to keep the girder load moments over the pier
within allowable limits.
Assuming that this moment is of exactly the
same magnitude as the fixed end moment of a
typical span under the same unit load W, one may
write:
z
Wa2
WL’
-=-
2
FIGURE
12
6.56. Progressive construction, deck reactions on piers.
315
Design Aspects of Segmental Progressive Construction
for a constant-depth girder, which is the general
case for- progressive construction. Thus:
over the support
(15%), 2.6 x 1.15 =
3.0
12.5 ksf‘
0 = 0.408L = 0.4OL
For (1 = 0.4OL the moment over the pier is equal to
,M = 0.08WL’. l‘he moment over the preceding
pier, for a structure with a large number of’ identical spans, is equal to 0.26&\1. Therefbre. the reaction over the pier at the end of’ this first stage of
construction can be easilv computed as:
-I‘he dif‘ference i s s m a l l a n d usuallv m o r e t h a n
offset hy the fact that horizontal loads during construction are smaller than during service.
R = 0.4OWL + 1.268 x 0.08WL = O..5OW’L
During the second construction stage the lveight
of the remaining part of’ the span is supported b:,
t h e temporarv stays, which are anchored in the
rear span as close as possible to the previous pier so
as not to induce undesirable variations of. moments
i n t h e last c o m p l e t e d s p a n . ConsequentI!,, t h e
lveight of’t hat part of’t he span induces in the pier a
react ion equal to:
1 .io
0.6W’L + - = 1.02WL
1 . o o ___
The total reaction during construction applied to
the pier is t bus:
R = 0.5OWL + 1.02WL = 1.62WL
as opposed t o R = N’L for cast-in-place or spanh\--span construction. ‘l‘his temporary increase of’
girder load reaction of’ 62% \vill eventuall!- \,anish
Ivhen construction proceeds. It is important to
\.erifv how critical this pier temporary overload
ma! be f’or the design of’ the substructure. Taking
the example of’ a 150 to 200 f’t span, the average
loads are as follows fi)r a 40 f’t wide bridge designed f’or three lanes of’ traf’fic:
Girder load
Superimposed load
Equivalent live load including impact
8.0 ksf
1.5 ksf
2.6 ksf
The maximum reaction during construction compares jvit h that after completion as follows (values
given are the ratio between reaction and span
length):
1. During construction, 1.62 X 8.0 =
2. Completed structure:
a . Girder load
8.0
b. Superimposed load
1.5
C.
Live load, including
provision tot- continuity
13 ksf
As shown previously, progressive construction of’ a
typical span entails two successive stages:
Cantilever construction on a length (I
~l‘emporary suspension by stays on the remaining
part of’ the span (L - n)
.I‘his second stage induces small deflections and
rotation, provided that the vertical component of’
the sta!- loads balances the total deck weight. On
the other hand, the first-stage construction not
only creates substantial deflections but also changes
the geometric position of’the entire span, as mav be
seen in Figure 6.5f.
The xveight (Wa) of’ the deck section produces:
A rotation of’ the previous span, w,, which will
project at the f’ollowing pier and create a vertical
deflection, J,
a deflection of’ the cantilever proper, yr
a rotation at the end of the cantilever, wL, which
Lvill p r o j e c t a g a i n a t t h e f’ollowing p i e r i n t o a
deflection mt (L - n)
Altogether the total deflection is:
Wa’ (2Ll v5 + 4nL - n’)
I’= 2 4 E I
If’we let 14 = N/L, the deflection can then be written
as:
WL1
u’(2 fi + 4U - u’)
)‘= 2 4 E I
With u = 0.4 as assumed betore, the total deflection is:
WL4
y = 0.0327 EI
where W = unit deck load,
L = span length,
316
Progressive and Span-by-Span Construction of Segmental Bridges
IZ’ = concrete modulus,
I = sectio n inertia.
A simple parametric analysis will reveal the importance of this problem. If W is the specific gravity
of concrete and A the cross-sectional area, then W
= GA. It was shown in Chapter 4 that the efficiency
factor of a box section is:
=
I
- = 0.60 to 0.63
Ac,c,
If the section is symmetrical, c, = cz = 0.5 h (h =
sec tio n d ep th), and I = 0.157 Ah” m ax. If c, =
0.33 h and cg = 0.67h, which is the practical dissymmetry of a box section, I = 0.133Ah2
min. For
all practical purposes, assume I = O.l4Ah*.
The deflection then becomes:
Ey = 0.23ZL2 + *
t 1
L-o
Bec ause the c o nstruc tio n p ro c eed s rap id ly , E
should be taken for short-duration loading; that is,
E = 800,000 ksf; the specific gravity of concrete is W
= 0.15 kcf. The slenderness ratio L/h varies between 18 and 22. Results are shown in Figure 6.58.
Construction of a 200 ft span, for example, with
a slenderness ratio of 20 will be accompanied by a
deflection under girder load (without prestress) at
the next p ier o f 8.3 inc hes. The c o nstru c tio n
method is therefore very sensitive to concrete
deflections, which are magnified by the great lever
arm of the first-stage construction of the span
projecting its intrinsic deformation to the following pier.
Fortunately, prestress will give a helping hand
and c o ntrib u te to su b stantially d ec reasing the
girder load deflection. The minimum prestress required at this stage is to balance the tensile stresses
induced by the girder load moments. With the
same notations as above, one may compute the
prestress force and the corresponding moment for
three positions of the neutral axis:
Efficiency factor
Distance fi-o m
centroid of‘ prestress to top fiber
Eccentricity o f’ p restress
Lower central core
Lever arm of prestress
Prestress mo ment
(ratio o f’ g ird er lo ad mo ment)
1
FIGURE 6.57. Progressive construction. def’ot-matio ns.
For an efficiency factor p = 0.65 the corresponding
values would be:
0.58
0.47
0.39
The prestress will therefore reduce the deflections
by the same amount-that is, approximately half
the to tal g ird er lo ad d eflec tio ns. The resu ltant
deflection (girder load + prestress) still remains
very significant as soon as the span length is above
150 ft. These deflections must be taken into full
account to compute the camber diagram (for segment precasting).
The next important point to consider here is the
second-stage construction of a typical span when
the remaining part of the girder is suspended from
the temporary stays. The concrete girder and the
group of stays form an elastic system that supports
the applied loads: girder load for the segments alread y in p lace, swivel crane and new segment
c,lh = 0.5
c,lh = 0.5
c,lh = 0.4
c,lh = 0.6
c,lh = 0.33
c,lh = 0.67
p = 0.60
d = 0.05h
p = 0.60
d = 0.05h
p = 0.60
d = 0.05h
e = 0.45h
72/c, = 0.30h
flc, = 0.36h
0.75h
0.71h
0.45 = 0.60
0.75 E
c = 0.35h
c = 0.28h
0.40h
r21c, =
0.68h
317
Design Aspects of Segmental Progressive Construction
zyxwvutsrq
IC
I
I
where progressive construction was contemplated
for a viaduct with a large number of identical 260
ft (80 m) spans all made up of 26 segments 10 ft (3
m) long. Figures 6.60 and 6.61 show the distributio n o f m o m ents b etw een c o nc rete g ird er and
temporary stays at three successive stages of segment placing: segments 15, 20, and 25, respectively. The first nine segments are placed in cantilever; the following 15 segments are suspended
from tern porary stays, while the last typical segment and the adjacent pier segment are placed
without stays.
The proportion of the load (and corresponding
moment) taken by the stays increases as the cantilever length increases and, when the last segment
is placed, more than half the load is supported by
the stays. For verv-long-span stayed bridges, this
distribution of load between stays and concrete
girder reaches the situation where the load is almost entirelv supported by the stays and the concrete girder’is subjected only to an axial force, except in the area of the longest stays.
The consideration of distribution of loads and
moments between stays and concrete girder has an
important aspect during construction-that is, the
accuracy of the tension in the stays and consequences of an accidental deviation between computed values of stressing loads in the stays and
their actual values in the field. For example, take
the simple case of a span L with 40% built in pure
cantilever and the remaining 60% suspended by
stays (see Figure 6.56). The moment over the pier
due to the second-stage construction load is M =
0.42WL’. Assume that an accidental deviation took
place of 5% between the design loads for the stays
and the actual values obtained in the field (owing to
friction in the jacks, inaccuracy in the pressure
gauges, and so on). As a result, an additional moment will appear over the pier of AM = & 0.42 WL’
= 0.021 WL’. The corresponding tensile stress at
the top fiber (assuming the error in stay loads was
to reduce the theoretical values by 5%) can be easily computed by:
FIGURE 6.58. Progressive construction, deflections.
traveling over the bridge with the trailer and tractor. -Two examples have been considered to show
the relative response of the various components of
this elastic svstem toward the application of a load.
1. 108) (JJm) s p a n This was one typical span
of’ the Rombas Viaduct. The span has been assumed to be completed except for the pier segment
over the next pier. For this construction stage, the
swivel crane and the new segment apply to the
staved cantilever a load of 88 tons (80 mt). In view
of the great stiffness of the concrete girder compared to the group of stays, the total moment induced bv the load remains ahnost entirely in the
concrete girder and there is only a small spontaneous increase of the stay loads, as shown in Figure
6.59. The magnitude of temporary prestress in the
deck must be designed accordingly to keep all
jo ints u nd er c o m p ressio n f o r all interm ed iate
loading cases.
Af=AM’-cI
0.02 IZAL2c,
APC,C,
= 0.0217X~
PC2
With W = 0.15 kcf, p = 0.60, and cp = 0.60/ z:
2. 260 ft (80 m) span
This example is taken
from a recent design for a large project in Europe
Af=
0.0088:
318
Progressive
and
FIGURE 6.59.
Span-by-Span
Pt-ogwssi\t’
Construction
co11structiotl.
‘l‘hc stress in kst’ t’or L//r = 20 (slcndert~ess ratio) i3
the l’ollo~vit~g 1’01. sewx~l spmi l e n g t h s :
L (11)
.I/ = O.li.il.
100
(tdl)
1X
130
“00
250
of
Segmental
incwase
Bridges
of’ hta\ Ioatlittg
A-\ssutiie that the itiaccutx~~ of the sta\ lo;itts lea\,es
itI the concrete girder 5% of its O\VII lveigtit to be
carried lx bending: the resulting deflection m.et
the pier ~vould be:
“6zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
3 .i
-l-l
I‘his stress is not critkil 1’01. short spans but I~;II
hecome sigtiificatIt t’or lotig ones. .I‘he simple dervation given above sho~3 that control of the stabtetisiottitig operations at the site shoitld albx~~s be
ott the salt side Ivith due allo\vatice fol. iti;icciIr;ic~~.
.-I dcviatioti in the tension of‘ the stavs bill aiso
al‘f’ect the deflections during constructio;l. Without
the presence of the stavs the total deflection over
the next pier due to the load 011 the length (L - n)
w o u l d be:
which gives t’or u = 0.4 as befixe:
.I‘his value should be compared to ttte ef‘tect of‘ the
first-stage cotistructiotl, lvhich bxs pre\iortsl\ gilen
:1s:
Ill sLltllm;II‘~
( il 5% de\iatioti
of the st;t\. tctisiotl
loads will increase the cantilever defectiotl due to
girder load by 36%. Considering the twneficial etfeet of prestressitlg fi)r the latter, \ve see that approximately 7% deviation ot’the sta!. l0;1d produces
the same defection 2s the first-stage cotistructiot~
loads including prestressing. .I‘tiis 41~0~3 that the
d e f l e c t i o n s a r e itnportant, particulart\ fi)r lotlg
spa~is built in progressive construction, hut that
proper deflection control is an excellent tool to
319
References
BY CONCRETL
3~
S.GML”T,
*~ ~
~~~
I” CANmEYER
o
o
f
lttolttclt1
I’ ro g rc \ si\ c
tx~t\\wll
sl;l\s
~o tt~tt~tt~~tio tt. c listt- ib rttio tt
;Illcl
gitxlcT.
I5 ?2WR”DED
SCfMLNTS
- - - - iw’ 9*h
zyxwvuts
FIGURE
FIGURE 6.60.
OIMDCR
Let-it\ t ttal st xwcs in t tie c-ottc‘rele girder ,tt-e ;11\\-;t\ 5
Lcpl \vittiitt alfo\v;lt)fe littiila.
f
6.61.
I’ro~t~c’~~i\
c
(
IlIoItlcI1t twt\\cul
\t,t\s
lion. (li\tt~itttttiott
c o nsli~uc
‘Incl
~itxlcl
Possit)l\. ;I third fatnil\ 01’ tendotta mtde of intet~tt;il
st;i\ s b,‘itli ;I dl-aped profile a nd attchot-ed over the
piet-s iii t tie di~tptirqp~. the put.posc of’ lvttictt is to
sttpplettietit both ottiet- f;itiiilies Ichile t~ecittcittg the
ttet stie;tt- sttwses in the \vet)s txcattse of’ the \,ctTiC d c o tiip o tie tit
o f prestt‘ess.
References
Because the silatic wttente at the cticl of’ cacti conslrttctiott 5tep ia idctttical to that of‘ ;I ca\t-itt-place
slntct tire, (tie pet~tttatietit tendotis c a n Ix ittstalletl
i i i t t t r sttxtctttw itiitiiediatel~, vAthut the tratisit i o t i s i t u a t i o n s I-eqttil-ed t,y o t h e r cotistrttctioti
ttt~tlioclologies
such 2s itict-emetital lautictiitig.
~1 t\ pical pi-esttws la \ o u t for progressive COIIstruction b,iff thus include:
1
\Vitttoht.zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPON
"l'lY?sllw\ed ~:oIIcl~cIc
Rlxlge
c:or1-
H .
sti- tic tio tt
b itti
sig n,
t’ ;~ p c r
SP
H .
\ Vittto ttt.
2% 2X.
tuiig e il b ic tn
.A sso c ia tio n
Siitttt
3
H.
Itlstitute .
iii Brid g e
B ri d g e
He ft
M;rl~-<:tlullg
a ~ td
im
4”. Ik/ c tttt,e l~
I‘ mg,
lkti- o il.
I‘ ttc
\~ol-scttt)l-its-
L‘ w
Sttxc tui- a l
.\ l;r\ H-
lb
SI’- 23.
lWi9.
VOII
C o nstrttc tio ii),
C o ng re ss. ;\ ttta te t- da ~tl,
13ritlg c
l’t~t~lic ~ttio tt
\‘ei-~eiidrtttg
(
E:q ttip ilit’ itt.”
(Zorict-ctc
.- \ C:l
Bt~ iic ke ithu”
I‘hul. “Sp a tttttxto n
He m.
-I
“Die
Fo rm \ \ - 0t.L
tttlx~sit~itt
S\
.\ tttt~ C c ;ttt C:ottc t-e te
”
Foi~rn~~o~L
.-I fit-t f;ttttil\ of. tettclotts located in t h e t o p flange
o\‘et‘ 1tie \~t~ious piers, h,itti atictiol-s s~tiitiietl-icall~~
loc;tlecl in hfistet-s, the purpose of Ivtiicli is to resist
ticgtti\ c tttotttetits 01 er t tic suppot-ts.
Ste p p ittg
Firs t Iitte t- ~ ta tio rta l
of
I‘ i- :t\ e liitg
Itttc ’ i~ n;ttio tt;tI
kkg ittc e riitg ,
13,
Briic Le lltxttt.”
lYi2.
Za wrtl
rrtc d
196X.
“ Re c e n t lkve lo l~ itie rtt
o f C o n-
.A wcottcl f’atttil\ of‘ tendons located a lo ti~ the spa ti
struc tio tt
~I‘c c hnic ltie s in (Ionct-etc
Brid g e s . ” I’ ra nsiii the tmtlom flatige a nd Am ;itictiored in blisters
p o t- ta tio it Ke swrc h
Ke wi- tl
66.5. Krid g r Ettg tne e l- inside the tms section. L’sua ll\ . the top a tid bo tto m
ing . \‘ot.
2. Pro c wd~nq ~ of the 7‘,n,~.rportcctrorl
RPWIW/I
hlisterh ~tre,joitieci to ;t \veb rib, allowing tetiipotxt-1
Rof~rd Co,,Jrrf~rrcf~, Se p te m b e r 2.5’Li . 19i8. S t . Im ttis;,
pi-estt-css hi-s t o Ix atictiored during s e g m e n t
.\fo.,
Sa tio ita l Ac;tdeiii\~
o t Sc ie n c e s . ~ l’ a stiittg to tt.
placittg.zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
D.C .
Progressive and Span-by-Span Construction of Segmental Bridges
320
5. U. Fin ste rw ald e r an d H .
“ D i e Elztal-
9.
AIlon.,
“ D e n n y
Der Buuzngenieur, H e f i 6, Ju n e 1966, an d
Washington,”
Heti
January
Cement
6. H .
1,
1967.
.I‘ hul, “ Bt-iickenbau,”
Heft 5, May
7.
Schambeck,
briicke,”
Anon.,
Beton- und Stnhlhetorrbnu,
1966.
“Ba u del- Loisachbriicke
Dyu&g-Berichte
bei O h l s t a d t , ”
19713, D y c k e r h o f t 8s N’idmann,
8. .A non., “ Bauausf‘iihrLlng
Loisach bei
Dvckerhoff
Anon.,
Association,
“ Florida’s
Ohlstadt,”
& Widmann,
d e r XutobahnbrLicke
Dydq-Betichte
AG. 1lunich.
ilbel
1972-5,
Creek-FrallkliII
Report SK
Skokie,
111..
E‘,~lls \‘i;ttluct.
202.01
E, l’ortl;t11(1
1978.
L o ng E(e) Bridge to Ltilile
l’rcca\t
Segmental Box Girder Span-b\-Spa11
<:onstI.uctiot~,”
Bridge Report, Post ‘I‘ensioning
IIIstitute.
l’hoellix.
Arizona, January
1979.
11. W alte r Podolny,
AC, M u nic h.
die
10.
Bt-idge
Jr. , “ ;\n O\ et-k iew
of t’reca\t Prc-
stressed Seynental Bridges.“ Jo~c,-,/ NI of / AC, ~w\l,r\wtl
Concrete Ztt,ditute. 1.01. 24, So. 1, jaI ~I I ;in -Fe111 udn
1979.
7
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCB
zyxwvu
Incrementallv Launched Bridges
J
7.1 INTRODUCTION
7.2 RIO CARONI, VENEZUELA
7.3 VAL RESTEL VIADUCl-, ITALY
7.9
7.9.2
7.9.3
7.4 RAVENSBOSCH VALLEY BRIDGE, HOLLAND
7 . 5 OLIFANT’S RIVER BRIDGE, SOUTH AFRICA
7.6 VARIOUS BRIDGES IN FRANCE
7.8.1 Mtihlbachtalbriicke,
Germany
7.8.2 Shepherds House Bridge, England
7.1
Introduction
-The concept of’ incrementally launched segmental
pres’ressed concrete bridges was described in Sectio n 1.9.5. .Fhis chap ter w ill d escrib e the im plementation of this innovative concept in several
representative
projects.
Since the in~plementation o f the inc rem ental
launching techniq ue on the Rio C aro ni Brid g e,
some eight\ bridge superstructures have been consrructed 1;~ t h i s m e t h o d t h r o u g h 1 9 7 6 , w i t h
g rad u al ref inem ents and im p ro v em ents in the
method.’ Bv concentrating the casting of segments
behind an ;Ibutment with a temporary shelter, if
required, this method can provide the same quality
control procedures and quality of concrete that can
b e ac hiev ed in a c o nc rete ‘ p rec asting p lant. It
minimizes temporarv falsework, extensive forming, and o ther teniporary exp ed ients req u ired
during construction bv the conventional cast-inplace on falsework meihod. Basically the method
entails incremental fabrication of the superstructure at a stationarv location, longitudinal movem ent o f the fabridated seg m ent an inc rem ental
OF
INCREMENTALLY
LAUNCHED
BRIDGES
Type, Shape, and Dimensions of Superstructure
Span Arrangement and Related Principle of Con-
struction
7.9.4 Design of Longitudinal Members for Flexure
7.6.1 Luc Viaduct
7.6.2 Creil Viaduct
7.6.3 Oli Viaduct
7.7 WABASH RIVER BRIDGE, U.S.A.
7.8 OTHER NOTABLE STRUCTURES
DESIGN
7.9.1 Bridge Alignment Requirements
7.9.5
and
Tendon Profile
Casting Area and Launching Methods
7.9.6 Launching Nose and Temporary Stays
7.9.7 Piers and Foundations
7.10
DEMOLITION OF A
TAL LAUNCHING
REFERENCES
STRUCTURE
BY
INCREMEN-
length, and casting of a new segment onto the one
previously cast. In other words, the procedure can
be considered as a horizontal slip-form technique,
except that the fabrication and casting occur at a
stationary location. Stringent dimensional control,
however, is an absolute necessity at the stationary
casting site, since errors are very difficult to correct
and result in additional costs in launching.’
Straight superstructures are the easiest to accommodate; however, curvature (either vertical or
horizontal) can be accomplished if a constant rate
of curvature is maintained. If the grade of the
structure is on an incline, it is preferred to launch
the stru c tu re, w herev er p o ssib le, d o w nw ard .
Where the fall is 2’$%, the superstructure has to be
p u shed o r held b ac k, d ep end ing u p o n the
coefficient of friction. Where the fall is in excess of
4%, special provisions are required to prevent a
“ runawav” superstructure during launching.’ To
the authors’ knowledge, this situation has never occ u rred . Piers, either tem p o rary o r p erm anent,
should be designed to resist the lateral force produced by the launching operation. A friction force
varying from 4 to 7% has been considered for de-
321
zy
Incrementally Launched Bridges
322
sign purposes, although values of’ only 2 to 34%
have been observed in the field.
At present, it is felt that this system cm be used
for superstructures up to 2000 t‘t (610 m) in length;
fbr longer structures incremental launching is acco m p lished f’rom b o th ab u tm ents to w ard the
center of’ the structure. .l‘he technique has been
ap p lied f’or s p a n s u p to 200 f’t (60 m) lvithout the
use of’ temporal-v supporting bents and for spans
u p t o 3 3 0 ft ( 1 0 0 m ) with such bents. Girders IISUally hav e a d ep th- to - sp an ratio rang ing f‘rom
one-tbvelfth to one-sixteenth of’ the longest span
and are of’ a constant depth. ‘l-he launching nose
has a length of’ approximately 60% of’ the longest
either straight or curved: holvever, cur\‘ature,
either vertical or horizontal, must be of’ a COW
stant radius.
2.
As mentioned above, strict dimensional control
d uring casting is req uired . .4n\ m istakes in
casting are difficult and expensive to correct,
especially if the\. are not discovered until af.ter
some length of’ bridge has been launched.
3.
l‘he superstructure must be of’ a constant section and depth. .l‘his is a disad\.antage in long
s p a n s , lvhere a v ariab le-d ep th sec tio n \vould
provide a better econom\ of’ materials.
4.
Considerable
area is req u ired b ehind the
abutment(s) for casting the segments. In some
project sites this may not he feasible.
spa11.
.I‘he p rincip al advantages of the
launching method are the following’:
1.
incremental
No f’alseworh is required f’or the construction
o f ’ the su p erstru c tu re o t h e r t h a n possibl!
f’alsework bents to reduce span length during
construction. In this manner cantilever stresses
d uring launching can be m aintained lvithin
allo~vable lim its. If‘ fhlsework b ents sho u ld
prove to be impractical, then a system of‘temporare stays can be used as indicated in Figure
I .63. Obviously, depending on site conditions,
;Inv or all combinations of’
temporary bents,
launching nose, and temporar!. stays may be
used, the point being that conventional use of’
f’alsework is cqreatlv minimized. -l-his is partitularly interesting f’or projects in urban areas
or spanning over water, highways, or railroads.
2.
Depending on the size of’the prqject there can
be a substantial reduction in form investment.
Because casting of’ the segments is centralized
at a location behind the abutment, the economic advantages of mass production and a
precasting plant operation can be duplicated.
3.
‘l-he method eliminates transportation costs of
segments cast at a fixed plant and transported
to the site.
4.
It eliminates heavy cranes or launching trusses
and associated erection costs.
5.
It eliminates epoxy joints. Since epoxy is not
involved, construction can continue at lower
temperatures.
6.
Camber control and other geometry controls
are easily obtained.
Disadvantages are as follows:
1. As mentioned in Section 1.9.5, bridge alignment fbr this type of’ construction must be
In the present state 01’ the art of i~icrementall\
launched bridges there appear to be basicall\ tlvo
methods of’ construction, \\,tiicli we shall call co?/ tirluou.c ctstrng and trnluncd c a s t i n g . .l‘hey are diff‘erent in mode of’ execution and in their areas
of‘ utilization. The continuous casting method is
some\\.hat
analogous to the span-by-span method.
and halanced casting is similar to the cantilever
method.
.l‘ lie continuous casting method is generall! used
for long viaduct-type structures with numerous
equal (or nearI>, equal) spans. Its principal characteristics are the f,llo\ving:
1.
Entire spans, or portions of’ spans, are concreted in fixed forHIS. The f’orms are reused, as
in the span-bv-span method, except that the
f o rm a are fised instead o f m o b ile and are
m o v e d from s p an to s p an . Su b seq u ent sp ans
(or portions of a span) are cast and joined to
the one previously cast, and the superstructure
is progressively launched.
2.
Usually the casting area behind the abutmeIlt
is long enough to accommodate either a span
leng th p lu s lau nc hing - no se leng th o r so m e
multiple of span segment length plus launching-nose length.
3.
Operations involve successive concreting and
launching. The principal phases aI-e: forming;
placing of’reinf’orcing and tendons; concreting
and curing; tensioning and launching.
4.
The two types of’ superstructure cross section
used ha1.e been box girder and double I‘.
5.
Lo ng itud inal p restressing c o nsists o f’ tw o
f hm ilies o f tend o ns: tend o ns concentricall\
placed and tensioned before launching, and
tend o ns p lac ed and tensio ned af’ter launch-
Rio
Caroni,
Venezuela
323
ing-that is, negative-moment tendons over
the supports and positive-moment tendons
in the bottom of the section in the central portion of the span.
The balanced casting method is used for smaller
projects up to a total length of 650 ft (200 m). It is
used for symmetric three-span structures where
the central span is tw ice the end span. Its principal
characteristics are:
1.
Concreting of segments is accomplished symmetrically with respect to a temporary support
located in the embankment behind the abutments. This method is similar to the balanced
cantilever except that the forms are supported
on the embankment fill.
2. Two areas of casting are required, one behind
each abutment. The half-superstructures are
constructed at opposite ends of the project.
The distance between the abutment and the
axis of the temporary massive support is generally slightly less than one-fourth the length
of the project.
3. After the two half-superstructures have been
concreted on the access fill, the two halves are
launched over the piers and joined at midspan
of the central span by a closure pour, which
usually has a length of 3 ft (1 m).
4. Longitudinal prestressing consists of three
families: cantilever tendons for each segment,
located in the upper portion of the cross section and stressed before launching; continuity
tendons, tensioned after closure and situated
in the lower flange; and provisional tendons,
located in the lower flange, tensioned before
launching, and opposing the cantilever tendons.
There are tw o m etho d s o f lau nc hing . The
method used on the Rio Caroni Bridge, Figure
1.67, has the jacks bearing on an abutment face
and pulling on a steel rod, which is attached by
launching shoes to the last segment cast. The second, and more current, method is essentially a
lift-and-push operation using a combination of
horizontal and vertical ja c ks, Figure 7.1. The vertical ja c ks slide on teflon and stainless steel plates.
Friction elements a t the top of the ja c ks engage
the superstructure. The v ertic al ja c k s lif t the
superstructure approximately & in. (5 mm) for
launching. The horizontal jacks then move the
superstructure longitudinally. After the superstructure has been pushed the length of the hor-
FIGURE 7.1. Incremental launching-jacking mechanism (courtesy of Prof. Fritz Leonhardt).
izontal ja c k stroke, the vertical ja c ks are lowered and the horizontal jacks retracted to restart
the cycle.’ Figure 7.2 is a schematic depiction of
this cycle.
To allow the superstructure to move forward,
special temporary sliding bearings of reinforced
rubber pads coated with teflon, which slide on
chrome-nickel steel plates, are provided at the
permanent piers and temporary bents, Figures 7.3
and 7.4. A sequence of operations showing the
bearing-pad movement on the temporary bearing
is depicted in Figure 7.5. A temporary bearing with
a lateral guide bearing is shown in Figure 7.6.
7.2 Rio Caroni, Venezuela
The design for this structure was proposed by consulting engineers Dr. Fritz Leonhardt and Willi
Baur of the firm Leonhardt and Andra, Stuttgart,
West Germany, in an international competition.
Design and planning occurred in 1961 and construction in 1962 and 1963. This structure, Figure
7.7, consists of a two-lane bridge with end spans of
157.5 ft (48 m) and four interior spans of 3 15 ft (96
m), for a to ta l length of 1575 ft (480 m).’ The site
provided some formidable construction problems.
The Rio Caroni River during flood stage reaches a
depth of 40 ft (12 m) with velocities of 13 to 16
ft/ sec (4 to 5 m/ set), thus eliminating the consideration of a cast-in-place concrete superstructure on
falsework. Balanced cantilever segmental construction was considered; however, the interruptions during high-water periods would require an
extensive construction period with attendant high
co sts.3
The proposed method consisted of assembling
and prestressing the entire length of bridge on
324
Incrementally Launched Bridges
(a)zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
FIGURE 7.3, Inc rem ental launching-longituclinal
section of launching bearing, from reference 3 (courtes,
of the American Concrete Institute).
fb)
FIGURE 7.4. Launching
BI idge, Indiana.
Cd)
FIGURE 7.2. Schematic of’ launching jack operation.
(cr)‘Lit‘t. (h) Push. (c) Lower. (cl) Retract.
land adjacent to the bridge site, using precast segments, and launching in a longitudinal direction,
over the piers, into final position. Temporary piers
were used at midspan of’ each interior span to produce ten equal spans of 157.5 ft (48 m) during the
launching of the superstructure. Accommodation
of’ on-site assembly of’ the total superstructure required a 1600 f’t (500 m) long f’abrication bed to the
rear of one abutment, which was partly excavated
in rock and had to be backfilled and compacted
upon completion of’the project. At the f’ar end of
bearing, I+‘abash
River
this fabrication bed stationary steel forms were installed to cast the precast box segments, which were
18 ft 4 in. (5.6 m) high and cast in 30 ft (9.2 m)
lengths.
After the precast segments attained sufhcient
strength thev were stripped from the f’orm and positioned in the fabrication bed to correspond with
their location in the final structure. The segments
were moved f’rom the form on wooden rails accurately positioned in the assembly bed, employing
formica sheets and a petroleutn-base lubricant
between the bottom of the segment and the top of
the wood rails, Figure 7.8. A space of 1 ft 4 in. (40
cm) w x lefi between the precast segments f’or an in
situ joint. Accurate positioning of’ the segments in
the assembly bed was required before casting of
the joints. To avoid shrinkage damage, the joints
were cast during the second half’ of’ the night so
that the temperature expansion of the precast
segments during the heat of the day wc~ulcl compensate for the shrinkage in the cast-in-place joint.”
After the joints were cast, concentric prestressing located inside the box and passing through
openings in the web stif’f’ening ribs, Figure 7.9, was
prestressed with a force of’ 5000 tons in one opera-
Rio Caroni, Venezuela
325
iI v - 5’ +
FIGURE 7.7. Completed Rio Caroni Bridge, from
reference 3 (courtesy of the American Concrete Institute).
FIG U RE 7.8. Precast
seg m ents in assembly
(courtesy of Arvicl Grant).
FIGURE 7.5.
l‘emporarv
sliding bearing, sequence of
operations.
I;tut~chitlg-tetll~)ot~~l~~
bearing and lateral guide bearing (courtesy of Prof. Fritz
Leonhardt).
FIG U RE 7.6. I~~c~xmrc~r~al
I)ed
tion. The prestress tendons were continuous
around a large half-round concrete block at one
end of the structure, Figure 7 .1 0 . This block
reacted against a number o fja c ks and a 10 ft (3 m)
thick concrete bulkhead wall. Bv activating the
jacks between the block and the bulkhead and
causing a movement of 9 ft (2.X m) in the stress
block, the initial prestress force was induced into
the tendons. The prestressing tendons were not
attached to the webs. To reduce the hazard of an\
accidental elastic instability condition, temporarv
steel bracing frames were installed at 60 ft (20 m)
intervals.” The 33 ft 10 in. (10.3 m) top flange of
the box girder section was transversely prestressed,
Figure 7.9.
Upon completion of the prestressing operations
the superstructure was ready for launching over
the temporary and permanent piers to its final POsition. To maintain acceptable levels of concrete
stresses, as the girder was launched over the 157.5
ft (48 m) spans, a 56 ft (17 m) tapered structural
steel launching nose w as attached to the leading
Incrementally Launched Bridges
326
-’
/r9’-l0’16’-51-y.---uI-9~-10’~
FIGURE 7.9. Rio Caroni, girder cross section, fl-om reference 3 (courtesy
of the American Concrete Institute).
FIGURE 7.10. Kio Ckror~i, patressing
tesy of Awiti Grant).
Mock (wur-
end of the superstructure, Figure 7.11. Two double jacks with a total capacity of 600 tons, mounted
against the bridge abutment and pulling on steel
rods fastened to the girder, provided the horizontal force required for the longitudinal launching
movement. To accommodate movement over the
piers, two sliding bearings were provided at each
temporary and permanent pier top. These bearings conststed
of chrome, polished steel plates
which supported teflon covered bridge bearings
w hich w ere placed in an inverted position such that
they bore against the underside of the girder and
slid on the steel plates. After a launching movement of 3 ft (96 cm) in the longitudinal direction
FIGURE 7.11. Rio Carom. Ltu1~111ng nose, 11 OIII refe~ence 3 ((ourtesy of the American Concrete Institute).
the operation was halted to allow the entire
superstructure to be jacked vertically, simultaneously at all piers. The teflon plates were then
moved back to their original position (the one they
occupied when the launching operation started)
and rotated 180 degrees, with respect to a vertical
axis, to compensate for any one-directional movement of the teflon coating. Longitudinal launching
movement occurred at a rate of 24 in./ min (6 cm/
min); thus, one 3 ft (6 cm) increment of movement
took 16 minutes. A total cycle of operation, after
subsequent synchronization, w hich included the
simultaneous jacking at 22 locations and repositioning of 22 teflon bearings, required 30 minutes
Val Restel Viaduct, Ztaly
for each 3 ft (96 cm) of movement. In this manner,
a daily movement of 63 ft (19.2 m) could be accompiished. The required initial jacking force for
launching was 220 tons; this gradually increased to
400 tons f’or the total girder weight of 10,000 tons,
which indicates a friction of 2 to 47c.3
After the launching operation was completed,
the initial concentric prestressing tendon profile
was changed to accommodate the loading condition in the superstructure after temporary piers
were removed. To accomplish the change in tendon profile, special L-shaped rods were installed
so that the! p ro jec ted u p w ard thro u g h the to p
flange or downward through the bottom flange,
the tendons being cradled in the U rods. The rods
ivere then jacked simultaneously at 24 points up~\a~-ct or downward, depending on their location.
During this operation the half-round stress block,
Figure i. 10, ~vas gradually released such that upon
final positioning of the tendons it had retracted 8 ft
6 in. (2.6 III). After the tendons had been relocated,
they lvere attached to the \veb and concreted for
corrosion protection.”
The procedure used for the construction of the
Rio Caroni Bridge, although technically adequate,
is prohibitively expensive. The methodology has
since been refined such that segments are cast directly behind the abutment in lengths of 33 to 100
ft (10 to 30 m) and incrementally launched after
curing of the last segment cast.’
7.3
Val Rested Viaduct, Italy
Because of rugged mountain terrain the alignment
of’ a 1050 ft (320 III) portion of this viaduct re-
327
quired a sharp horizontal curvature of 492 ft (150
m) radius, and a vertical curvature of approxim ately 8860 f t ( 2700 m ) rad iu s, Fig u re 7.12.
Maximum pier height is 212 ft (64.61 m). Site conditions and alignment precluded construction by
the balanced cantilever method or conventional
cast-in-place on falsework, leading to the decision
to construct by the incremental launching method.
The curved 1050 ft (320 m) length of this viaduct consists of 52.5 ft (16 m) long segments, which
were fabricated in an enclosed shed behind an
abutment. The bottom flange and bottom stubs of
the webs of the first segments were cast first, Figure
7.13~1, 6, in a 52.5 ft (16 m) length, and approximately 118 ft (36 m) behind the first abutment.
After curing and stressing of the partial segment it
was jacked forward an increment of 52.5 ft (16 m)
toward the abutment, where the balance of the
section was cast, Figure 7.13~2, c. At the same time
the formw ork vacated by the first-segment bottom
flange was reused for the casting of the bottom
flange of the second segment, monolithically with
the previous segment. A f ter lau nc hing ano ther
52.5 ft (16 m) increment the cycle was repeated
until the superstructure was completed.4
Placement of the bottom flange mild steel reinforcement is shown in Figure 7.14, with the web
forms in the background. The side forms for the
webs and underside of the top flange cantilever,
and the hydraulic jacking arrangement for stripping, are illustrated in Figure 7.15. Reinforcement
in the top flange is shown in Figure 7.16 and the
completed top flange with the following segment in
the b ac kg ro u nd in Fig u re 7.17. The c o m p leted
segment with rails in place as it emerges from the
casting shed is shown in Figure 1.6 1.zyxwvutsrqponmlkjihgfedcbaZYXWVUTS
e=t
t
320.00m
t
Elevation
FIGURE 7.12. Plan (n) and longitudinal profile (6) of the Val Restel Viaduct, showing:
.-\, shed for the construction of the deck segments; B, hydraulic equipment used for
Iannching.
From reference 4.
(b)
Cc)
FIGURE
7.13.
Construe
zyxwvuts
Cd)
tion stages Val Rested \&duct, from 1 eference 4.
FIGURE 7.14. Val Restel, placement of bottom flange
reinforcement, from reference 4.
FIGURE 7.16. Val Restel, top flange reinforcement.
from reference 4.
FIGURE 7.15. Val Restel, side form stripping mechanism, from reference 4.
FIGURE 7.17. Val Restel, completed top flange. with
reinforcement for next segment in background, from
reference 4.
328
Ravensbosch Valley Bridge, Holland
The superstructure cross section is shown in
Figure 7.18~. Width of the segment is 29.5 ft (9.0
m). Total depth ofsegment is 8.13 ft (2.48 m), for a
depth-to-span ratio of l/13. The top flange has a
thickness of 9.8 in. (250 mm) and the bottom
flange a thickness of 5.9 in. (150 mm). Figure 7.186
is a longitudinal section of the superstructure
showing a layout of the second-stage prestressing
tendons required after launching to accommodate
loads on the final structure. Figures 7.19 and 7.20
show the interior anchorage blocks for the
second-stage prestressing before and after concreting,
respectively.
A complete cycle of fabricating and launching a
52.5 ft (16 m) segment was accomplished in four
nine-hour working days. Actual launching time for
one segment was 60 to 65 minutes.4 Figures 7.21
and 7.22 show the launching nose approaching
and landing on a pier. Views of the completed
structure are shown in Figures 7.23 and T.24. Construction of this bridge was accomplished in ten
months, from Januarv 1972 through October
1972.
329
of 6 ft (1.8 m) by 19 ft (5.8 m) with wall thickness of
1.3 ft (0.4 m), Figure 7.26.
The superstructure consists of two siigle-cell
trapezoidal box girders connected at the interior
upper flange tips by a 8.3 ft (2.5 m) slab and prestressed transversely, Figures 7.26 and 7.27. Each
box has a width of 56.8 ft (17.32 m) and a constant
depth of 10.8 ft (3.3 m) for a depth-to-span ratio of
l/17. The top flange has a thickness of 9.8 in. (250
mm) and the bottom flange a thickness of 7.9 in.
(200 mm). Top flange cantilever is 13 ft (4.01 m).
Each dual structure consists of 22 segments approximately 62 ft 4 in. (19 m) in length. The con-
7.4 Ravensbosch Valley Bridge, Holland
The 1378 ft (420 m) long Ravensbosch Valley
Bridge near Valkenburg represents the first bridge
in Holland built by the incremental launching
method of’ segmental construction, Figure 7.25.
‘This dual structure has end spans of 137.8 ft (42
m) and six interior spans of 183.73 ft (56 m). Hollow rectangular piers vary in height from 21 ft (6.5
m) to 77 ft (23.5 m) and have exterior dimensions
FIGURE 7.19.
Val
Restel, second-stage prestressing
anchorage block before concreting, from reference 4.
FIGURE
7.20. Val Restel, second-stage prestressing
anchorage block after concreting, from reference 4.
:I tt40
32.00111
ICavi 4207mm-Cab/es
4207mm
FIGURE 7.18.
t
32.00m
Ccrvo 16 17mm - Cob/e /617mm
(b)
Val Restel. (a) Cross section of deck. (b) Longitudinal section of deck.
From reference 4.
t
Incrementally Launched Bridges
FIGURE 7.21. VA Kc\td, launching IIOW
ing pier, from reference 4.
C~l~l~~
o,~h-
FIGURE 7.24. Val Kesrel, ~on~plctctl
reference 4.
\‘iaclur.c, 1’ 1 o rn
FIGURE 7.22. VA Kestel, launching nose landing on
pier, from reference 4.
V alley IS1itlge, g eneral
FIGURE 7.25. Ka\;ensbosch
view (courtesy of Brice Bender, BVNiS’TS).
F I G U R E 7 . 2 3 . Vitl Kestcl, \,ic\c o f incrcnlcntally
launched curved viaduct after launching, from reference 4.
struction of’ the superstructure was based upon a
cycle of one segment per week.
To ac c o m m o d ate b end ing m o m ents d u ring
launching operations a 52.5 fi (16 m) long launching nose was used, Figure 7.28, in conjunction with
a concentric first-stage prestressing consisting of’26
12 in. (32 mm) diameter Dywidag bars per box
girder. In addition, temporary piers were used at
midspan, Figure 7.28. During launching, friction
amounted to 2 to 4%, equivalent to a maximum
pushing force of 430 tons for a completed box
girder.
Olifant’s
River Bridge, South Africa
331
FIGURE 7.26. Ravensbosch Valley Bridge, dual structure cross section (courtesy of
Brice Bender, BVNISTS).
After completion of the launching, second-stage
prestressing following a parabolic profile and consisting of 12-0.62 in. (16 mm) diameter strands was
installed and stressed. This structure was completed in 1975.
7.5
FIGURE 7.27. Ravensbosch Valley Bridge, girder
cross section (courtesy of Brice Bender, BVNISTS).
FIGURE 7.28. Ravensbosch Valley Bridge, view of
launching nose (courtesy of Brice Bender, BVNLSTS).
Olifant’s River Bridge, South Africa
This railroad structure, upon completion, held the
world’s record for the longest bridge accomplished
by incremental launching. It has a total length of
3395 ft (1035 m), consisting of 23 equal spans of
147.6 ft (45 m). The final structural arrangement
consists of 11 continuous spans on each side fixed
at the abutment and one simply supported center
span-that is, an expansion joint on either side of
the center span. With this structural arrangement
the braking force of the trains (transporting iron
ore) is transmitted to the abutments (10% of live
load). In this manner the flexible piers can be used,
resulting in an economy in the foundations by
comparison w ith the classical solution, w here the
longitudinal force is transmitted through the piers
to the foundations.
A ll 23 spans w ere incrementally launched as 23
continuous spans from one abutment, Figure 7.29.
During launching the two expansion joints were
made temporarily continuous by temporary prestressing. The joints were released after the structure was in place and before it was rested on its
permanent bearings. A launching nose, 59 ft (18
m) long, was prestressed to the first segment to
maintain the cantilever stresses, during launching,
in the concrete within allowable limits. The tip of
Incrementally Launched Bridges
332
END BENT
PO
/
Pl
Construction of the superstructure \vas accomp lished in nine m o nths. Seg m ents lvere span
length, with the theoretical cycle per span of ten
hours attained in the tenth operation and grx!ually reduced to seven hours at the conclusion of’
casting operations. Reinforcing cages Ivet-e pref’Ahricated in templates at the side of. the tornis. A
qcle of operations consisted of the follo~ving:
Clealling
and adjustment of forms
Plac em ent o f reinf o rc ing
loiver flange and \+.ebs
and
t e n d o n s for t h e
Concreting of this first phase
4
Placement of‘
upper flange
reinforcing
and
t e n d o n s t’or t h e
Concreting of’ this phase
FIGURE 7.29. OMant’s
launching awangenwnt.
Rive]- Brid g e. inc rem ental
Tensioning of’tendons in second phase of’ pre\
ious
span cast
the launching nose had a -jacking arrangement to
accbmmodate d eflec tio n o f’ the no se as it ap proached the pier.
In cross section, Figure 7.30, the superstructure
is a c o nstant-d ep th rec tang u lar sing le-c ell b o x
girder. Depth is 12.5 f‘t (3.80 tn); the top flange
is 18 f‘t (5.50 m) wide and the bottom flange 10 f’t
(3.10 m) wide. The webs and flanges are of a constant thic kness thro u g ho u t the stru c tu re. W eb
thickness is 13.75 in. (0.35 m) and contains vertical bar prestressing tendons to carry shear. Longitudinal prestressing is straight and contained in
the flanges. Anchorage blocks for the longitudinal
tendons at-e continuous across the width of both
flanges (interior buttresses) to assure a more favorable distribution throughout the section. There
are no diaphragms at the piers; the interior corner fillets are such as to permit the ef’f’ect of’ torsio n to b e ac c o m m o d ated b y a transv erse b o x
frame.
r
$
/ ......... ‘:. “., -JL
____41
5.50
-I
Safety platform
.:
s
l-4
Lm
I””
L
3.10
FIGURE 7.30. Olifant’s River
I
4
Bridge, cross section
-1‘ensioning of’ tendons in first phase of span in
forms
Stripping of forttis
Launching
Af‘ter launching, and before placing the structure
its final bearings, it \vas necessary to adjust the
joints lvithin 2 itt. (10 mm). l‘he principal &f.ticuties in accomplishing this operation lvere:
on
‘l‘emperature
differential between night and da!.,
w hich p ro d uced a \ariation in leng th o f ’ the
superstructure of 9.X in. (250 mm)
Age of‘ concrete at time of‘ adjustment, lvhich
ied t‘rom nine months to ten hours
V;II‘-
Ja c k i n g op e1‘ ; 1t ’IOIIS, w hic h c o u ld no t retrac t the
structure in case of an error in pushing forward
.I‘he solution of the temperature problem ~~1s to
q uic kly ac c o m p lish the ad justm ent early in the
morning. Because of’ the constant temperature
during the night the temperature of the superstructure was known, and its length was determ inab le in- sp ite o f the therm al inertia o f the
concrete.
The su p erstru c tu re w a s then j a c k e d into its
theo retic al p o sitio n on the ab u tm ent and firnil\,
maintained by a system of blockage. The temporary
tendons that had fixed the first joint were released
and jacks were placed into the joint to push the
remaining 12 spans and place the central simple
span in its exact position. The second joint was
then opened, and jacks at the other abutment po-
Various Bridges in France
sitioned the last 1 l-span portion of the superstructure.
M’hcn rhc superstructure had thus been placed
in position, it was -jacked up off the piers, and the
temporar\ sliding bearings were replaced bv rhe
petmatiet~l bearings.
7.6 Various Bridges in France
7.6.1 I.1.C
I~I.-lDl’C7
‘I‘his is a dual structure 912 f’t (278 in) long on a
curve of a 3280 ft (1000 m) radius. The superstructure \vas constructed by incretnental launching of’ complete spans on sliding bearings. Resistance of rite structure to its dead load during
l a u n c h i n g \vas ~iccotntiiod~tretl
b!- a temporal-!
cable-stay s! h,tetn in which the tension \vas adjusted
a s c.otistt~uctioti
proceeded, Figure 7.3 1. No
supplementary prestressing \vas provided during
the taunching phases. A 26 ft (8 m) launching nose
leas pro\Gied at the leading end in order to reduce
the Jveight of’ the cantilevered structure.
It is a continuous structure supported on neoprene bearings and has a double-T cross section,
as indicated in Figure 7.32. Roadway width is 46 ft
(14.0 m), and depth of superstructure is a constant
10.3 f’t (3.15 in). Spans at-e 133.5 ft (40.i in).
‘l‘his structure consists of eight continuous s p a n s
having a total length of 1102 ft (336 m), crossing a
railroad and the Oise Ri\-er. The project is of interest in that it \vas launched from both abuttnents
\vit bout the use of a launching nose or a tetnporark
cable-sta\svstem. However, tetnporarv bents were
used to control the cantilever stresses. In cross section the superstructure is a single-cell bos, Figure
7 .3 3
S e g m e n t s f’or e a c h o f t h e t w o h a l f - s u p e r
structures were from 65.6 to 98.4 ft (20 to 30 m)
in length. .A launching \vas effected upon completion of’ each segment. After the two halfsuperstructures had been launched to their final
position, a closure pour of 3.3 ft (1 m) in length was
c.onsutiitiiated to provide continuit\..
333
Concentric tendons frotn one end to the other of
each half-superstructure, coupled together at each
phase of concreting of segments
Straight, short tendons in the top flange over the
piers and in the bottom flange, centered in the
span and tensioned after launching
Continuity tendons, tensioned af’ter launching,
situated in webs and anchoring at the upper flange
Short parabolic tendons, located in the webs and
a n c h o r i n g i n t h e t o p flange, t e n s i o n e d a f t e r
launching
‘retnporary tendons in the upper flange, having
the satne effect as the cantilever rendons
i.h.3
0I.I
L’I‘-tDI’CT
I‘his viaduct spans the valley of Oli in 15 spans of
134.5 ft (41 m) for a rotal length of 2017 ft (615 m)
at a height of 197 ft (60 tn). The structure has
a grade of 5.355% and a horizontal curve lvith a
radius of 6700 ft (2046 tn). Total weight of the
superstructure is 16,500 tons (15,000 mt).
Incremental launching in this structure, rather
than pushing the superstructure out over the piers,
was accomplished bv a restrained lowering down
the grade. The fo;-ce required in braking the
structure was approximately 660 tons (600 tnt) as
compared to the estimated force of 1540 tons
(1400 tnt) to push the structure uphill.
In its final configuration, because it was difficult
to accommodate horizontal forces due to braking
and seismic effects in the tall flexible piers, the
superstructure is anchored in the terrain in the
area of the abuttnents by a tie of a large stiffness.
All of this longitudinal global force is accomtnodared in the large stiff tie, the abutments, and the
relativelv short stiff piers in each bank. A central
joint diiides the structure into two independent
srructures.
Upon cotnpletion of launching and before placing the superstructure on its pertnanent bearings,
it was necessary to “unlock” the joint that held the
two half-superstructures together during construction and to adjust its position within approximatelv i in. (10 mm). This operation was conducted as follows:
Longitudinal prestress consists of six sets:
Cantilever tendons, tensioned bef-ore launching,
located in the top flange and anchored in fillets at
the intersection with the web
The superstructure was restrained at the upper
abuttnent until the distance between its theoretical
position and the end of the lower abutment was
approximately 8 in. (200 mm).
334
Incrementally Launched Bridges
.6 .
r.
zyxwvutsrq
F+,,.
1n
c;:--..
placing of the launching nose
concreting and prestressing of the first span
launching of the first span
concreting and prestressing of the second s?an
erecting the cable-stay system
launching of the first two spans
concreting and prestressing of the third span
launching of the first three spans
concreting and prestressing of the fourth span
launching of the first four spans
thing operations
disassembling of the launching nose and
cab1 e-s Lay systm
placing on permenant bearings
placing and tensioning of phase 2 prestressing
FIGURE 7.31. Luc \‘iaduct. incremental launching phases. ((I) Placing of the launching
IIOSC. concreting atd prestressing of the fit-st span. launching of the first span. (b) (:OIIcreting and prestressing of the second span. erecting of the cable-sta! s\ stem, launching
of the first t\vo spans. (c) Concreting and PI estressing of the third span. launching of‘ the
first three spans. (rl) (:oncreting and Prestressing of the fourth span, launching of‘thc firat
four spans. (P) Completion of launching operations. disassembling of’ the launching 11osc
and cable-stay system, placing on permanent bearings, placing md tensioning of’
phase-two prestressing.
.The t e m p o r a r y t e n d o n s c o n n e c t i n g t h e t w o h a l f s u p e r s t r u c t u r e s w e r e successivelv
detensioned.
However, two temporary tendons’ restrained the
l o w e r h a l f - s u p e r s t r u c t u r e . T h e u p p e r half-superstructure was fixed to the upper abutment by a
s y s t e m o f ‘ p r e s t r e s s b a r s a n d c.otn~,lementar\
reinforcement installed in the upper abutment.
.l‘he t w o temporat-!- t e n d o n s r e s t r a i n i n g t h e lowet
half-superstt-uctttre
lvere d e t e n s i o n e d i n increments, allowing the lolver half-super-strttcrure
to
Wabash River Bridge, U.S.A.
t-
14.65
I
1
FIGURE 7.32. Luc Viaduct, cross section.
FIGURE 7.33. Creil \‘iaduct.
cross section.
descend to a blocking system in the lower abutment.
Fixing o f the lo w er half - su p erstru c tu re to the
lower abutment was then accomplished.
The sup erstruc ture w as p o sitio ned o n its final
bearings.
7.7 Wabash River Bridge, U.S.A.
‘I-his structure, the first incrementally launched
segmental bridge constructed in the United States,
c arries tw o lanes o f U.S. 136 o v er the W ab ash
River near Covington, Indiana. It is a six-span
structure with end spans of 93 ft 6 in. (28.5 m) and
four interior spans of 18i ft (57 m), Figure 7.34.
Roadway width is 44 ft (13.4 m). Pier heights are
approximately 40 ft (12 m); average river depth is
11 ft (3.35 m) with low water at 8 ft (2.4 m) and
high water at 24 ft (7.3 m). The superstructure is a
t\\‘o-cell box girder with a constant depth of 8 ft
(2.4 m). .I‘he prqject was awarded in September of
1976 lvith a completion date of October 1978. The
entire superstructure was completed in November
of 1977.
O rig inal d esig n p lans p rep ared by A m eric an
Consulting Engineers, Inc., of Indianapolis for the
State Highway Commission called for a precast
segmental balanced cantilever design; however,
the bid documents permitted alternative methods
of constructing the superstructure. The successful
contractor, a .joint venture of Weddle Bros. Con-
335
struction Co., Inc., and the Ralph Rodgers Construction Co., both of Bloomington, Indiana, investigated three alternatives for the superstructure
co nstructio n. These alternates includ ed cast-inplace segments supported on falsework, incremental launching, and the cast-in-place segmental
balanced cantilever method. Incremental launching w as the suc c essful m etho d and reportedl)
saved $100,000 over the other precast segmental
method:j The V .S.L. C o rp o ratio n o f Las G ato s,
California, was the subcontractor for prestressing
and launching.
A 140 ft (42.7 m) casting bed was located behind
the west abutment of the bridge and could acconmodate three 46 ft 9 in. (14.25 m) segments. .I‘he
forms for casting were supported on I beams,
which were supported on steel piling to provide a
solid foundation and prevent any settlement of the
casting bed, Figure 7.35. The bottom third ot. the
two-cell box superstructure was cast at the most
westerly end of the casting bed, Figure 7.35. It was
then advanced 46 ft 9 in. (14.25 In), w here forms
for the balance of the section were positioned, mild
tendons
steel reinf o rc em ent and prestressing
placed, and the balance of the segment cast, Figure
7.36. A f ter the seg m ent had b een p o u red and
cured, the 20-ton jacks that held the forms in position, Figure 7.37, were released to break the bond
and rem o v e the f o rm s. ‘I-he larg e m etal f o rm s
stayed in place and were simplv swung in and out
as needed. The segment was then advanced to the
f o rw ard third o f the c asting b ed f o r su rf ac e
finishing by a conventional Bidwell screed, Figure
7.38, before launching over the abutment. In this
methodolog!
W;lS
m a n n e r a production-line
maintained. Three segments were always in various stages of fabrication, with reinforc&,lent
and
p restressing tend o ns c o ntinu o u s b etw een seg ments.
53
The first-stage pour required approsimately
vd3 (40.5 m”) and the second pour required from
iO1 to 130 yd3 (77.2 to 99.4 111”). It took approximately four hours for each pour. ‘I‘jventy-eight-da)
design strength \\‘as 4800 psi (3.37 kg/ mm”),
and
6000 to 7000 psi concrete strengths were actualI>
attained (4.2 to 4.9 kg/ mm’). A 3500 psi (2.46 kg/
mm2) strength was required before stressing, and
this was normally achieved in 24 to 30 hours. As
segments were completed, each was stressed to its
predecessor by first-stage prestressing consisting of
eight tendons of twelve f in. (12.7 mm) diameter 27
ksi (190 kg/ mm2)
strands, Figure 7.39. Initially the
contractor was able to complete one cycle of segm ent fab ric atio n and lau nc hing in tw o w eeks:
A T
MIDSPAN
AT PIERS
PLAN
D IREC TIO N O F MO VEMW
Fa b ric a tio n
w
a re a
CON>mJON
-.
ELEVA.
FIGURE 7.34. Wabash River Bridge: cross section of girder, from reference 6; construction details, from reference 2.
however, as experience was gained, two cycles per
week were attained.
To accommodate the launching stresses a 56 ft
(17 m) launching nose was attached to the lead
segment, Figures 7.34 and 7.40. In addition, the
four interior spans had temporary steel bents at
midspan,
Figures 7.34 and 7.41. In this manner
the total structure length was divided into ten
equal spans of 93 ft 6 in. (28.5 m) during the
launching procedure.
Because of the longitudinal force on the piers
during launching, the permanent piers were tied
back to the abutment with four prestressing
strands each. These strands were stressed to 96
kips (43,545 kg) before launching commenced.
Each temporary pier was tied back to the preced-
FIGURE
support.
7.35.
Wabash
River
Bridge,
casting-bed
FIGURE 7.39. Wabash River Bridge, first-stage prestressing.
FIGURE 7.37.
Wabash River Bridge, side form jacks.
FIGURE 7.38.
top flange.
Wabash River Bridge, surface finishing
incrementally
FIGURE
7.40.
Wabash River Bridge, launching nose.
Launched
Bridges
horizontal jacks an 18 in. (457 mm) stroke. The
vertical jacks lifted the superstructure about 4 in.
(13 mm) and the horizontal jack pushed it forward
17 in. (432 mm). Each jacking cycle required about
five minutes, and the entire launching of a 46 ft 9
in. (14.25 m) segment required about three hours.
Temporary bearings, Figure 7.4, were located at
each temporary bent and permanent pier. During
the launching operation workmen were stationed
at each bearing location to insert the teflon pads as
the superstructure slid over the bearings. To
maintain lateral alignment of the superstr&ture,
lateral guide bearings, Figure 7.43, were also located at each temporary bearing and also used
teflon pads. Workmen would tighten bolts on one
side of the superstructure and loosen them on the
opposite side to push the superstructure laterally.
Final positioning of the superstructure on the east
abutment was within & in. (0.8 mm) of its prescribed location.
7.8 Other Notable Structures
7.8.1 MiiHLBACHTALBRiiCKE,
GERMANY
ing permanent pier by two stays of 10 in. by 10 in.
(254 mm by 254 mm) structural steel tubing, Figures 7.34 and 7.42.
The jacking procedure during launching used
the two-jack system (one vertical and one horizontal) and teflon pads, as described in Figure 7.2. The
vertical jacks had a 2 in. (50 mm) stroke and the
Another example of this type of construction is the
Miihlbachtalbriicke
about 30 miles (50 km) southwest of Stuttgart, West Germany, Figure 7.44. This
structure has an overall length of 1903 ft (580 m)
with 141 ft (43 m) spans. The far-side trapezoidal
box girder is shown in Figure 7.44 completed from
abutment to abutment; the near-side trapezoidal
box girder has been launched from the left abutment and the launching nose has reached the first
pier. A general view of the structure is presented in
Figure 7.45.
FIGURE 7.42. Wabash River Bridge, structural steel
tubing tie.
F I G U R E 7 . 4 3 . Walmsh
bearing .
FIGURE 7.41. W,lb,1sh Rile1
BI Age, temporary steel
bent.
Ki\;er Br-idge,
lareral g u i d e
FIGURE 7.44. zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
Miihlbachtalbrticke. aerial view
FIGURE 7.47. Miihlbachtalbriicke,
first-stage
pre-
stressing tendon anchorage.
FIGURE 7.45. Miihlbachtalbriicke, general view.
Some idea of the size of the box girder may be
obtained from Figure 7.46, showing the interior of
the formwork at the rear of the abutment. Firststage prestressing tendon anchorage at the top of
the web may be seen in Figure 7.47. The anchorage block for the second-stage prestressing is located inside the completed box, Figure 7.48.
FIGURE 7.48. Miihlbachtalbriicke, second-stage prestressing anchorage block.
7.8.2 SHEPHERDS HOUSE BRIDGE. ENGLAND
FIGURE 7.46. Miihlbachtalbriicke, segment in statio nary fo rms.
The Shepherds House Bridge is the first incrementally launched bridge constructed in England.
This highway structure crosses four railroad tracks
at Sonning Cutting, near Reading, about 30 miles
(48 km) west of London. The new structure contrasts sharply with an existing brick arch structure
built in 1838 by Brunel, a famous English engineer. The existing structure consists of three circular brick arches supported on tall brick piers
with the abutments founded in the sides of the
cutting.’ A general plan showing the existing
bridge, railroad tracks, and alignment of the new
structure is presented in Figure 7.49.s
340
Incrementally Launched Bridges
FIGURE 7.49. Shepherds House Bridge, general plan, from reference 8 (courtesy of
Institution of Civil Engineers).
In 1971 the no rth ab u tm ent settled and the
existing bridge was temporarily closed for repairs.
In March of 1972, because the life expectancy of
the existing structure was in question and because
it did not comply with current highway standards,
the Ministry of Transport instructed consulting
engineers, Bullen and Partners, to prepare a study
to determine the type and method of construction
for a new structure. The new bridge provides a
dualing of the existing road, and in the future the
existing bridge will be replaced by a parallel structure.
Because British Rail was engaged in extensive
maintenance and upgrading of the tracks prior to
introduction of high-speed trains, there would be
severe limitations on track possession. Further, it
was dictated that piers between tracks were to be
avoided and that f-oundations on the north slope of
the cutting were not to disturb the foundations of
the existing bridge abutment. Construction working area was restricted because traffic was to be
maintained on a residential street at one end and a
trunk road at the other end. Soil conditions required that any temporary conditions that would
load or disturb the slopes was to be avoided, thus
requiring pile foundations with the pile caps at the
surface to avoid extensive excavation in the slopes.s
The consultants initially studied five possible
schemes for construction of a bridge. Schemes
using cast-in-place construction on falsework had
earlier been rejected.
An incremental launching scheme was recommended, even though there were no accurate cost
data for construction in the U.K. The consultants
concluded that this scheme, although of shorter
length than customary for this type of construction, would solve the problems of restricted working space and interference with residential streets
and would require the least track downtime.
The west elevation of the bridge is shown in Figure 7.50. Span lengths, determined by track location, are 75.5 ft (23 In), 121.4 ft (37 m), and X2 ft
(25 III). The bridge is fixed at the south abutment
with an expansion joint at the north abutment. ‘The
casting bed for the production of 31.5 ft (9.6 rn)
segments was located to the rear of the south
abutment. The south abutment was located to provide maximum work space for the casting bed and
to clear a large number of Post Office communication cables. Interior piers b and c were designed to
w ithstand the fric tio n fo rc es exerted d u ring
launching operations. In addition, pier c, located
close to the railroad tracks, was subject to damage
or complete demolishment in the event of a derailment. Therefore, the superstructure was designed to withstand the removal of pier c by an accid ent. Six untensio ned but ancho red M acallo v
tendons in certain segments were added so as to
preclude ultimate collapse with no live load on the
bridge and pier c removed.7*H
Normally, in this type of construction, the casting bed is of sufficient length to accommodate at
least two and sometimes three segment lengths,
such that the bottom flange may be cast separately
in advance of the webs and top flange. In this project, with restricted space for the casting bed, it was
decided to cast one complete segment in one pour.
341
Other Notable Structures
- a ohllk rmll
‘we,. h n,ns 00.
Dbnwula h mWumb*.
Shepherds House Bridge, west elevation, from reference 7 (courtesy of
The Concrete Society, London).
FIGURE 7.50.
A maximum of three weeks was allowed for construction and launching of a segment. This time
was later reduced to two weeks except for those
segments with a diaphragm.’ A typical cross section
of the box girder segment is shown in Figure 7.51.
The launching sequence is shown in Figure 7.52.
The steel launching truss nose was first erected
using a temporary intermediate support. The first
segment was cast against the launching nose and
post-tensioned by Macalloy bars, some of which
were used to connect the launching nose to the first
segment. The launching nose, in position, before
the launching of the first segment is shown in Figure 7.53. After the first segment had been
launched forward, the next segment was cast and
post-tensioned to the previous one. This procedure was repeated until the completed bridge was
FIGURE
launched to the north abutment. The launching
nose passing over pier c is shown in Figure 7.54.
Arrival of the launching nose at pier b is shown in
Figure 7.55. The launching nose was removed
after the concrete superstructure arrived at pier b,
Figure 7.56.
The superstructure was launched over temporary bearings, which consisted of high-grade concrete pads with a +Z in. (1 mm) thick stainless steel
plate clamped and tensioned across the top surface. Lateral guide bearings were also provided to
keep the superstructure on line. Upon completion
of launching the superstructure was jacked in a
predetermined s e q u e n c e a n d t h e t e m p o r a r y
bearings were replaced with permanent bearings8
The jacking force for launching was provided by
two jacks pulling on a set of nine 0.6 in. (15 mm)
7.51. Shepherds House Bridge, girder cross section, from refer.
ence 8 (courtesy of The Institution of Civil Engineers).
Incrementally Launched Bridges
342
Stage 1: Cast first unit and
connect to launching
nose
Stage 2: Launch to pier C
Stages 3-5: Launch over tracks
FIGURE 7.54. Shepherds House Bridge, launching
nose passing over pier c, from reference 7 (courtesy of
The Concrete Societv. London).
Stage 6: Launch to per B
Stage 7: Conttnue launch
Stage 8: Reach pw 9 and remove “018
Stages 9 and 10: Complete launch
FIGURE 7.52. Shepherds House Bridge, sequence of
incremental launching, from reference 8 (courtesy of
The Institution of Civil Engineers).
FIGURE 7.55. Shepherds House Bridge, launching
nose at pier b, from reference 7 (courtesy of The Concrete Society, London).
FIGURE 7.53. Shepherds House Bridge, launching
nose in position before launching, from reference 7
(courtesy of The Concrete Society, London).
diameter cables passing under the casting bed and
anchored to the front of the abutment. The load
was applied to a fabricated bracket secured to the
rear of the segment by bolts coupling with the
projecting ends of the M acalloy bar tendons in
the top and bottom flanges of the segment, Figure 7.57. The two jacks were operated in tandem
by a single pump. This system required 30 seconds
for jacking and 30 seconds for retracting for each
10 in. (254 mm) str0ke.s
FIGURE 7.56. Shepherds House Bridge, superstructure launched to pier b and launching nose removed,
from reference 7 (courtesy of The Concrete Society,
London).
Design
of
Incrementally
Launched
Bridges
The dimensions for
sented in Section 4.5.4
thickness, but the top
thickness may have to
the type of prestressing
7.9.4).
7.9.3
FIGURE 7.57. Shqhtwls 1 lowc~ Brid g e, segment
being launched from f’ormwork,
from reference 7
(courtesy of’ The Concrete Society, London).
7.9
Design of Incrementally Launched Bridges
7.9.1 BRIDGE ALIGNMENT REQUIREMENTS
The designer must always remember that in order
to construct incrementally launched bridges, the
horizontal and vertical alignment must be either
straight or constantly curved or twisted. This is
generally not the case, as road planners are not
bridge builders. As a matter of fact, it is the soffit ofthe bridge deck that has to be designed with a constant radius of curvature; the transverse cantilever
of the deck flange can be varied to accommodate
possible small deviations.
7.9.2
343
typical cross sections preremain valid for the web
flange and bottom flange
be increased, depending on
layout adopted (see Section
SPAN ARRANGEME,VT
AND RELATED
PRI,VCIPLE
OF CONSTRUCTION
The constant-depth requirement limits the economical use of this construction method to spans
not longer than 160 to 200 ft (50 to 60 m). It is
advantageous if all the spans are equal in length.
However, much longer spans have been built by
utilizing special techniques in conjunction with the
basic principle of incremental launching.
A three-span construction may be launched
from both sides. In this way the center span can be
twice the length of the edge spans without increase
of the stresses in the deck. The span configuration
then becomes: L-2L-L (see Figure 7.58).
Champigny Bridge near Paris was the first structure of this type. Longer bridges are often
launched from one side only (the record length is
that of Olifant’s River Bridge in South Africa, in
excess of 3300 ft). Auxiliary temporary devices are
used to reduce the bending moments in the front
portion of the deck (launching nose or tower stays)
TYPE, SHAPE AND DIM ENSIONS OF
SUPERSTRUCTURE
This method of construction requires a cross section with a constant depth, since the designer has
to insure the resistance of the superstructure,
under its own weight, at all sections as the launching proceeds. Economic considerations dictate a
constant moment of inertia.
Two types of cross section have been used to
date: the box girder and the double T. The box
girder provides a better stiffness and resistance to
torsion and at the same time an easier placement of
the prestressing tendons in the cross section. The
depth of the box is usually one-twelfth to onesixteenth of the longest span, the first value applying to larger and the second to smaller spans.
Table 7.1 summarizes the characteristics of several
incrementally launched bridges.
FIGURE 7.58. Three-span symmetrical incrementally
launched bridge.
TABLE 7.1. Characteristics of Incrementally Launched Bridges
1‘) pical
Name
Year
Cross Section
Span
(t’t)
Nuel Viaduct,
France
~I‘otal
Length
(W
Launched
Weight (t)
Vertical
Curve
Horizontal
Curve
1976
135
807
6,000 Slope 6%
R = 2,460 ft
Borriglione
Viaduct,
France
1976
135
807
6,000 Slope 5.5%
R = 2,460 ft
Kimonkro
Bridge,
Ivory Coast
1978
118
709
3,600
Straight
Tet Viaduct,
France
141
660
Luc Viaduct,
France
135
915
7,900 Slope 3.8%
Straight
1976
135
1,151
Slope 1.3%
Curve
1976
135
2,018
1972
131
345
1978
194
1,102
Paillon
Bridge,
France
Oli Viaduct,
France
Marolles
Bridge,
France
Creil Bridge,
France
344
15,000 Slope 5.85% R =
6,712 ft
zyxwvut
Design of Incrementally Launched Bridges
TABLE 7.1. (Continued)
l‘otal
.\ ‘I I 1 1 C
Gro nachtal
Brid ge,
Gerlnan!
Length
(it)
l’eal-
Launched
Weight (t)
345
Vertical
Curve
Horizontal
Curve
4 6 .3 ’
1978
t-
1
262
1,732
13,000 Slope 0.7%
R = 7,217 ft
\
L17.5’
Var Viaduct,
France
1976
138
1,107
Bridge,
Kufstein,
German\
1965
335
1,476
169
1,562
138
1,398
1967
108
469
1978
6 spans
9,700
Straight
Inn
\
Koches Valle)
Bridge,
Gerlnan\
Querlin Guen
Bridge.
German\
Abeou
Aqueduct,
France
Ingolstadt
Bridge,
Danube
Brid ge,
Gertnant
197 to
377
as previously indicated in some of the examples described in this chapter.
When the spans become too large, intermediate
temporary bents are used. This was done for the
first bridge over the Caroni River in Venezuela.
The reco rd sp an leng th fo r inc rem entally
launched bridges was obtained by a structure over
the Danube River designed by Prof. Leonhardt,
the originator of the method, Figure 7.59. The cost
of the temporary bents depends greatly ox the
2x
1,246
foundation conditions; it may be prohibitive if the
bent height is greater than 100 ft (30 m) and soil
conditions require deep piling.
For very long bridges, intermediate expansion
joints are needed, much the same as for cantilever
bridges. The expansion joints are temporarily
fixed by prestressing during launching and are released at the end of construction to allow for thermal expansion in the structure during service. A
very ingenious variation of this principle was de-
346
Incrementally Launched Bridges
FIGURE 7.59. Ih~ltrlx
Ki\-e t.
Brid g e ,
.- \ ustria .
veloped for the Basra Bridge in Iraq, where a concrete swing span was launched together with the
approach spans as a single unit and later arranged
to serve its purpose as a movable bridge over the
navigation channel, Figure 7.60.
7.9.4
DESIGN OF LOlVGITUDINAL
M EM BERS FOR
FLEXURE AND TENDON PROFILE
During launching, the superstructure is subjected
to continually alternating bending moments, so
that any one section is subjected to a continual
variation of bending moments, both positive and
negative, as shown in Figures 7.61 and 7.62. These
bending moments are balanced by internal uniform axial prestressing.
In the final stage, additional tendons are required to supplement the uniform axial prestressing in order to carry the service loads. Conventional solutions are applied to this problem, and in
the present discussion we need only enlarge upon
the specific problem of the axial prestressing. For
this prestressing, tendons are so arranged that the
compressive stresses are the same over the entire
cross-sectional area. The required tendons are
placed in the top and bottom flanges of the box
section. They are usually straight, tensioned before
launching, so couplers are needed at each joint
between successive segments.
Segment length may vary from 50 ft (15 m) to
100 ft (30 m). As noted in our discussion of- the
progressive construction method, there are limitations to the deck’s capacity to carry its own weight
during launching when the front part is in cantilever beyond a typical pier. To keep bending
moments and stresses within allowable values, it is
usually necessary to use a launching nose, a light
steel member placed in front of the concrete
structure to allow support from the next pier,
rather than launching the concrete deck all the way
with no support. Numerical values are given in
Figures 7.61 and 7.62 for the critical maximum
positive and negative moments during launching.
Assuming the unit weight of the launching nose
to be 10% of the weight of the concrete deck (a
value somewhat lower than average), the critical
7.61. Critical negative moments during launching with nose. M,
(W’L2/12)[6a’ + 6y(l - &)I. Multiplier: WL2/12. For y = 0.10:
FIGURE
N
P
M”
0.20
0.30
0.40
0.50
0.80
0.70
0.60
0.50
0.82
1.09
1.46
1.95
1 .oo
0.00
6.00
k yo/rp,n
=
.&
7.62. Critical positive moment during launching with nose. M, =
(WL”/12)(0.933 - 2.96#*). Multiplier = WL’l12. For y = 0.10:
a
M,
P
FIGURE
0.20
0.30
0.40
0.50
0.80
0.50
0.60
0.50
0.74
0.79
0.83
0.86
1.00
0.00
0.93
347
3
4
8
Incrementally Launched Bridges
moments are as follows for various lengths of the
launching nose:
Nose Length,
Percent of
Typical Span
50
60
70
80
Maximum Moments
Support OKJ
Span (Ml)
M&f,
1.95
1.46
1.09
0.82
0.86
0.83
0.79
0.74
2.27
1.76
1.38
1.11
Moment factor is WL2/12
(W = weight of concrete per unit length and L =
span length)
FIGURE 7.65. Sathorn
Technologically, the uniform axial prestress may
be installed in the superstructure in several different w ays:
above the concrete deck with steel deviation saddles at intermediate joints.
The three solutions above have their relative
merits and disadvantages:
1.
2.
Straight tendons running through the top and
bottom flange of each segment, joined by
couplers at the joints between segments.
Straight tendons running through the top and
bottom flanges, anchored in block-outs inside
the box girder, Figure 7.63.
3. Temporary curved tendons may be used to
balance the final continuity tendons during
construction. These tendons are outside the
concrete section between supports, Figure
7.64. This method has been used for several
large projects.
Figure 7.65 shows the Sathorn Bridge in Bangkok,
Thailand, with the temporary tendons installed
FIGURE 7.63. Lapped prestressing tendons.
TEMPORARY
I
PRESTRESSING
SUPPORT5
F I N A L PRES’RESSING
I
FIGURE 7.64. Temporary external prestressing system.
Hr idge, Thailand.
1. The first solution may require local thickening of the concrete flanges for placement of
the couplers. However, it is often preferred to increase the thickness of the flanges over the entire
bridge length to simplify casting of the segments.
Axial prestressing tendons are permanent and
cannot be removed. They must be incorporated in
the final prestressing layout. The joints between
segments have to be carefullv designed, owing to
the presence of couplers and concrete voids that
may significantly weaken the section.
2. The main advantage of the second solution
pertains to the removal and reuse of those tendons
not required in the final prestressing layout. However, the cost and difficulty of providing a large
number of block-outs offsets a significant part of
the advantage of removing the temporary tendons.
In order to obtain a satisfactory shear resistance
from the webs, particularly during launching with
alternating shear and b end ing stresses, the
configuration of the box section and location of the
upper and lower blisters must be carefully considered. This problem was mentioned in Chapter 4 as
presenting potential difficulties. A satisfactory solution is shown in Figure 7.66, where upper and
lower blisters are not in the same vertical plane. A
sufficient amount of vertical prestress will insure
the resistance of w ebs against shear during all construction stages.
3. The third solution is theoretically a satisfactory one, allowing the permanent prestress to be
installed during construction and the temporary
prestress to be designed only to counteract the un-
349
Design of Incrementally Launched Bridges
reinforcement must be made in an area already
densely prestressed.
FIGURE 7.66. Offset lapped prestressing tendons.
desired effects of the former during ‘moment reversals created by the successive launching stages.
In practice, installation of the tendons passing
from the inside to the outside of the box section is
not particularly simple. An attempt should be
made to reuse these temporary tendons to reduce
the investment in nonproductive materials.
A comparative analysis between the first two
methods of temporary prestressing has been made
for a typical railway bridge. Solution 2 requires
19% more conventional reinforcement than solution 1 because of the many blisters and more elaborate tendon layout. The total cost of materials
(concrete prestress and reinforcement) is 9%
higher for solution 2 than for solution 1. These results may be significantly different for highway
bridges, where the ratio between girder load and
superimposed dead and live loads is very different.
7.9.5 CASTING AREA AND
LAC’,VCHING METHODS
The precasting area is located behind one abutment and has a length usually equal to that of two
or three segments. T h e r e a r e t w o d i f f e r e n t
launching methods:
1. The launching force is transmitted from the
ja c ks bearing against the abutment face to the
bridge by pulling tendons or steel rods anchored in the bridge soffit.
2. A launching device consisting of horizontal
and vertical ja c ks is placed over the abutment.
The vertical ja c k rests on a sliding surface and
has a special friction gripping element at the
top. The vertical jack lifts the superstructure
for launching, and the horizontal jack pushes it
horizontallv.
The designer should be concerned with the following items:
The first launching method applies high local
forces to the concrete soffit where the pulling device is anchored. Careful design of the passive
The second launching method requires sufficient
vertical reaction on the vertical ja c k. This could be
critical at the end of launching, when the required
launching force reaches its maximum with a corresponding small vertical reaction.
A very precise geometry control is required during
launching. The possibility of foundation settlement must be considered in the design. Whichever
launching method is used, after completion of the
launching procedure the deck must be raised successively at each pier so that the permanent bearings may be installed. This phase also calls for careful analysis.
7.9.6 LAUNCHING NOSE AND
TEMPORARY STAYS
The large cantilever moments occurring in the
front part of the superstructure that is being
launched from pier to pier inevitably call for special provisions to keep the bending stresses and the
temporary prestress within allowable and economically acceptable limits. Two methods have
been used together and separately, as previously
mentioned:
nose: A steel member made either of
plate girders or of trusses is temporarily prestressed into the end diaphragm of the concrete
bridge, which is the front section of the deck during launching.
Tower and stays: This method was described in
Chapter 6 for progressive construction. Its application to incremental launching, however, needs a
special approach, because the relative position of
the tower and the stays changes constantly with regard to the permanent piers.
Launching
The advantage of the launching nose to reduce
cantilever moments in the concrete superstructure
was discussed in Section 7.9.4. It is important not
only to select the proper dimensions of the
launching nose but also to ta ke into proper account
the actual flexibility of the steel nose in comparison
to that of the concrete span. This relative flexibility
may be characterized by the following dimensionless coefficient:
++
c c
Incrementally Launched Bridges
350
where E, and E, refer to steel and concrete moduli,
and I, and I, are the moments of inertia of the steel
nose and concrete superstructure. Figure 7.67 presents the results of a study analyzing the variation
of the maximum support moment in the concrete
deck for different launching stages with the relative stiffness K. This chart confirms the obvious
fact that a flexible nose has only a limited efficiency
in reducing the moments in the concrete deck. The
following table gives the characteristics of several
structures using a launching nose and serves as a
reference for preliminary investigations of the optimum launching method.
Bridge
Launching
Nose Length
[ft (Ml
Wabash River
Oli River
Saone
Roche
56
59
93.5
124.5
(17)
(18)
(28.5)
(38)
Weight of
Launching
Nose (tons)
Stays
30
36
65
90
No
Yes
No
No
To allow the method to be effective in all
launching stages, it is necessary to constantly control the reaction of the tower applied to the concrete deck. When the tower is above one pier, it is
totally efficient. When launching has proceeded
for another half-span length, the tower and stays
produce additional positive moments at midspan,
exactly contrary to the desired effect. For this reason the tower may be equipped with jacks between
the concrete deck and the tower legs, and the tower
reaction may be constantly adjusted to optimize the
stresses in the concrete superstructure. Figure 7.68
shows a device being successfully used for the first
time in the construction of the Boivre Viaduct,
near Poitiers, France.
7.9.7 PIERS AND FOUNDATIONS
The loads applied to the piers and foundations
during the incremental launching procedure are
very different from those appearing during service. The static configuration of the piers is also
For longer spans the launching nose is not necessarily the optimum solution, while temporary bents
may also be expensive. A tower-and-stay system
has been successfully used either alone or in conjunction w ith a launching nose to reduce the cantilever moments in the front part of the superstructure.
FIGURE 7.67. Variation of the maximum support
moment.
FIGURE 7.68. Boivre Viaduct. nwr I’oiliers. France.
351
Design of Incrementally Launched Bridges
different. During construction, the bridge slides
over the pier tops and the buckling length of the
pier is larger than that during service. The horizontal force applied to the pier top is also higher
than during service, thus requiring a close study of
this construction phase.
Lou& <4cting on the Piers The various systems of
horizontal forces that may act on the piers depend
on the following:
Longitudinal
Direction of’
profile
of
the
superstructure
launching
E‘riction coef‘ficient of sliding bearings
Notation:
H= angle ot bridge superstructure lvith respect to
the horizontal; tan 0 = r
4= angle of’ f’riction of sliding bearings; tan C$ = p
R = to tal reac tio n o f the su p erstru c tu re o n the
piel-: \,ertical and ho riz o ntal c o m p o nents V
and H, normal and tangential components A:
an d 7
The f’ollowing four cases will be considered (see
Figure 7.69):
1. H > 4, upulard launching: Sliding starts on the
bearings \vhen the inclination of the reaction R
\\,ith respect to the vertical is:
cY=t)++,
FIGURE 7.69. Reactions on piers during launching.
(a) upward launching. (b) downward launching.
F=N(tan8-tan+)
o
r
F=N(r-p)
For the same reasons as above, the safe value
of F is equal to Nr.
3. 0 < $, upward launching: As above, the horizontal load applied to the pier is:
H = V tan (0 + 4)
H = (r + p)V
For small values of 0 and 4:
H = (r + p)V
2. H > 4, downward launching: Sliding starts
Lvhen cy = 8 - 4. The horizontal force on the
pier acts in the direction opposite to that of
movement irith a value:
H = V tan (0 - 4)
For small values of the angles:
H = (I - p)V
Because p varies with environmental conditions (cleanness of the plates in particular), the
launching equipment and the pier will be designed for H = 4’ . The downward movement
of’ the bridge is controlled by a restraining
jacking force:
4. 0 < 4, downward launching: In this case the
horizontal load on the pier is applied in the direction of the movement with a value of‘:
H = (r - p)V
Because of the possible variation in the angle
of friction, it is safer to provide a braking system to control the movement of the bridge.
Pier Cap Detailing The pier caps must be carefully detailed in order to provide room for the following devices:
Temporary
sliding
bearings
Vertical ja c ks to lift the bridge after launching to
install the permanent bearings
Horizontal guiding
devices during
launching
352
Incrementally Launched Bridges
Adjusting jacks for correction of the relative displacements between piers and deck
and the principle of the method are shown in Figure 7.70. The 900-ton structure had a width of 26
ft and the following spans: 46, 55, 55, 46 ft.
The existing reinforcing did not provide the necessary strength to resist superstructure dead load
during launching. Therefore, a rear launching-out
tail 26 ft long was installed at the end opposite the
direction of launching, while exterior post-tensioning tendons were placed above the deck to
strengthen the structure.
The bridge was lifted off its bearings 7 in. to install sliding bearings and lateral guiding devices in
preparation f’or the operation. The whole operation was performed in 54 weeks as f’ollo\vs:
Moreover, to reduce the pier bending moments
induced by launching, the sliding bearings are
often eccentric. However, it is possible to reduce or
balance this horizontal force by installing ties anchored in the ground. If the piers are very high,
the horizontal force can be eliminated by using
jacking equipment directly installed on the piers.
7.10
DemoEition of a Structure by
Incremental Launching
We close this chapter with an unusual application
showing the interesting potential of incremental
launching. An overpass structure over the A-l
motorwav north of Paris needed to be demolished
for replacement by another structure as part of a
highway relocation program. I‘he limited headroonl between the existing bridge soffit and the
clearance diagram, together with the considerable
traffic on the major motorway providing permanent access from Paris to Charles de Gaulle Airport, made all conventional methods of demolition
extremely difficult and unadapted.
A ver! simple scheme was devised whereby the
deck was launched away from the traffic onto the
approach embankment to be conventionallv demolished at leisure. The dimensions of the bridge
Design and preparation of the contract
Mobilization and purchase of equipment
Launching
2
2
1;
5;
rraffic \vas interrupted for only f’our nights between 10 P.SI. and 6 .a.~. The operation turned out
to be a complete success in spite of its originality.
L4UNWlt4G
I
46’
1
55’
1
55’
:
T
I
46’
t
T O T A L WEICUT
soot
PROCEDURE
I/ L I F T T O T A L
2)
PLACE
SCHEDULE
BRIOSC 7’
ROLLERS OV ER
PI ERS
A N D 48lJlMCNl5
3) INSTALL
APPROACH FILL AND
CONCRETE
T O T A L 5’/2 W E E K S
_ DESIGN & CONTRACT :
2
.MOB.
2
PURCUAsLS :
_ LAUNCHING
:
BEAM5
TRAFFIC INTERRUPTION :
A)
PLACE
A
PROVISIONAL
REAR NOW
zyxw
IV!?
5k
P/l A N D
26 FT. LONG
FIGURE 7.70. Bridge over A-1,
(IO PM.
launching out.
TO
6
A.M .)
4
N;OUTs
References
2. :~non..
“First Incrementally Launched Post-Tensioned Box Girder Bridge to Be Built in the United
Swtes.” Bridge Report, December 1976, Post-Tensioning Institute, Phoenix, Ariz.
3. .ir\ icl (;I.d1it. “Increment;il
Launching of Concrete
Stn~cturcs.” Jorrmnl of thr .-lvwriccl~r Courwtp Imtltutr,
\‘()I. T”. s o . 8.
4 .
.-\ug”st
19i5.
(:c.IllcIlto
.-\llllatc,
E:
ssoc-id/ione It~rliamr
(.\I IX(:). Rome 1971.
.\
.-h~Il., “Segmental Box Girder Bridges \lake the Big
‘I‘ime in U.S..” Engiuuwiug .\‘~!\-RPcoM/. \Iarch 2.
1978.
6. Xnon., “Wabash Rive]- Bridge. Covingtot). I ndian;l.”
P o r t l a n d C e m e n t .-\ssociation, Bridge Repot“,
SR201 .Ol E, lYi8, Skokie. 111.
i. Xl. .\Iaddison, “Crossing the Cutting with Segments at
Sonning,” Coucwtp, 7%~ Jou,-r~u/ of tha Corlo.rtr Socirt!
(Lot~rlor~j, Yol. 12. S o . 2, Februar\ 19iH.
8. K. H. Best, R. H. Kingston, and 11. J. \Vhatle\, “lncremental Launching at Shepherd House Bridge,”
Pwcfwfirrgc, Instztution of Cnfil Eugrnrfm, \‘ol. 64. Part
I, Fehruar\ 1978.
5 .
zyxwvutsrq
“\‘a1 Restel Viaduct tol- the Provincial Road
S o . 89 .Se;rr Ko\ereto. .I‘rento,” Prr.clrr.c,wd Couovtr
.Strrlttrrtfs I)/ I/n/y 1970/l 974, .-\ssociarioIle Itdli;~na
.4IlOII..
353
l’l~ec.oIllpl~esM ~
~cononica
(A IC.-\I’)
and
Del Cement0
zy
8
Concrete Segmental Arches,
Rigid Frames, and Truss Bridges
8.1
8.2
8.3
8.4
8.5
I N T R O D U CT I O N
S E G M E N T AL P R E CAS T B R I D G E S O V E R T H E
M AR N E R I V E R , F R AN CE
CAR ACAS V I AD U CT S , V E N E Z U E LA
G LAD E S V I LLE B R I D G E , AU S T R ALI A
AR CH E S B U I LT I N CAN T I LE V E R
8 . 6 R I G I D - F R AM E B R I D G E S
Saint Michel
Bridge in Toulouse, France
8.6.1
8 . 6 . 2 B r i e s l e Maas B r i d g e , N e t h e r l a n d s
8 . 6 . 3 B o n h o mme B r i d g e , F r a n c e
8.6.4
M o t o rway O ve rpasse s i n t he M i ddl e E ast
8 . 7 TR US S BR I D GES
8.7.1
8.5.1
R e vi e w o f Co nc e pt ; S umma ry o f S t ruc t ure s w i t h
T e mp o r a r y S t a y s
8 . 5 . 2 N e c k a r b ur g B r i d g e , G e r ma n y
8 . 5 . 3 N i e s e n b a c h B r i d g e , Aus t r i a
8 . 5 . 4 K i r k B r i d g e s , Y ug o s l a vi a
8.1 Introduction
An arch bridge, in a proper setting, is an elegant
and graref‘ul structure with aesthetic appeal. Instinctivelv, a layman relates to an arch bridge as a
form that follows its function. Long before prestressed concrete was developed as a technology,
concrete arches were used for long spans, taking
advantage ot the compressive stress induced b\
gravitational- forces into a curved tnetnbet- much as
earlier
generations o f b uild ers had d o ne w ith
niasotirv arches.
Three b rid g es d esig ned and b u ilt b y Eu g ene
Frey ssinet b etw een 1907 and 1910 in c entral
France were to become a tnajor landmark in the
development of concrete structures. In the \‘eut-d re Brid g e, Fig u re 8.1, the three hing ed reinforced concrete arches had a clear span of 238 ft
(72.50 m) and an unusual rise-to-span ratio of l/ 15
dictated by the topography of the site and the sudden floods of the Allier River. The \‘enture Fvas an
unqualified success both during load testing and
after opening to traffic. As Freyssinet wrote in his
memoirs:
354
R e t ro s pe c t o n Co nc e pt s f o r Co nc re t e Trus s
B ri dg e s
8 . 7 . 2 Mangfall B r i d g e , Aus t r i a
8 . 7 . 3 R i p B ri dg e , Austral i a
8.7.4
Co nc e pt f o r a Cro s s i ng o f t he E ng l i s h Channe l
R E F E R E N CE S
Lord testing um (I triumph. 0~ the Gght bank, (1 hill
oz~erlookirrg the bricige site UYS occupied by .\e-c~ewl
thou,wd spectators do had trrken thei?- plclce c~lretrd~ crt
dnulrl to ulatch the j<lilure of‘ the bricl<ge predicted by n
10~~1 riea$mper .cold to some ur~happ~ competitor. These
hopes were deceiz~ed, c~rrd ule had (I -corltirruom lnrle oj
henry .cteclm rollers trm~eling the bridge brick CIH~ for-th
quite unable to produce ar+hing more th(crl the corn&ted
e(n.rtic de$ections.
Betw een 1907 and 191 1, ho w ev er. f ears d eveloped in Freyssinet’s mind. It seemed that the
hand rails, which had been properly aligned at the
time of the load test, were showing some convexit!
toward the skv at the nodes of the cro\vn hinges. Br
the spring of 191 1 the crown had moved do~vnw a r d a s m u c h a s 5 i n . ( 0 . 1 3 m), a n d cot-t-espondingly the springings had raised appreciablv.
W itho u t telling any o ne, Frey ssinet mobilized a
team of four devoted tnen and placed hydraulic
ratns at the arch crowns to raise the bridge spans to
their original profile; he then replaced the hinge
by a rigid concrete connection between the trvo
abutting
half-arches. This near-d isaster \v;ts the
355
Introduction
FIGURE 8.1. Veurdre
first consequence seen in a structure of a phenomenon theretofore completely ignored: long-term
concrete creep.
Other beautiful concrete arches were also constructed in the sam e p erio d . The V illeneuv e
Bridge over the Lot River in southwestern France,
Figure 8.2, is an interesting example. The twin
arch ribs are of plain concrete with a clear span of
316 ft (96 m) and a rise of 47 ft 4 in. (14.5 m). Each
rib has a solid section 10 ft (3 m) w ide and 4 ft 9 in.
(1.45 m) deep built in at both ends into the concrete abutments. The reinforced concrete deck
rests upon the arch ribs through a series of thin
spandrel columns, fa c e d with red brick.
Construction began shortly before World War I
and was interrupted for four years, fortunately not
before the concrete arch ribs could be cast on a
wooden falsework, Figure 8.3. Immediately upon
completion, hydraulic rams were used at the
midspan section to lift the concrete arches off the
falsework and actively create the compressive stress
in them, a technique from Freyssinet’s fertile mind
that already contained the germ of the idea of prestressing.
FIGURE 8.2.
\~illc nc u\c
HI
itigc O\~I chc Lot Ki\cr.
Bridge.
The bridge was completed in 1919 and kept the
world’s record for long-span concrete structures
for several years. The photograph appearing in
Figure 8.2 wa s ta ke n by one of the authors in the
summer of 1980; it shows that beautiful structure
in a remarkable state after sixtvI vears of continuous operation under constant urban traffic.
Another Freyssinet design, the Tonneins Bridge
over the Garonne River, was built at the same time,
and he considered it to be one of his nicest bridge
structures, Figure 8.4.
The Plougastel Bridge in Brittany, Figure 1.38,
reached for longer spans with concrete arches. For
the first time a box section was employed, calling
on an ingenious method of construction in which a
wooden falsework was floated into position and reused several times for the various arch ribs. Dimensions of the structure and typical details of the
arches are show n in Figure 8.5, w hich is a facsimile
of a document published in 1930.
The three arches have a span length of 611 ft
(186.40 m) and carry a single-track railroad and a
two-lane highway. The reinforced concrete trussed
double de c k accommodates the train track on its
lower level and the highway on the upper. Near
the arch crow n in each span, the train passes
through the arch rib.
The arch ribs were only slightly reinforced and
the quantity of steel was 39 lb/ y&’ (23 kg/ m’{), in
spite of the relatively thin walls used for the box
section.
The three arch ribs were constructed one after
the other on a temporary wooden arch built on
shore and floated into position for each of the
three concrete arches, Figures 8.6 ,and 8.7. This
wooden arch was 490 ft (150 m) long and weighed
550 tons (500 mt), including the two reinforced
concrete end sections, which allowed the thrust
created by the concrete arch ribs to be transferred
.
--
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,I,.
-r
Segmental Precast Bridges Over the Marne River, France
357
know it today. It incorporated so many innovations
in a single structure that it would not be out of
place in today’s modern bridge technology.
The single-span structure, Figure 8.8, is a
double-hinged arch with a distance between hinges
of 180 ft (55 m) and a very tight clearance diagram
for river navigation that allowed only 4 ft 3 in.
(1.30 m) below the finished grade of the roadway.
Consequently, not only is the bridge structure very
shallow, 4.16 ft (1.27 m), at midspan, but the riseto-span ratio of the arch is unusual: l/23. The
bridge consists of three parallel box sections made
up of precast segments 8 ft (2.44 m) long, connected after placement in the structure by precast
slab sections at both top and bottom flanges, Figure
8.9.
The bridge is prestressed in three directions:
to the arch springings completed earlier on the
foundation
caissons.
Two barges and a temporary steel tie slightly
above the water level, with the help of the large
tidal range, allowed the transfer of this falsework
from the construction area to the three positions of
use and its final return after completion of the concrete structure.
As this outstanding undertaking neared completion in 1930 after five years of uninterrupted
effort, Freyssinet expressed his thoughts as follows:
In Brittany light is like a fairy who constantly plays at
covering nature with [many] changing coats, now of
lead, noul of silver or of pearls, or of something immaterial and radmnt.
Toward the evening oj the load testing of the bridge, she
had spread her most sumptuous treasures on the roadstead
and each line of the work, changed into a long rosary of
unreal light, added another touch of beauty to the marvellous whole, proving in this way that the Fairy of the
Roadstead had already adopted the child that men had
imposed on her and had known how to weave for him
garments magn$cent enough to hide all the imperfections
of the work.
8.2
Segmental Precast Bridges over the
Marne River, France
Located some 30 miles (50 km) east of Paris, the
Luzancy Bridge represents probably the first application of truly segmental construction as we
The 4 in. (0.10 m) webs are vertically prestressed to
resist shear.
The longitudinal box girders are then prestressed
to connect the precast segments and resist bending.
The negative-moment prestressing tendons at the
top flange level over the arch springings are located in grooves provided at the top surface of the
precast segment upper flange and are ultimately
embedded in a 2 in. (50 mm) concrete topping.
This dense, high-quality concrete pavement provides the sole protection for the high-tensile steel
wires and also serves as the sole roadway wearing
course. In spite of the excellent behavior of this
structure after more than 34 years of operation, it
would probably be difficult to envisage duplicating
it today.
Transverse connection between the box girders
and the connecting slabs is achieved by prestressing.
There was no conventional reinforcing steel in
the bridge superstructure except in local areas,
such as the Freyssinet concrete hinges at the arch
springings. The erection was just as remarkable as
the conception of the bridge. Each box girder consisted of 22 segments, which were cast in a central
yard at the rate of one a day (little progress has
been achieved after thirty years). Afterward they
were carefully aligned on concrete blocks to take
the profile of the finished structure with proper
provision for camber. The 2 in. (20 mm) wide joints
were dry packed to allow segment assembly by prestressing. In fact, the 22 segments of each box
girder were assembled at this stage in three units:
two side units made up of three segments each, and
the center unit incorporating the remaining 16
L E
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CIVIL
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FONT A.LOUPPE,EN BQTONARMG, SUR L'ELORN,PRlj:S DEPLOUGASTEL (FINIST~RE)
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la dqositmn des armatures
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FIGURE 8.5. Plougastel Bridge, dimensions of the structure and details of the arches, a
facsimile of a document published in 1930.
Segmental Precast Bridges Over the Mame River, France
segments with a length of 170 ft (52 m) and a
maximum weight of 134 tons (122 mt). All three
units were assembled on the bridge centerline immediately behind one abutment, while the deltashaped sections representing the arch springings
were cast in place over the abutment in their final
location in the structure.
A special aerial cableway made up of two steel
towers resting on both banks and properly anchored to the rear, a system of suspended winches,
and a unique elliptical drum allowed the transfer
of the precast girder units from their assembly po-
Pou 1 re
Demo-coupe
359
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FIGURE 8.8. Luzancy
midlane - Demo coupe dans he
i l a
cli
Bern/. coupe i 24 “ho de la c/i
FIGURE 8.9. Luzancy Bridge, concrete dimensions.
Blitlgc.
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
360
sition on the banks to their final location in the
structure. In spite of a seemingly involved concept,
the operations were carried out safely and rapidly;
a center beam was placed in only eight hours and a
complete arch including all preparatory and
finishing operations was assembled in 120 hours,
Figure 8.10.
Another interesting feature of this structure was
the incorporation at both arch springings of Freyssinet flat jacks and reinforced concrete wedges
between the arch inclined legs and the abutment
sills, to adjust and control the arch thrust and the
bending moments at midspan.
The bridge was opened to traffic in May 1946
after successfully proving its structural adequacy
through a comprehensive series of static and
dynamic load tests, following a custom still in use
today in several European countries. Figure 8.11
gives a view of the finished structure.
This first precast segmental arch bridge was followed a few years later by a series of five other
structures, all of the same type and in the same
geographical area, the valley of the Marne River,
FIGURE 8.12. One of the five hlarne River Bridges:
Esbly, Anet, Char@, ‘Trilbardou, and Ussy.
Figure 8.12, at the following locations: Esbly, Anet,
Changis, Trilbardou, and Ussy. All five bridges
have the geometric dimensions shown in Figure
8.13:
Distance between hinges: 243 ft (74 m)
Rise of the central axis at the crown over the abutment hinge: 16.3 ft (4.96 m)
Depth at crown: 2.82 ft (0.86 m)
Deck width: 27.5 ft (8.40 m)
The deck structure is made up of six precast girders, each consisting of:
Two precast delta-shaped sections at the
springings
Thirty-two precast segments 6.8 ft (2.07 m) long
and weighing from 2 to 4.2 tons (1.8 to 3.8 mt).
FIGURE 8.10. Luzancy Bridge, erection of central
section.
FIGURE
rise.
8.11.
Luzancy Bridge, view shmving flat arch
The same design and construction principles
used at the Luzancy Bridge were repeated for this
series of five bridges, except for some improvements commensurate to the experience gained
from the first structure and taking into account the
importance of the project. Precasting of the 960
segments was achieved in a factory completely enclosed and using the most modern concrete manufacturing techniques of that period.
Each segment was fabricated in two stages in
heavy steel forms. Top and bottom flanges were
cast first, with high-strength steel stirrups embedded in both units. After strength was achieved,
a set of steel forms equipped with jacks was placed
between the flanges, which were jacked apart to
stress the web pretensioned stirrups. Then the web
was cast between the flanges. There was no need
for any conventional reinforcing steel in the precast segments.
The concrete was vibrated with high-frequency
external vibrators, then compressed for maximum
FIGURE
8.13.
Marne River Bridges, typical longitudinal and CTOSS sections.
362
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
compaction and steam cured for a fast reuse of the
forms. The equivalent 2%day cylinder strength
was in excess of 6500 psi.
Near the precast factory, an assembly yard allowed the segments to be carefully aligned and assembled by temporary prestressing into sections,
which were transferred into barges to be floated to
the various bridge sites. Each longitudinal girder
was thus made up of six sections:
The two delta springing sections
Two
intermediate
Two
center
five-segment
ten-segment
sections
sections
Handling of these various sections was performed
by the Luzancy cableway properly rearranged for
the purpose.
The stability of the side sections, at both ends,
was obtained by temporary cantilever cables anchored in the abutments, while the two center sections were suspended to the cableway until casting
of the wet joints was completed and longitudinal
prestressing installed to allow the arches to support
their own weight. Figures 8.14 through 8.16 show
the various sequences of the arch construction,
while one of the finished bridges is shown in Figure
8.17.
The quantities of materials for the superstructure were very low, considering the span length
and the slenderness of the structure:
FIGURE 8.15.
tral section.
,tIanre
FIGURE 8.16.
tral section.
11~1 nc Kixcl RI idgcs, t‘~ cc tton of (cn-
Kiwx Rridgm, L’IWJ ion ot’ ccn-
Precast concrete: 353 yd:’ (270 m3)
Reinforcing steel: 13.2 tons (12 mt)
Prestressing steel: 13.2 tons (12 mt)
For a deck area of 6540 ft2, the quantities per
square foot were:
Precast concrete: 1.46 ft 3/ft 2
Reinforcing steel: 4.0 lb/ft2
Prestressing steel: 4.0 lb/ft2
FIGURE
tion.
8.14.
YI,II IIC Ki\ (‘1 RI iclga, Ed ectetl end scc-
As in the Luzancy Bridge, the high-density concrete placed over the exposed longitudinal prestressing tendons was also used for the roadway
Caracas
FIGURE 8.17.
Viaducts,
Marne River Bridges, completed strut-
Venezuela
363
ture.
FIGURE 8.19. C;lr-atas Viadutr\,
wearing course. The behavior of these bridges has
been excellent f’or thirty years.
length as shown in Table 8.1 .2 Preliminary investigations indicated that adequate soil material would
probably be found irregularly at great depths.
Construction of abutments to resist large bending
moments under these conditions would be difficult
if not impossible. The decision was therefore made
that the abutments would resist only the centered
thrust of the arches and that the bending moments
applied to the abutment would be reduced, as far
as practical, to zero. This required that hinges be
located as near as possible to the points of origin of
the arches. Because of consideration of long-term
creep deformation on buckling of the arch and
possible consequences of abutment displacement
as might be caused by an earthquake, the decision
was made to eliminate a crown hinge, thus resulting in two hinged arches.’
Although the bridges vary considerably in dimensions, they are quite similar in appearance. Because of the valley profile, it was possible to use the
same basic design for all three structures. All were
designed for AASHO H20-44 loading. Wherever
possible, the elements were standardized in order to
minimize design and maximize precasting and prefabrication.
Pilasters were placed at each end of the arch in
Bridge 1 so as to avoid an unpleasant appearance
of a change without transition from the main
structure to the approach viaducts.
8.3 Caracas Viaducts, Venezuela
In Venezuela in 1952 a highway was being constructed between Caracas and La Guaira airport.
Alignment of this highway necessitated crossing a
gorge at three locations with relatively large
bridges. These structures were designed and constructed under the direction of Eugene Freyssinet.’
Although the three bridges are similar in appearance, Figures 8.18 and 8.19, they vary in
FIGURE 8.18. ~&acas
Viaducts, Bridge 1.
Britlgcs 2 ,mct 3.
TABLE 8.1. Caracas Viaduct Arches
Height from
Bridge
1
2
3
Total Length
Bed of Gorge
1013 ft (308.8 m)
830 i’t (253 m)
700 ft (213.4 m)
230 ft (70.1 m)
240 ft (73.2 m)
170 ft (51.8 m)
Main Span
498 it (151.8 m)
478 it (145.7 Ill)
453 tt (138.1 m)
364
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
FIGURE 8.20. Caracas Viaducts, elevation of Bridge 1, from reference 1 (courtesy of
Civil Engineering-AXE)
An elevation of Bridge 1, Figure 8.20, shows the
principal dimensions and foundations of the arch.
The three bridges have identical cross sections,
Figure 8.2 1. The poured-in-place concrete deck
topping varies in thickness from 2 in. (50 mm) at
the edges to 7$ in. (190 mm) at the center to provide a transverse slope of 1.5% for drainage. Each
deck span, except at the crown, consists of eight
precast prestressed I girders. Variations in span
length of the deck girders are accommodated by
adding or removing standard form units. Identical
transversely prestressed precast stay-in-place deck
slabs span transversely between the deck girders.
Continuity of the deck girders is accomplished by
longitudinal tendons placed in a groove in the top
of the top flange of the girders.’
Approach piers and spandrel columns over the
arc hes c o nsist o f three I- shap ed c o lu m ns o f a
standard cross section shown in Figure 8.21. A
five-segment precast cap beam on the columns re-
Cross section of pier A-A
ceives the eight deck I girders. A perspective of the
deck over the piers is shown in Figure 8.22. The
precast deck girders, cap beams, and slab are supported on the cast-in-place piers, and the whole assembly is prestressed vertically, transversely, and
longitudinally.
The c enter sp an c o nsists o f three p arallel
d o u b le- hing ed arc h rib s 27 f t 6 in. ( 8.4 m ) o n
center, Figure 8.21. Each arch rib is a box with a
width of 10 ft 6 in. (3.2 m) and a slightly varying
depth from 9 ft 6 in. (2.9 m) to 10 ft (3.05 m) at the
su p p o rting p o ints o f the d ec k. To p ro v id e increased capacity to resist end moments developed
by horizontal loads, the width of the ribs is increased to 17 ft (5.18 m) at the spring lines. The 5
in. by 5 in. (127 mm x 127 mm) fillets provided at
each inside corner of the box are to reduce the
concentration of torsion stresses. Thickness of the
b o tto m f lan g e o f th e b o x rib w as ke p t to a
minimum to reduce weight on the falsework. The
FCross
5’3” 4
section of arcqly
’
1
7”
itkA
FIGURE 8.21. Caracas Viaducts, typical cross section, from reference 1 (courtesy
of Civil Engineering-ASCE).
Caracas Viaducts, Venezuela
365
FIGURE 8.22. Caracas Viaducts, perspective of deck over piers, from reference
1 (courtesy of Civil Engineering-ASCE)
thicker top flange provides the box rib with the required area and moment of inertia for resisting
thrust and live-load moments.
Design of these structures considered a design
wind pressure of 50 psf (2.4 kN/m*). The arch ribs
carry part of the wind pressure to which they are
directly subjected; the remainder is transmitted to
the deck structure by bending of the spandrel columns and the connection of the arch rib to the
deck at the crown. The arches were assumed to be
transversely fixed in the foundations, the end moment developed in the springings resulting in a
slight transverse displacement of the pressure
line.*
Thus, the deck structure was chosen as the principal member to resist wind loads, requiring the
exclusion of all joints in the deck from abutment to
abutment. The condition of deck continuity led to
the attachment of the deck to the arch on both
sides of the arch crown. This was accomplished by
prestressing the continuous cables provided over
the top flange of the girders and anchoring them
into the arch. Six girders were connected to the
arch in this manner: the two intermediate girders
that do not rest directly on the arch were
lengthened to the crown, Figure 8.21.*
During construction, an open joint was provided
at the crown. In this joint Freyssinet flat jacks
staggered with concrete wedges were inserted,
acting as a hinge for the arches to adjust the pressure line during different phases of construction.
Expansion and contraction of the deck due to
temperature, creep, and shrinkage take place over
an approximate length of 1000 ft (305 m), developing approximately symmetrically on both
sides of the arch crown. Free movement of the
deck structure over the pilasters was accommodated by providing two concrete rockers over each
transverse wall of a pilaster. The rockers consisted
of a 3 ft 6 in. (1.07 m) high continuous wall
throughout the width of the bridge with a continuous Freyssinet-type concrete hinge at both the top
and bottom. Approach piers were fixed in the deck
at the top and hinged at their footings. Because of
their height, these piers have sufficient flexibility to
allow movement of the deck without developing
appreciable bending moments, the exception being
the short stiff piers next to the abutment, which
were hinged both top and bottom.*
We shall describe the construction procedure for
the superstructure of Bridge 1, which was also
used for the other two bridges. Because the cableway did not have the capacity to transport the deck
girders across the canyon, precasting operations
were established at both ends of the bridge. During
construction of the foundations, precasting operations were started at both sites at either end of the
bridge.
When the foundations for the approach piers
were completed, the cableway transported and positioned the precast Freyssinet pier hinges to their
respective locations, where they were grouted to
their respective foundations. Pouring of the piers
then commenced, using special steel forms attached to the hinge blocks. Two sets of forms were
used in leap-frog fashion to maintain a pouring
366
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
the cantilever method, this formwork being placed
rate of 5 ft (1.52 m) per day. Because of the hinge
by the overhead cableway and held in place by a
at the base of each pier column, the piers required
system of cable stays. Thus, the arch rib was essentemporary support until the deck girders could be
tially constructed to the quarter-points. The center
placed. The first 25 ft (7.62 m) lift of each column
half-span formwork was constructed as a light
in each pier was supported by a light steel scafwooden trussed arch assembled at the bottom of
folding that surrounded each column; the scafthe canyon and winched into position from the
foldings, in turn, were braced together. Succeedends of the quarter-span cantilevers. .I‘he timber
ing 25 ft (7.62 m) lifts were braced to the previous
falsework truss was wedged against the concrete
lift by light timber trusses. As the columns in the
arch ribs already erected. It acted as an arch under
piers rose, steel reinforcement was placed: at the
same time, holes for vertical prestressing tendons
the weight of the bottom flange concrete, transwere cast in the concrete by the insertion of l+ in.
mitting its thrust to the cantilevered arch sections
previously
erected. Later the timber falsework
(38 mm) steel tubes, which were withdrawn lfzyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFE
acted compositely with the hardened bottom flange
lio~~t-s after concrete placenlent.3
concrete to support the webs and top Hange of the
Upon completion of the three columns of an aphollowbox arch ribs when they were placed.’
proach pie r, precast segments of the cap beam
The following discussion describes the erection
were placed atop the columns and prestressed
sequence of the center-span arch ribs.” The first
vertically to them as indicated in Figure 8.22. The
falsework unit in the quarter-span for each arch
two intermediate cap beam segments were placed
rib consisted of a timber platform 31 ft (9.45 m) in
by the cableway and temporaril!; held in position
length with a width of 27 ft 8 in. (8.43 m) at the
by steel brackets. Four prestresslng tendons were
spring line and a width of 17 ft 2 in. (5.23 m) at the
then placed through the cap beam segments and
opposite end, Figure 8.23 (Phase 1). This platform
the four vertical 14 in. (38 mm) joints between the
was constructed of 3 x 10 in. (76.2 x 254 mm) timsegments were packed with a rich mortar. After
bers on edge at l@ in. (267 mm) centers covered
eight to ten hours the longitudinal tendons in the
cap beam \vere stressed and anchored to complete
on the upper face with 1 in. (12.7 mm) thick
plywood. It provided the form for the bottom of
a pier bent, which was then readv to receive the
deck girders and slabs. The 137 ft (41.75 m) high
the arch rib. For the first section of the quarterpilasters at each end of the arch are four-celled
span, three of these units (one for each rib) were
hollow boses 20 by 80 ft (6.1 s 24.4 m) in plan with
placed by the cableway, supported by cable stays A
all walls 4: in. (120.65 mm) thick. They were conand B, and their position adjusted by hydraulic
structed in lifts with special steel forms that were
jacks at the ends of the anchor cable stays. Next
leap-f‘rogged. ‘l‘en vertical prestressing tendons
four precast Freyssinet hinge blocks were posianchored into the foundation provided stabilit\
tioned at the spring line and assembled into one
against wind forces.3
hinge block by prestressing them together. Forms
Upon completion of the abutments and the first
were then erected on the falsework for the webs of
approach piers, erection of the bridge deck girders
the arch rib, and placement of concrete commenced, Figure 8.23 (Phase 2). As the weight of
and slabs commenced. It was accomplished with a
126 ft (38.4 m) long structural steel lattice girder
each increment of concrete came onto the forms,
gantry, 60 f’t (18.3 m) of which extended as a canthe cable stays elongated and the geometry of the
arch-rib soffit had to be carefully adjusted by the
tilever. One 48 ft (14.6 m) span, consisting of eight
precast beams and 112 precast slabs, required nine
hydraulic jacks.
Upon completion of the concreting for the first
working days and a crew of 16 men. When the approach viaduct decks were in place, they were presection of the quarter-span, falsework section 2 was
attached to it and supported by two more cable
stressed longitudinally by prestressing tendons
stays, C and D. After geometry adjustment, conplaced in the grooves of the top flange of the deck
creting continued, Figures 8.23 (Phase 3) and 8.24.
girders, which were anchored at one end into the
As a result of the position of the cable stays and the
abutment and at the other end over the arch pilasconcreting
sequence, angular deformations were
ters.
possible
between
falsework sections 1 and 2.
The three arch ribs of the main span were cast in
Therefore,
a
temporary
concrete hinge was placed
place on a light wooden falsework, which was rein
the
lower
used almost in its entirety for the two other
flange of the arch rib, which would
bridges. Basically, the system adopted was to erect
allow angular deformation but transmit the thrust
the timber formwork for casting the arch ribs by
to maintain equilibrium. When the concreting of
PHASF3
3As.u
/
/’
/
-. -.._c
/
/
/. i’
;rF---_
‘\(
,/
/I
/zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
FIGURE 8.23. Caracas Viaducts, erection and construction sequence, from refet-ence
(courtesy of Civil Engineering-ASCE).
3
367
368
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
FIGURE 8.24. Caxcas Viaducts, comtruction of wch
springings on suspended scaffolding.
the second portion of arch rib was completed and
geometry adjustment made, the temporary hinge
was blocked and the two sections were prestressed
together. In the same manner, temporary hinges
were used for the remaining sections of the
quarter-span arch rib and at each end of the central half-span arch section.
The first two sections of arch rib thus became a
continuous member supported at the outer end by
cable stays, and during construction of the rest of
the arch its geometric position was adjusted by
cable stay D.
The next operation was the erection of the third
falsework unit consisting of a trusswork. Its weight
was such that it could not be accommodated by the
cableway, Therefore, it was assembled at the bottom of the canyon below its position in the arch.
The outer end was lifted by the cableway and the
inner end by a winch located at the end of the previously concreted section of the arch, Stay cables E
passing over the pilaster were attached, and the
bottom flange of the new arch rib section was cast,
Figures 8.23 (Phase 4) and 8.25,
In like manner the next section of trussed
falsework was positioned and supported by cable
stay F. Next, concrete for the bottom flange of the
rib was placed, including small concrete brackets
which protruded below the bottom flange to take
the thrust of the 267 ft (81.4 m) central falsework
after its positioning, Figure 8.23 (Phase 5).
In the last phase of the quarter-span concreting,
the vertical webs were formed and concreted, as
well as a few narrow strips across the top to provide
stiffness to the arch-rib members, which at this
stage had a U-shaped cross section, Figure 8.23
(Phase 6). The anchor stay cables were again adjusted to bring the 125 ft (38 m) quarter-span into
its proper position.
The central 267 ft (81.4 m) falsework span had
been assembled at the bottom of the canyon below
its final position in the arch, Figure 8.26. The ends
of the timber falsework arches were tied together
by steel cables acting as ties to keep the arch
falsework rigid. The whole central falsework was
hoisted into position by winches located at the ends
of the cantilevered quarter-span units, Figure 8.27.
Once the central falsework was in place and the
location of the crown exactly positioned, cement
mortar was packed in the gap between the ends
of the central falsework and the quarter-span
falsework, and extra-flat sand boxes were embedded in the joint for subsequent stripping of the
central falsework.
After two days, the steel tie cables on the central
falsework were released and the winches support-
GENERAL
,/ - -n
DETAIL OF JOlNT OF TOP MEMBER
AND
DIAGONALS
ELEVATION
OF
FALSEWORK
I A’ ._
- \
a,,II\
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
ing the section were loosened. At this point the
combination of the central trussed falsework and
the concreted quarter-span units acted as a complete arch from abutment to abutment.
Next, the bottom flanges of the arch ribs were
concreted, in a previously arranged sequence, up
to the crown on each side, and temporary crown
hinge blocks were placed. The other temporary
hinges between elements of the quarter span were
blocked and the cable stays up to stay D removed.
‘rhe combination of timber falsework and partly
built concrete arch ribs continued to be held in position by stays D, E, and F, with a temporary hinge
at F onlv.
The vertical webs of the arch ribs over the central section were then concreted up to the crown
hinge; cable stav D was released; crown concrete
was completed; the remaining construction joints
were tied with prestressing tendons; and the last
cable stays E and F were released. At this point the
concrete arch ribs, less the top flange over the
center 260 ft (79.25 m) section, carried themselves
as well as the dead lo a d of the entire falsework.
Next, the cement joints at the ends of the
falsework were destroyed, sand boxes emptied,
and, after the steel cable ties had been retightened,
the central section of falsework was lowered, Figure 8.28. Falsework elements in the quarter-spans
were lowered by hand winches.
Spandrel columns were constructed next. Then,
following a carefully worked out sequence, the top
flanges of arch ribs over the central section were
concreted. Upon completion of the arch ribs the
de c k beams and slabs were placed, in the manner
previously described for the approach viaducts, in
a symmetrical and simultaneous manner on both
sides of the crown. After the deck had been prestressed transversely, it was prestressed longitudinally in the same manner as the approach viaducts.
Finished Viaduct 1 is shown in Figure 8.29.
In 1973, twenty-one years after the construction
of these arches, they were reevaluated to see how
they would now be designed and constructed. Figures 8.30 and 8.3 1 compare the actual project constructed in 1952 with the structure as it would have
been designed in 1970 (two boxes) and in 1973
(single box). The three-arch-rib and eight-beam
superstructure would be replaced by a variabledepth box section (cantilever construction using
precast segments) supported on slip-formed piers.
The arch remains an appealing and aesthetic
structure and might still prove to be competitive;
but perhaps the construction technique sug g e ste d
in the Neckarburg Bridge (Section 8.5.2) might
be more appropriate today, either cast in place or
precast.
FIGURE 8.28. Caracas Viaducts, lowering center
falsework.
FIGURE 8.29. Caracas Viaducts, finished Viaduct 1.
FIGURE 8.27. Caracas
false\\x)rk.
Viaducts,
lifting
center
Gladesville Bridge, Australia
371
As constructed in 1952
Possible alternative in 1973
FIGURE
8.30.
Caracas
Viaducts, comparison of longitudinal sections.
8.4 Gladesville Bridge, Australia
This precast segmental arch bridge, completed in
1964, spans the Parramatta River between Gladesville and Drummoyne and serves a large section of
the northern area of the Sydney Metropolis, Figure 8.32.
After award of contract the contractors submitted an alternative design. They proposed that the
arch be built on fixed falsework, whereas in the
original design part of the arch was to be built on
floating falsework and towed into position. The
original design called for an arch span of 910 ft
(277.4 m). The alternate design increased the clear
span of the arch to 1000 ft (305 m) and eliminated
the necessity for deep-water excavation for the
arch foundations on the Gladesville, or northern,
side of the river.4
Total bridge length between abutments is 1901
ft 6 in. (579.6 m). The 1000 ft (305 m) clear span
arch consists of four arch ribs, Figure 8.33, supported on massive concrete blocks, known as
“thrust blocks,” founded on sandstone on each side
of the river. Roadway width is 72 ft (22 m) with 6 ft
(1.8 m) wide sidewalks on each side. The roadway
has a grade of 6% at each end, and the grades are
connected by a vertical curve 300 ft (91.4 m) in
ngth over the center portion of the structure.
he arch has a maximum clearance, at the crown,
of 134 ft (40.8 m) above the water and not less than
120 ft (36.6 m) above water level for a width of 200
ft (61 m) in the center of the river.
Construction of the bridge involved the following main operations4:
1. Excavation for foundation of:
a. Arch thrust blocks on each side of the river
at the shoreline and partly below water.
b. Abutments at the ends of the bridge.
Shore pier columns of the approach spans
C.
on each side of the river.
2. Concreting of the arch thrust blocks, the
abutments and columns.
3. Driving of falsework piles in the river and
erection of steel falsework to support the hollow concrete blocks and diaphragms forming
each of the four arch ribs.
4. Casting of the box-section segments of the arch
and diaphragms and the erection of the four
arch ribs one at a time.
5. Jacking each rib to raise and lift it off the
falsework.
6. Casting of concrete deck beams on each side of
the river.
7. Erection of the deck beams to form the roadway over the arch.
8. Paving of the concrete roadway and final completion of the structure.
As constructed in 1952
20.50
--wft
-t--
I
II
I
1
R
-.-
Possible alternative in 1970
zyxw
-
20.50
Possible alternative in 1973
FIGURE
372
8.31.
Caracas Viaducts, comparison of cross sections.
FIGURE 8.32. Gladesville Bridge, aerial view, from
reference 4.
The roadway deck is supported on pairs of prestressed concrete columns, Figure 8.33. The wall
thickness is 2 ft (0.6 m), except in the tall columns
above the arch foundation where the wall thickness
is increased by 6 in. (152 mm). At the top of each
pair of columns there is a reinforced concrete cap
beam to support the deck girders.
During construction it was necessary to provide
falsework to support the box segments and diaphragms that make up each of the four arch ribs
in the arch. The falsework was made up of steel
tubular columns on steel tubular pile trestles carrying spans of steel beams 60 ft (18.3 m) long and a
steel truss span of 220 ft (67 m) over a navigation
opening in the Gladesville (northern) half of the
FIGURE 8.34. Gladesville Bridge, arch rib falsework
and positioning of arch rib segment, from reference 4.
falsework. These falsework units were tied together and anchored at each end to the thrust
blocks, Figure 8.34. Piling was taken down to rock
in the river bed.
Steel columns, braced together, formed a tower
extending transversely the full width of the bridge
at the center of the falsework. Transverse mem-
FIGURE 8.33. Gladesville Bridge, schematic of four arch ribs, columns, and deck, from reference 4.
374
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
bers, extending the full width of the bridge, above
the waterline connected the pile trestles, Figure
8.34. The balance of the falsework was of sufficient
width to support one arch rib. Upon completion of
erection of an arch rib, the falsework was moved
transversely on rails on the transverse members of
the pile trestle to a position to enable erection of
the adjacent arch rib, until all arch ribs were
erected.
Equipment installed on the central tower lifted
the arch box segments and diaphragms from water
level and positioned them. The tower also served
as a lateral bent to stabilize the individual arch ribs
after they were self-supporting and until they were
tied together. 4
The hollow-box segments and diaphragms were
cast 3 miles (4.8 km) downstream from the bridge
site. The casting yard was laid out to accommodate
the manufacture of one arch rib at a time. Each
arch rib consists of 108 box segments and 19 diaphragms. Each arch-rib box segment is 20 ft (6 m)
wide, with depths decreasing from 23 ft (7 m) at
the thrust block to 14 ft (4.3 m) at the crown of the
arch, measured at right angles to the axis of the
arch. The length of the box segments along
the arch varies from 7 ft 9 in. (2.36 m) to 9 ft 3 in.
(2.82 m). After the box units were manufactured,
they were loaded on barges and transported to the
bridge site. The box segments and diaphragms
were lifted from the barges to the crown of the
arch falsework and winched down to their proper
position, Figure 8.34. Diaphragms are spaced at
intervals of 50 ft (15.24 m), serving not only to
support the slender columns that support the
roadway above but also to tie the four arch ribs together.
When the units were located in position on the
falsework, a 3 in. (76 mm) joint between the precast segments was cast in place. At two points in
each rib, four layers of Freyssinet flat-jacks were
inserted, with 56 jacks in each layer. The rib was
then jacked longitudinally by inflating the jacks
with oil one layer at a time, the oil being replaced
by grout and allowed to set before the next layer
was inflated. Inflation of the jacks increased the
distance between the edges of the segments adjacent to the ja c ks and thus the overall length of the
arch along its centerline. In this manner a camber
was induced into the arch rib, causing it to lift off
the supporting falsework. The falsework was then
shifted laterally into position to support the adjacent arch rib and repeat the cycle. Figure 8.35 is a
view of the completed four arch ribs, and Figure
8.36 shows the completed bridge.
FIGURE 8.35. Gladesville
Bridge, complctrd four
arch ribs, from reference 4.
8.5 Arches Built in Cantilever
Until the appearance of the concrete cable-stay
bridge starting in 1962 (see Chapter 9), long-span
concrete bridges were the domain of the arch
type of structure. Until 1977, with the completion
of the Brotonne Cable-Stay Bridge in France with a
span of 1050 ft (320 m), the record length for a
concrete bridge had always been held by an archtype bridge. When the Kirk Bridges in Yugoslavia
were completed in 1980, the larger arch with a
span of 1280 ft (390 m) once again regained for the
arch the record of longest concrete span.
FIGURE 8.36. Gladesville Bridge, \ ICI\ of completed
bridge.
Arches Built in Cantilever
375
Here is a brief chronology of record concrete
arch spans up to 1964:
1930, Plougastel Bridge, France: three spans of
611.5 ft (186.40 m)
1939, Rio Esla, Spain: 631 ft (192.4 m) span
1943, Sando, Sweden: 866 ft (264 m) span
1963, Arrabida, Portugal: 886 ft (270 m) span
1964, Iguacu, River Parana, Brazil: 951 ft (290 m)
span
1964, Gladesville, Sydney, Australia: 1000 ft (305
m) span
The concrete arch bridge does not enjoy the
favor it once did. Modern methods of bridge construction utilizing prestressing, cable stays, and
segmental construction have all but eliminated it
from contention as a economical bridge type.
However, with the application of these modern
methods to the older form, and given the proper
site conditions, concrete arches may regain some of
their lost popularity.
fc)
8.5.1 REVIEW OF COSCEPT; SUMMARY OF
STRUCTURES WITH TEAZIPORARY STz4YS
The use of temporary stays to facilitate the construction of arch bridges began, perhaps, with the
Plougastel Bridge. Temporary prestress tendons
were used to provide stability to the short arch
cantilever sections emanating from the arch foundations (see Figure 8.5). Prestressing tendons were
used to support the f-alsework of the Rio Esla
Bridge and were incorporated into the structure.
However, the more novel method, which is the
birth of today’s technology, was employed in the
construction of the Saint Clair Viaduct at Lyon,
France, by M. Esquillan. The stability of precast
segments was obtained by the use of temporary
stays.
In the construction of the Caracas Viaduct,
Freyssinet extended this concept by using temporary stays to support the falsework and construct a
much longer cantilever section of the arch. This
same stay system was then used to accommodate
the forces produced by lifting the center arch section falsework (see Section 8.3). This concept was
partially recaptured for the construction of the
Iguacu Bridge in Brazil, where the falsework of the
central portion of the arch was supported by temporary stays.
The first arch bridge to be constructed using the
concept of supporting segmental sections of the
FIGURE 8.37.
Concrete arches built in cantilever with
temporary stays. (n) With stays and pvlons. (h) With stays.
spandrel columns, and pylons. (c) With spandrel colunins, tie diagonals and stay’s.
arch by temporary stays is the Sibenik Bridge in
Yugoslavia. Falsework f-or an approximate length
of 88.6 ft (27 m) was supported on Bailey trusses,
which were in turn supported by temporary stays,
Figure 8.376, consisting of a combination of cables
and structural steel rolled shapes. This arch was
constructed in nine sections, four on each side and
the central closure section. A modification of this
concept was used for a second Yugoslav bridge at
Pag with a 634 ft (193.2 m) span constructed in
seven sections. A further modification was used for
the Van Staden Bridge in South Africa, Figure
8.37a, with a span of 656 ft (200 m).
A somewhat different concept is where, with the
assistance of spandrel columns, the stays act as
temporary diagonals during construction, Figure
8.37~. In this manner, the structure is built as a
variable-depth Pratt truss. This concept was used
for the Kirk Bridges in Yugoslavia. In some instances these temporary diagonal stays may be incorporated into permanent diagonals such that in
the final configuration the structure is a truss and
not an arch (see Section 8.7.3).
In summarizing the construction methods using
temporary cable stays, we find two basic categories:
I
I
I
I
I
I
i
I
I
’
i
I
__~
Longitudinal
section
(a)
Erection
at
scheme
a-a
approaches
b-b
at arch
Cross-sections
376
Cc)
Arches Built in Cantilever
377
FIGURE 8.38. (Opposite) Neckarburg Bridge, erection scheme and sections, from reference 5. (a) Longitudinal section. (h) Erection scheme. (c) Cross section.
Where the arch is supported directly by the temporary stays
Where the temporary stays act as diagonals of a
Pratt truss during construction
Characteristics of the arch bridges using this concept of temporary stays during construction are
presented in Table 8.2.
8.5.2 NECKARBURG
BRIDGE, GERMANY
This unique and contemporary arch-supported
structure, some 50 miles (80 km) southwest of
Stuttgart, crosses the Neckar River near Rottweil,
Germany. It is a part of the federal expressway
A-81 from Stuttgart to the west of Bodensee with a
connection to Zurich, Switzerland.
The original scheme proposed by German authorities consisted of a steel girder structure supported on tall piers. Designer-contractor Ed. Zublin, Stuttgart, developed an alternative design
consisting of twin concrete arches to support the
roadway. The proposal was to construct the arches
segmentally by the cantilever method and construct the twin single-cell trapezoidal box girders
for the roadway by the incremental launching
technique (see Chapter 7). The Austrian method
called the Mayreder system was used to construct
the arches without scaffolding.5,6
The roadway of this 1197 ft (364.98 m) long
structure is approximately 310 ft (94.7 m) above the
Neckar River, Figure 8.38. The 507 ft (154.4 m)
arch span, Figure 8.39, has a rise of 164 ft (49.85
m). Total roadway width is 102 ft (31.0 m). The
FIGURE 8.39. Neckarburg Bridge, completed arch
(courtesy of Willhelm Zellner).
FIGURE 8.40. Neckarburg Bridge, arch just before
closure (courtesy of Willhelm Zellner).
structure is constructed as two independent parallel structures with a 1.8 ft (0.54 m) gap in the median. Roadway spans are 98 ft (30 m) in the approach sections and 72.6 ft (22.14 m) over the arch.
Each independent arch rib is a two-cell box. The
arch ribs were constructed in symmetrical halves,
Figure 8.40. The cuEved formwork was 43 ft (13.1
m) long, the first 23.3 ft (7.1 m) of the form
clamped to the previously constructed arch segment and the remaining 19.7 ft (6 m) remained to
cast the next segment increment. The first 23.3 ft
(7.1 m) of arch segment at the arch foundation was
constructed by conventional forming methods.
There are 14 segments on each side of an arch rib
and a closure segment at the crown of each arch.
The exterior dimensions of each two-cell arch rib
are 21.3 ft (6.5 m) wide by 9.8 ft (3.0 m) deep. Exterior webs vary in thickness from 10 to 11 in. (260
to 280 mm), and the interior web is 6.3 in. (160
mm) thick. The arch rib was cast in two operations
-first the bottom flange and second the webs and
top flanges.”
Piers supported by the arch or independent
foundations are of a constant section and slipformed by conventional methods. Sliding bearings
are used at the abutments and the short stiff piers 1
and 13. The remaining piers are hinged to the
superstructure deck such that the elastic piers can
follow the superstructure movement.5’6
During construction, as each half-rib was cantilevered out from its foundation, it was supported
by a temporary system of Dywidag bar stays, Figures 8.38, 8.41, and 8.42. After completion of the
arch, the temporary stays were removed, except
those required to stabilize the arch during the incremental launching of the superstructure deck.
Dywidag bar stays were anchored either to a pier
foundation or to Dywidag rock anchors in the side
of the valley.5
TABLE
Ye a r
Na m e
Sib &k
01
(:o nwuc tlo n
Yug fe la via
8.2
Span,
Charact erist ics of Arch Bridges Const ruct ed wit h Cable St ay s
It.
(In)
Sta ic
A rc h
Stia y
Mrtho d
C ~,tlSt~U~tlO tl
Arc
II Type
x 07
Nine wc tio n\ o n
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
I Y64 -66
(24 6)
De < k ‘I’ypc
Sim ple
,,rrC a \ t
A rc h Sc he m e
Po ssib ility 01 c a r-
la lse wo rk o f X 8 . 6
g lrd e rs o l 7 6 . 4 tt
I- e rtm g
a nd
( 2 3 . 3 0 Ill) spa t,
a t the c rown b y a
rm d e c o ntinuo tt9
ba tte ry 01 hy -
Y5 It ( 2 7 a nd
2Y 111)
Re m a rks
the
thrust
dra ulic ja c ks
I Y66-67
634
‘I‘hre r
( 19 3 . 2 )
Io rm e d
stay\
I brc e - c e ll
fro m
ICC 1611-
Pos5ibiltty
g ulnr box
01 c a r-
re c ting the thrust
ro lle d \ tre l
a t the c ~o wtt b y a
sha pe \
ba tte ry o l hy -
a nd c a b le s,
a uxilia ry
pylo n,
dra ulic ja c ks
the lo ng e r stq
b e ing a ddttxm a lly
su}q ”“te d
o n toI-
UrnllS
Va t1 Sta tle n
Nie se nb a c hb ruc ke
South
A ustria
A l&a
A b o u t I9 7 0
I Y73
6 5 6
Se g m e nt\ 1 9 . 6 It
(200)
(20
Sim ple
pre c a st
m ) sp a n
m nd e
~O IIIIIIUO U\
Multif~le sta ys
Mo b ile
( 120)
wppo rting
pe t- m ittinK thr
do uble - T
6 5 .6
a rc h dire <
tly with
suc c e ssive
the a id o f a n a ux-
struc tio n
ilia ry pykm
( 6.5
fo rm s
(16.8
394
the
1x1)
Fixe d
g ird e rs 01 55 tt
m ) lo n g
C o ntinuo us
c on
Fixe d
Ho rinm ta l
\ pa n 01
It ( 2 0
111)
1092 It ( 3 3 2 . 8 m )
o f 21 11
to ng seg-
“,e”t S
Hvka wa ru
Ja pa n
1973- 74
55x
Fo rm wo ~ - k
Re c ta ng ula r two-
Ho llo w sla b o f 2 It
( 170)
tta lly tufq>o rte d by
pa r-
c e ll
( 0 .6 0
a sta y fo r- the firs1
sp ring ing s,
SC<
bo x. Ne a r the
the
m ) thtc
nes5; 5 0
k-
I t ( I5 m )
width inc re a se s
c o ntinuo us
the a butm e nt a nd
Itne a rly fro m 26 to
< r,nstru<te d ,,I a
the hrst Fpa ndre l
52 It (X to I6 m ) to
\ pa n- b y- spa n
c o lum n. A lte r-
,m prwe
m o va ble la lwwo rk
wa rd, ~ o nstru-
e ra 1 sta hdttv
lio n be twe e n
tio n b y suc c e ssive
ca ntile ve rs by
\ c g m e nts
01 I I I t
( 3 . 4 111) Ie nfg h.
the tat-
spa n\
c urva -
ture o f d e c k is R =
Hm g rd
‘it th e two
qwing m g s
a rc h
o f thr
Krum m ha rhb r- ilrke
Switxrla nd
1976- 77
407
Constructrd
(124)
se g m e ntc
I”
rd 2 0 . 5
tt (6.25 m ) le ng th
‘fwo
pa ra lle l
Do ub le ‘I’ wth
a rc he s. dla -
c o ntinuo us
phra g m
va rymg
a t the
be twe e n
c o lum ns. t:a c h
32.8
a rc h
( IO to 20 111)
IS a so lid
Fixe d
c p ns
‘I”1 6 5 . 6 II
r- e < ta ng ula r rib 3 6
x 6 . 6 It (1.1 x 2 . 0
m)
Ne c ka rhurg
(&m a ny
I9 7 7
507
(:onstrurted of
I’wo p”‘“lle l
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
wc c e ssive se g a rc he s. e a c h < o n-
IG xe d
( 154.4)
m e nts 19.7
Sc hwa r/ wa swhrtrke
Switze rla nd
1977- 79
It (6.0
+Isting o f a re c ta w
rn) In le ng th
g ula r twwe ll box
374
Suc c e sive
Re c ta ng ula r \ la h
(114)
tile ve r se g m e nts
c url-
16.4 to 17.7
E‘ixe d
Ra ilro a d
b ridg e
Fixe d
Ra ilro a d
b ridg e
It (5
to 5.4 m ) 111 le ng th
Aka ya g wa
JZ3ptl
197R
4 I3
C o nstruc te d
with
A rc h c o nsistc
(126)
m o hile l.o rm s
suc h
re c ta ng ula r
of a
rwr,- <e ll box
thin
tha t a c o m ple te
sla b be twe e n two
p a nfzl wa s c a st, in-
C O lUm tIS
c tuding the a rc h.
the c o lum n, a nd
the de < k
Kirk Bridg e
(sm a lle r a rc h)
Yug o sla via
197x
x0 0
st‘ lys uwtl ‘IS
ue < t‘ rnp I‘ N
Fixe d:
(244)
dia g o na ls ot a
thre e - c e ll box
ot c o rre c ting the
Pra tt
tn,s\
po sslhltity
thrust a t the
c rown hy a ha tte ry
ot hyd ra ulic
jac ks
Re c ta npda r
(Sa m e a s ;tbove .
thre e - c e ll box
fo r sm a lle r a rc h)
FIGURE 8.41. TSecharburg
Bridge, temporaq
Dvwi-
dag bar stays supporting cantilevered arch rib (courtesy of Willhelm Zellner).
FIGURE
8.43.
Neckahur-g Bridge, hudling
of’ deck
girder.
FIGURE 8.44. Neckarburg Bridge, close-up of
launching nose.
FIGURE 8.42. Neckarburg Bridge, temporary Dywi-
dag bar stays supporting cantilevered arch rib (courtesy of Willhelm Zeliner).
The trapezoidal box girders of the superstructure deck were constructed behind the Singen
abutment and incrementally launched “downhill”
toward the Stuttgart abutment, Figure 8.43. A
close-up of the launching nose is shown in Figure
8.44. Overall girder width is 48.8 ft (14.9 m) with a
constant depth of 7.5 ft (2.3 m). Girder segments
were cast in lengths of 65.6 ft (20 m). The lift and
push combination of hydraulic ja c ks (see Chapter
7) launched the girder in 10 in. (0.25 m) increments. To maintain deformations of the arch and
piers, resulting from the horizontal forces of the
incremental launching operations, within allowable
limits, the tops of the piers were tied back to the
abutments and the arch was tied back by the temporary stays used during the arch construction. An
innovation introduced by Zublin on this project
was the use of bearings for the incremental
launching that remained as permanent bearings.
Prior procedure had employed a system of temporary bearings for the incremental launching and
then a transfer to permanent bearings5
8.5.3
NIESENBACK
BRIDGE,
AUSTRIA
This is a two-rib arch structure utilizing the free
cantilever construction method for each half-arch,
Figure 8.45. The arch has a span of 394 ft (120 m)
with a rise of 123 ft (37.5 m). Each arch rib is a
two-cell box with exterior dimensions of 16.4 ft (5
m) wide by 8.2 ft (2.5 m) deep. The roadway consists of a concrete slab and girder system with an
overall width of 57.7 ft (17.6 m). Although the longitudinal axis of the arch is in a straight line, the
Arches Built in Cantilever
Structure during construction
WJO
FIGURE
8.45.
-
Final structure
I
I
Hll‘SPYlO”
590
!=
s&m
cl
Niesenback Bridge, elevation, plan, and cross section, from reference 7.
roadwav it supports has a centerline radius, in
plan, of 1092 ft (332.8 m).
The curved roadway structure has spans of 65.6
ft (20 m) over the arch and is supported by two 3.3
ft (1.0 m) square piers, one on each arch rib. At
the arch foundations, roadway support is by a wall
pier with dimensions of 4.6 ft (1.4 m) by 33.8 ft
(10.3 m).
Each two-cell box arch rib is constructed by the
cantilever method, using a 41 ft (12.5 m) long
traveling form. The form clamps to the preceding
construction such that a 19.7 ft (6.0 m) segment
can be cast. A crew of seven men was able to cast a
segment on a weekly cycle.
To keep moments in the cantilevering arch to a
minimum during construction, the cantilevered
382
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
portion of the arch was supported by a system of
Dywidag bar stays, Figure 8.45. Stay stresses are
monitored at each stage of construction to maintain a nearly moment-free condition in the arch.
Dywidag bars used in the stays were 1 in. (26.5
mm) diameter and were used because they were
easily coupled and could be reused.’
8.5.4 KIRK BRIDGES, YUGO SLAVIA
These structures connect the mainland with the
Island of Kirk in the Adriatic Sea. In between is a
small rocky outcropping known as St. Mark, such
that from the mainland to St. Mark is the world’s
longest concrete arch with a span of 1280 ft (390
m) and from St. Mark to Kirk is the seventh longest
concrete arch with a span of 800 ft (244 m), Figures
1.40 and 8.46.
Because the distance between the shores of the
mainland and St. Mark is 1509 ft (460 m), the arch
support is partially founded in the sea, Figure 8.47.
The arch reaction of approximately 15,400 tons
(14,000 mt) is accommodated by the inclined pier
in the sea, which takes 9900 tons (9000 mt) to the
rock, while the nearlv horizontal box structure
1/3.00
above sea level takes the other reaction component
of 6600 tons (6000 mt).
A system of temporary stavs was used to support
the arch as it was progressively cantilevered out
from the springings, Figure 8.48. These temporar)
stays were used as the top chord and diagonals of a
temporary variable-depth Pratt truss during construction, Figures 8.48 and 8.49. The arch rib consists of a three-cell rectangular precast box, which
was cast in segment lengths of 16.4 ft (5 m) and
assembled with cast-in-place joints, Figure 8.48. .4
view of the completed arch with spandrel columns
is given in Figure 8.50.
8.6 Rigid-Frame Bridges
Another bridge type that lends itself to the contemporary segmental concept is the rigid-frame
bridge. Unfortunately, segmental construction has
not often been applied to this type of structure.
The reason is probably that the segmental concept
is associated with the conventional girder type
bridge, and designers have given little consideration to applying this method to the rigid-frame
bridge. Hopefully, the few examples that follow
will stimulate thinking about this type of structure.
!
Section 1
Section 2
ELEVATION
FIGURE 8.46. Kirk Bridges, elevation and sections.
zy
390m
L
I
/
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300 t
0 0 00
~
, -'lo.
-
,
33 50
1
-
00
, 0 00
-
-19OQ
33 50
FIGURE 8.47. Kirk Bridge, foundation detail.
383
FIGURE 8.48.
tion.
Kirk Bridge, erection of first arch sec-
FIGURE 8.50. Kirk Bridge, completed arch.
FIGURE 8.51. Saint Xlichael 131 itigc, LI~‘M ot the completed structure.
FIGURE 8.49. Kil k Bridge, erection apploarhing
crown.
7L’O’
_-
-
__- _-__. .--
FIGURE
8.6.1
8.52.
I-.--“..
Saint
--_
Michael
2Tr;LO’---
--.
Bridge,
partial
SAINT M ICHEL BRIDGE IN TOULOUSE,
FRANCE
This beautiful structure, Figure 8.51, appears as a
succession of arches w ith inclined legs, crossing the
two branches of the Garonne River in the southern
city of Toulouse, France. Typical dimensions of a
rigid frame are presented in Figures 8.52 and 8.53.
Because the bridge replaced an obsolete structure resting on masonry piers, it was possible to
longitudinal
section
construct the inclined legs on suspended scaffolding using temporary ties anchored to the masonry
piers before they were demolished, Figure 8.54.
The longitudinal girders were cast in place between the legs to complete the rigid frame. Over
each pier an expansion joint with laminated bearings is provided in the roadway slab, Figure 8.54.
Another view of the finished bridge is presented in Figure 8.55.
FIGURE 8.53. Saint Michael Bridge, typical section.
9
4
@oprm c be a ring
Q
4zyxwvutsrqponmlkjihgfedcba
4
Q
‘. .
‘A Extcling
moaonry pie r
FIGURE 8.541. Saint Michael Bridge, construction sequence at typical pier.
FIGURE 8.55. Saint Michael Bridge, finished structure.
FIGURE 8.56. Briesle Maas Bridge, general view.
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
386
8.6.2 BRIESLE MAAS
BRIDGE, NETHERLANDS
The Briesle Maas Bridge near Rotterdam, completed in 1969, is a distinctive structure with its
V-shaped piers, Figure 8.56. This bridge, crossing
the Meuse River, is situated in an area reserved for
pleasure boating and recreational purposes. It was
therefore considered essential to maintain a high
degree of bridge aesthetics. Although the design is
perhaps not the most economical, it was chosen to
meet the aesthetic requirements.
The three-span superstructure consists of a 369
ft (112.5 m) center span with end spans of 264 ft
(80.5 m). Transversely, the superstructure consists
of three precast single-cell boxes, joined at their
flange tips by a longitudinal closure pour and
transversely prestressed, Figure 8.57. The hollow
inclined legs of the V piers are structurally connected to the deck structure by post-tensioning, and
the V pier is supported at its base through neoprene bearing pads on the pile cap foundation, Figures 8.58 and 8.59. The superstructure, with the
exception of a few cast-in-place closure joints, is
composed of precast segments.
Shear forces, mainly concentrated in the webs,
normally are transferred to piers or columns by a
diaphragm. Prefabrication prevented this solution
as-25
485
CROSS
FIGURE
8.57.
in this project, however, as the additional weight in
the pier segments would have increased intolerably. Shear stresses were maintained at an acceptable level by increased web thickness and by triaxial
prestressing.
At the moment that the midspan closure pour of
the center span is consummated, the bending moment at this joint is zero. With time this moment
increases, as a result of creep, to a significant percentage of what would occur if the bridge were
built as a continuous structure on falsework. Prestressing to accommodate both conditions cannot
be given maximum eccentricity, and it becomes
both difficult to execute and expensive. .4 considerable amount of prestressing was saved by
eliminating the condition of zero stress at closure
and therefore preventing creep. This was accomplished by inducing an upward reaction under
segments 7 and 72, Figure 8.59, after joint closure.
Simultaneously with the increase of these reaction
forces, prestressing tendons in the central span
were stressed. Upon completion of the end spans
the induced forces were released automatically by
prestressing the end spans.
Segments were produced at an existing casting
yard 68 miles (110 km) from the bridge site. A
long-line precasting bed (see Figure 11.37) was
3525
i
485
Lc.75
/
SECTION
Briesle Maas Bridge, transverse cross section.
LONGITUDINAL SECTION WITH CABLE PROFILE
FIGURE
8.58.
Briesle Maas Bridge, longitudinal section with tendon profile.
Rigid-Frame Bridges
A=
B=
C=
D=
E =
Steel frame
Jacks
R u b b e r bearing p a d s
lolnts
Counter weight
387
F = Joint
G = Temporary support
H = Scaffolding
J = Joint
FIGURE 8.59. Briesle Maas
used w ith a length equal to a half-span-that is,
one cantilever. Three sets of segment forms were
employed to cast a total of 234 segments, averaging
78 reuses. Segments were transported to the
bridge site by barge.
The various stages of erection are indicated in
Figure 8.59. A special structural steel frame was
used to position the inclined precast hollow-box
legs of the piers and to support the seven precast
roadway girder segments before casting the joints
at the corners of the delta pier portion of the
structure. This frame was also utilized to balance
the pier during erection of the remainder of the
roadway girder segments and to adjust, by means
of ja c ks, the loads in the inclined legs of the pier
during various stages of erection.
Upon completion of the balanced cantilever
erection about both piers, temporary supports
were placed under segments 7 and 72 (the extreme
end segments of the partially completed end spans)
so that the temporary steel frames under the piers
could be removed. At this point both halves of the
structure were in an unstable equilibrium condition,
therefore, counterweights were placed over the
supported segments, Figure 8.59, to prevent the
half-structures from toppling over.
Jacks atop the temporary supports were used to
adjust the position of the bridge halves with respect
to one another and to induce the upward vertical
reaction forces previously discussed. Also, differences in elevation between the three box girders
Bridge, erection sequence.
were adjusted by these jacks. After casting the
center-span closure joint and stressing in the
center span, the remaining segments in the end
spans were placed on falsework, Figure 8.60; closure joints w ere cast; and longitudinal and transverse prestressing was completed.
All segments in the balanced cantilever portion
of the structure were placed b y a floating crane.
Because of the crane’s small reach, it could not
place the last five segments needed to complete the
end span. Therefore, it placed them on a small
dolly installed on top of the falsework, which
would roll them into their final positions. To avoid
dismantling the falsework after completing one
girder and reinstalling it under the next, it was
constructed so that it could be lowered and moved
transversely into position, Figure 8.60.
A close-up of the piers of the finished structure
is show n in Figure 8.61.
8.6.3 BONHOM M E BRIDGE, FRA.VCE
The Bonhomme Bridge over the Blavet River in
Brittany, France, was designed and built between
1972 and 1974, Figure 8.62. This three-span
slant-leg portal-frame bridge has a center span of
481 ft (146.7 m) and end spans of 223 ft (67.95 m),
Figure 8.63. The span between the foundations of
the slant legs is 611 ft (186.25 m). A tubular steel
framework was used to support the slant legs temporarily until closure at midspan, Figures 8.64 and
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
I
smcu A-.
-k--J
FIGURE
span.
8.60.
Briesle Maas
zyxw
Bridge, erection falsework for last five segments in the end
8.65. This structure was built by the cast-in-place
balanced cantilever method.
For adjusting the geometry of the bridge, flat
jacks were placed under the legs and at midspan. A
detail of the adjusting ja c ks placed on top of the
temporary support is shown in Figure 8.66. Flat
jacks and sand boxes were used both to adjust
the geometry of the bridge before closure was
achieved at midspan and later to release the energy
stored in the legs of the temporary supports, which
were loaded with the full weight of the bridge.
FIGURE 8.61. Briesle Maas Bridge, close-up \iew of
V piers.
FIGURE 8.62.
KERVIGNA&
&ORIENT
L
N
Ronhomme Bridge over Blavet River.
67.95
282.60
146.70
67.95
/
FIGURE 8.63. Bonhomme Bridge, elevation.
zyx
TW'FII FRS FOR CIP CONSTRUCTlON
A
FIGURE 8.64. Bonhomme Bridge, construction stages.
+
++
++
++
+
+
+
+ +
++ +i
\
FIGURE 8.65. Bonhomme Bridge, temporary support
389
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
390
The scheme is a very satisfactory one in terms of
both the aesthetics of the finished structure and
simplicity of construction. However, it may be used
only when site conditions allow the foundations of
the temporary supports to be established safely at a
reasonable cost. Figure 8.67 shows the temporary
supports during the balanced cantilever construction of the bridge.
8.6.4 MOTORWAY OVERPASSES IN THE
MIDDLE EAST
The use of precast segmental construction for the
Alpine Motorways in southern France was described in Section 3.15. It w as show n how mass
production could be applied to the construction of
a large number of similar overpasses.
This experience was repeated recently in a middle eastern country for the construction of 17
overpass structures over an existing freeway, Figure 8.68. To minimize disturbance of freeway
traffic, it was felt that a three-span rigid-frame
structure with inclined legs would be an attractive
solution.
Dimensions are show n in Figures 8.69 and 8.70.
The total deck length of 252 ft 3 in. (77 m) is divided into 32 precast segments for each of the twin
box girders. Deck width of the overpasses is either
36 ft (11 m) or 46 ft (14 m). The same box section is
used for all structures, and the cast-in-place longitudinal closure strip varies as required.
The slant legs are precast in the same plant
where the deck segments are produced. The typi_.
,.
cal erection sequence is show n in Figure 8.71. A
;zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
temporary bent founded at the edge line of the
new freeway is used to place and adjust the precast
legs on either side of the bridge. Segments are
placed in balanced cantilever from the special segment located atop the slant legs. A light temporary
bent in the short side spans is used to reduce the
bending moment in the slant legs during construction.
After completion of the deck and removal of all
temporary supports, the structure is in effect a
two-hinged arch with vertical restraints at both
ends. The bridges were analyzed for earthquake
13'- i 5
'
and large thermal variation loads (seasonal variation of 120°F and temperature gradient between
top and bottom flange of 18°F).
Figure 8.72 shows a detailed view of the inclined
legs and the temporary support during construcFIGURE 8.65. (Continued)
tion.
CONI’RFTF
CAF
FIGURE 8.66. Bonhomme Bridge,
details of bearing of concrete cantilever on temporary support.
zyxwvutsrqponmlk
b
. ,
h
392
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
FIGURE 8.67. Honl~ornrne
construction.
Bridge, during cantilever
I I--/ : r] ;-y
j /
+----: : ‘--:
I //j, /
B
$1
z
1
Plain
concrete
,
,9’-0”
19’-0”
FIGURE 8.70. Motorway Overpass Frames, cross section and elevation of inclined legs.
FIGURE 8.68. Motorway Overpass Frames, general
view.
8.7.1
RETROSPECT ON CONCEPTS FOR CONCRETE
TRUSS BRIDGES
Trusses were used in all long-span cantilever steel
bridges, and it was logical to conceive of the application of this type of structure to prestressed concrete. An interesting example of such an approach
is presented in Figure 8.73, in which an original sketch made in 1948 by Eugene Freyssinet for
the design of a precast prestressed concrete truss
is reproduced. The studies were applied to two
specific examples:
8.7 Truss Bridges
As with rigid frames, segmental construction has
seldom been applied to truss bridges. Once again
the designer must realize that the principles of
segmental construction, together with imagination,
can be applied to bridge structures other than the
conventional girder bridge.
149.3
I
3Segments-8’0”
FIGURE
I
51’-6”
“:
I
I
8.69. Motorway Overpass Frames, longitudinal section.
A
I
Truss Bridges
393
FIGURE
8.71. M otorway O verpass Frames, erection sequence.
(a) Stage 1. (b) Stage 2. (c) Stage 3. (d) Stage 4.
6
A bridge over the Hanach River near Algiers,
Algeria, with a clear span of 400 ft (123 m), Figures
8.74 and 8.75.
A major crossing of the Rhine River at Pfaffendorf, Germany, with a main span of 600 ft (180 m)
FIGURE 8.72. M otorway Overpass Frames, detail of
inclined leg and temporary support.
These studies were very encouraging from the
viewpoints of both economy of materials and
simplicity of construction. The deck was to be entirely precast, with members assembled by prestressing. Construction would proceed in balanced
cantilever from the main piers until reaching
midspan closure, where adjustment of the deck
geometry and loads in the members was provided
by jacks.
FIGURE 8.73. Original sketch of E. Freyssinet for a concept of prestressed precast
concrete truss (1948).
__ ELEVATION -
FIGURE
8.74.
Concept of a truss bridge.
l/2 COUPE B-B
l/2 C O U P E A - A
FIGURE 8.75.
Concept of a truss bridge.
The use of I girders at 7 ft (2 m) spacing for the
precast deck would not be considered today as the
optimum design. One of the authors, who was involved in the studies with E. Freyssinet, remembers
also that many technological problems such as the
connection details between diagonals and chords
were not completely solved.
Neither of these two designs reached the construction stage, and the concept was rapidly forgotten before its potential could be objectively ascertained.
Oddly enough, the designers of steel structures
followed a similar path. Abandoning prematurely
the concept of truss structures, which had allowed
such outstanding structures as the Firth of Forth
Bridge to be built all over the world, they turned to
web girder structures and closed box sections with
all the critical problems they entailed, such as elastic stability. Perhaps it is time to reassess some
major design approaches in both steel and concrete
for very long spans.
8.7.2 MANGFALL
anced cantilever, as construction started at one
abutment and proceeded to the opposite abutment
by progressive placement. Temporary intermediate piers were used as required to reduce the
cantilever stresses.
Figure 8.77 shows an interior view. The lower
flange is used as a walkway for pedestrians and for
bicycles. The railing in the center surrounds an
opening in the bottom flange where stress conditions do not require the concrete area. Figure 8.78
is an interior view looking through one of the floor
openings, and Figure 8.79 is another interior view.
8.73 RIP BRIDGE, AUSTRALJA
The recently completed Rip Bridge, Figure 8.80,
north of Sydney, Australia, has a center span of
BRIDGE, AUSTRIA
The Mangfallbriicke in Austria, Figure 8.76, on
the autobahn between Munich and Salzburg was
constructed in 1959. This structure is perhaps best
described as a large box girder with the webs being
a trusswork. Total length is 945 ft (288 m) from
abutment to abutment; the center span is 354 ft
(108 m) with side spans of 295.5 ft (90 m). It was
constructed as cast-in-place segmental using the
free cantilever method. However, it, was not bal-
FIGURE 8.76. Mangfallbriicke, general view.
396
Concrete Segmental Arches, Rigid Frames, and Truss Bridges
FIGURE 8.77. hl;~t~gf;llIbt.ucke,
trusswork.
FIGURE 8.78. Mangfallbl.iicke,
through floor opening.
FIGURE
8.79.
interim view showing
FIGURE 8.80. Rip Bridge, general view.
interior view looking
M,tngfallbrticke, general intertor view.
600 ft (182.88 m). The identical cantilever trusses,
which sit symmetrically on either side of the crossing, reach out 240 ft (73.56 m) toward each other
to support a 122 ft (37 m) drop-in simple span at
their extremities, Figure 8.8 1.
The erection scheme is illustrated in Figure 8.82.
Note that cable stays were used as diagonal members during construction to support the arch segments. Temporary falsework bents were used at
each panel point of the truss on the landward side
of the main piers. Precast concrete elements were
delivered from a precasting plant some 80 miles
(130 km) from the site.
Each panel of the lower chords of the truss was
assembled from five precast I-shaped elements
with a 1 ft (0.3 m) longitudinal pour strip between
the flange tips. Similarly, the upper chord was assembled from five rectangular two-cell precast
members. Erection of one of the lower chord
members is shown in Figure 8.83. The exterior two
I-shaped lower chord members are supported by
the diagonal stays, while the interior three elements of the lower chord are supported by a transverse beam arrangement from the exterior two
during construction.
Each diagonal member was assembled from longitudinally split halves, which, when brought together, encase the diagonal prestress tendon stays,
incorporating them into the structure by concrete
poured in place between the two halves. The upper
chord or deck members are erected after the vertical members along with temporary falsework to
support the deck panels, while the cast-in-place
concrete is placed between the deck elements and
transversely
prestressed.
Truss Bridges
397
8.81. Rip Bridge, elevation and cross sections.
FIGURE
Prestress cable to
support lower member
f
FIGURE 8.83. Rip Bridge, erection of knver chord.
Abutment
I.
,
The deck performs as a prestressed concrete
tension member. As construction proceeds, additional prestress is progressively added to ensure
that the deck remains in compression.
8.7.4
Location of
<-ggri.n
CONCEPT FOR A CROSSING OF THE
ENGLISH CHANNEL
Certain projects for crossings, such as of the English Channel between France and Great Britain,
the Straits of Messina, and even the Straits of Gibraltar, have exerted a powerful fascination on the
minds of the great engineers of this century.
FIGURE 8.82. Rip Bridge, erection sequence.
FIG U RE 8.84.
E‘t-eyssinet’h c-oncept of‘ preconfinetl
Channel with a set-ies of‘ 2000 f’t (612 111) slxtns.
concrete at-ch
crossing the English
Refmences
Eu g ene Frey ssinet w as no exc ep tio n, and he
spent the last years of his long professional career
studying the crossing of the English Channel with a
series of 2000 ft (612 m) long prestressed concrete
spans. The many worthwhile ideas contained in
this concept are not likely to be developed soon, or
even by the turn of the century.
Figure 8.84 presents an elevation of a typical
2000 ft (612 m) span, which was contemplated as a
prestressed concrete composite truss. Major members of the truss were not of conventional prestressed concrete, because such high stresses had to
be accepted to keep the weight of the span within
acceptable limits. A new material to be use d for
that purpose had occupied Freyssinet’s mind for
several years and had even been laboratory tested
for confirmation of the concept. When a concrete
member is completely confined in an envelope that
creates permanently biaxial transverse compressive
stress, it will resist safely much higher stress than if
subjected to a monoaxial stress or reinforced conventionally with untensioned transverse reinforcing (such as spirals in a circular column).
From a technological point of view, the permanent active restraint creating the biaxial transverse
compressio:l is easily achieved in a member that
has a circular cross section by confining it in a
high-strength steel pipe or within a continuous spiral o f p restressing steel w ires, w hic h are prestressed at the time the concrete is cast.
This m aterial, w hich co uld be called “ preconfined
concrete,” has extraordinary properties
such as total absence of brittleness and a capability
399
to sustain several times as much longitudinal compressive stress as a reinforced concrete member
without excessive strains, provided it is initially
loaded to offset the initial strain.
Such a project and such a material could not be
developed in a short period of time. They are
mentioned here at the close of this chapter as a
conceptual heritage, which it is our duty to m a ke
functional.
References
1. E. Freyssinet, “ Largest Concrete Spans of the
Americas-Three Monumental Bridges Built in
Venezuela,” Cizd Engineering-ASCE, March 1953.
2 . Je a n
M uller,
“ L arg e s t C o n cre te S p an s o f th e
A m ericas- H ow
signed,”
3.
the
T hree
Brid ges
Civil Engineering-ASCE,
W ere
D e-
M arch 1953.
Ro b ert Shama, “ Largest C oncrete S p ans of the
.Americas-How
They W ere Built.” C/zlil Et/g/tff’f’r-
rng--AXE,
4. Anon.,
M arch 1953.
“ N ew
Bri d g e
o v er
Parram atta
Ri v er
at
Main Roaak, Journal of the Department
of Main Roads, New South Wales, December 1964.
5. Anon., “ Talbriicke Rottweil-Neckarburg,” ZublinRundschau, Heft 7/ 8, Dezember 1976, Stuttgart,
Gladesville,”
Germany.
6. “ Arch Slipformer Shuns Ground Support to Cross
Valley,” Engineering LXrews-Record, June 1, 1978.
7. Anon., “Niesenbachbriicke,
Bogen i m Frei en V orbau,” A ustria
York.
1970-74,
FIP C o ng ress 1974, N ew
.
zy
9
Concrete Segmental Cable-Stayed Bridges
9.1
INTRODUCTION
9.1.1
9.2
9.3
9.4
9.5
9.6
Historical
Review
9.1.2 Advantages of Concrete Cable-Stayed Bridges
9.1.3 Structural Style and Arrangement
LAKE MARACAIBO BRIDGE, VENEZUELA
WAD1 KUF BRIDGE, LIBYA
CHACOlCORRIENTE.5
BRIDGE,
MAINBRiiCKE,
GERMANY
TIEL BRIDGE, NETHERLANDS
9.1
ARGENTINA
Introduction
The concept of supporting a beam or bridge by inclined cable stays is not new , and the historical
evolution of this type of structure has been discussed in the literature.‘-‘j Although the modern
renaissance of cable-staved bridges is said to have
begun in 1955, with steel as the favored material,
in the last two decades a number of cable-stayed
bridges have been constructed using a reinforced
or prestressed concrete deck system. In recent
years several concrete cable-stayed bridges have
been built in the long-span range. In at least four
current projects, alternative designs in concrete
and steel have been prepared for competitive bidding. Cable-stayed bridges are extending the competitive span range of concrete bridge construction
to dimensions that had previously been considered
impossible and reserved for structural steel. To
d a te ,
approximately 2 1 concrete cable-stayed
bridges have been constructed, and others are
either in design or under construction. A tabular
summary of concrete cable-stayed bridges is presented in Tables 9.1 and 9.2.
400
9.7
9.8
9.9
9.10
PASCO-KENNEWICK
BRIDGE,
U.S.A.
BROTONNE BRIDGE, FRANCE
DANUBE CANAL BRIDGE, AUSTRIA
NOTABLE EXAMPLES OF CONCEPTS
9.10.1
Proposed Great Belt Bridge, Denmark
9.10.2
Proposed Dame Point Bridge, U.S.A.
9.10.3 Proposed Rock-A-Chucky Bridge, U.S.A.
REFERENCES
9.1.1 HISTORICAL REVIEW
Since the beginning of the cable-stay renaissance in
1955, whether for technical or other reasons,
structural steel has been the preferred construction material. In 1957, however, considerable excitement was generated when Prof. Riccardo
Morandi’s prize-winning design of a prestressed
concrete 1312 ft (400 m) center span cable-stayed
bridge for the Lake Maracaibo crossing was announced. Regrettably the Lake Maracaibo Bridge
was not constructed as originally conceived. The
modified structure, built in 1962, is generally considered to be the first modern cable-stayed bridge.
However, the Lake Maracaibo Bridge was preceded by two little-known concrete cable-stayed
structures.
The first concrete structure to use cable stays w as
the Tempul Aqueduct crossing the Guadalete
River in Spain. ’ Designed by the famous Spanish
engineer, Prof. Torroja, who has introduced many
original concepts in prestressed concrete, this
structure has a classical three-span symmetrical
cable-stayed bridge configuration with two pylons.
zyxwvu
Zntroduction
T ABLE
9.1.
Bridge
8
9
10
11
12
13
14
15
16
17
18
19
20
21
22
23
24
25
26
27
28
Tempul
Benton City
Lake
Maracaibo
Dnieper River
Canal du Centre
Polcevera
Viaduct
Magliana
Danish Great Belr
Danish Great Belt”
Pretoria
Barwon River
Mount Street
Wadi Kuf
Richard Foyle
Mainbrticke
ChacolCorrientes
River Waal
Barranquilla
Danube Canal
Kwang Fu
Pont de Brotonne
Carpineto
Pasco-Kennewick
M-25 Overpass
Ruck-A-Chucky’
Dame Poinr
East
Huntington”
Weirton-Steubenville’
Concret e
Cable-St ay ed
401
Bridges-General
Location
Type
Guadalete River, Spain
Yakima River, Wash., U.S.A.
Venezuela
Kiev, U.S.S.R.
Obourg,
Belgium
Genoa, Italy
Rome, Italy
Denmark
Denmark
Pretoria, S. Africa
Geelong,
Australia
Perth,
Australia
Libya
Londonderry, N. Ireland
Hoechst, West Germany
Parana River, Argentina
Tiel, Holland
Barranquilla,
Columbia
Vienna,
Austria
Taiwan
Normandy, France
Province Poetenza, Italy
State of Wash., U.S.A.
Chertsey,
England
Auburn, California, U.S.A.
Jacksonville, Florida, U.S.A.
East Huntington, W.Va., U.S.A.
Weirton, W.Va., U.S.A.
Aqueduct
Highway
Highway
Highway
Pedestrian
Highway
Highway
Highway & rail
Highway & rail
Pipe
Pedestrian
Pedestrian
Highway
Highway
Highway & rail
Highway
Highway
Highway
Highway
Highway
Highway
Highway
Highway
Rail
Highway
Highway
Highway
Highway
Dat a
Spans (ft)d
66- 198-66
2@57.5-170-2@57.5
525-5@771-525
216.5-472-2
16.5
2@220
282-664-689-460
476-176
multispans 1132
multispans 1148
2@93
180-270-180
2e116.8
320-925-320
230-689
485.6-308
537-803.8-537
312-876-312
228-459-228
182.7-390-182.7
220-440-440-220
471-1050-471
100-594-100
406.5-981-406.5
2e180.5
1300
650-1300-650
158-300-900-608
820-688
Year
Completed
1925
1957
1962
1963
1966
1967
1967
Delayed by funding
Delayed by funding
1968
1969
1969
1971
Project
abandoned
1972
1973
1974
1974
1974
1977
1977
1977
1978
1978
Design
completed
Design
completed
Under constt-uction
In design
“Design by White Young and Partners.
bDesign by Ulrich Finsterwalder.
“Alternative design with structural steel.
1 ft = 0.305 m.
The stays were introduced to replace two piers that
were found to be too difficult to construct in deep
water. Thus, the stays were introduced to provide
intermediate support in the main span.
On July 5, 1957, a stayed structure crossing the
Yakima River at Benton City, Washington, was
opened to traffic. Designed by Homer M. Hadley,
the structure has a total length of 400 ft (122 m)
with a center span of 170 ft (51.9 m) flanked on
each side by tw o continuous spans of 57.5 ft (17.53
m) each. A 60 ft (18.3 m) central drop-in span of 33
in. (0.84 m) deep steel beams is supported by
transverse concrete beams, supported in turn by
structural steel wide-flange stays. Continuous longitudinal concrete beams comprise the remainder
of the structure and receive support at their extremity, in the center span, from the transverse
concrete beams and steel stays.4*8
In the more than half-century that has elapsed
since Torroja’s Tempul Aqueduct, 2 1 cable-stayed
bridges have been constructed (Table 9.1). Thirteen, or 62%, of these structures have been con-
strutted in the past decade. In the last five years
nine have been completed, representing 43% of
the total. Within the last three years the span of
1000 ft (300 m) has been exceeded, and a current
design contemplates a span of 1300 ft (400 m). It
has taken almost a quarter-century to reach a span
contemplated by Prof. Morandi in his original design concept for the Lake Maracaibo Bridge. Be
that as it may, it is obvious from the statistics that in
recent years the concrete cable-stayed bridge has
been accepted as a viable structure.
9.1.2 ADVANTAGES OF CONCRETE CABLE-STAYED
BRIDGES
As engineers, we are aware that no particular concept or bridge type can suit all environments, considerations, problems, or site conditions. The
selection of the proper type for a given site and set
of circumstances must take into account many
parameters. The choice of material, in addition to
Concrete Segmental Cable-Stayed Bridges
402
T ABLE
9.2.
Concret e
Cable-St ay ed
Bridges-Dimensional
Paramet ers
P)hl
Stay
Planes
Bridge
I
2
3
4
5
6
7
8
9
10
I1
12
I3
14
15
16
17
18
19
20
21
22
23
24
25
26
27
28
Tempul
Benton City
Lake
Maracaibo
Dnieper River
Canal du Centre
Polcevera
Viaduct
Magliana
Danish Great Belt”
Danish Great Belt”
Pretoria
Barwon R i v e r
Mount Street
Wadi K u f
River Foyle
Mainbrticke
Chaco/Corrientes
River Waal
Barranquilla
Danube Canal
Kwang Fu
Pont de Brotonne
Carpineto
Pasco-Kennewick
M-25 Overpass
Ruck-A-Chuck)
Dame Point
East
Huntington
WeIrton-Steubenville
2
2
2
2
2
2
2
3
2
2
2
1
2
1
2
2
2
2
2
2
1
2
2
2
2
2
2
2
No.
stay
Stays Arrangement
1
1
1
3
4
1
1
2
16
2
2
2
1
2
13
2
2
1
1
2
21
1
18
2
20
21
15116
24
Height
Above
Deck
(f0
Pylon
Heightto-Span
Ratio’
14.1
139.4
95
65.6
148
111.5
0.07
315
41
43
49
177.5
360
172
155
151.8
0.27
0.44
0.16
0.42
0.19
0.52
0.38
0.19
0.17
-
52.5
-
0.15
-
Radiating
Fan
231
94.75
220
71
0.22
0.16
0.22
0.39
Harp
Radiating
Radiating
302
279.4
333.2
0.23
0.31
0.41
Radiating
Radiating
Radiating
Harp
Radiating
Fan
Harp
Harp
Radiating
Radiating
Radiating
Fan
0.18
0.20
0.30
0.21
0.23
Deck
Width
cf.0
Girder
Depth
(ft)
6.9
57
5.87
59
79
51.75’
46
15.8
6
15.75
42.5
98
101.5
47
101
37
51.8
67
63
41.3’
79.8
39
54
105.75
41
103.5
SpantoDepth
Ratio”
3.25
16.4
4.8
1.94
15
9.8-13.2
23.5
2.95
3
7
2
11.5-23
11.5
8.5
11.5
11.5
10
9.2
28.7
52.3
46.7
98.75
113
46
36
48
390
31
38.5
58.4
70
60
57
70
76
46
42.5
12.5
11.5
7
9
8.5
5-6
5
8.5
84
52
140
20
153
260
180
96.5
Girder
Construction
Typed
CIP
CIP
CIP/PC d-i-s
PC
PC
CIP/PC d-i-s
CIP/PC d-i-s
PC segments
CIP segments
CIP
CIP
CIP
CIPiPC d-i-s
PC segments
CIP
PC/CIP d-i-s
PC and CIP
CIP segments
PC and CIP
PC
PC and CIP
CIP
PC segments
CIP
PC segments
CIP and PC
Composite
Composite
“Design by White Young and Partners.
hDesign by Ulrich Finsterwalder.
( See Table 9.1 for major span dimensions.
“CIP = cast-in-place, PC = precast, d-i-s = drop-in-span.
’ Form hyperbolic paraboloid in space.
‘Per single-cell box.
1 ft = 0.305 m.
zyxwvutsrqponml
material properties, depends on availability and
the prevailing economics at a particular time as
well as the specific location of the site. The process
of weighting and evaluating these parameters for
various types of bridges under consideration is
certainly more an art than a science.
In evaluating a concrete cable-stayed bridge, the
designer should be aware of the following advantages:
1.
The main girder can be very shallow with respect to the span. Span-to-girder-depth ratios
vary from 45 to 100. With proper aerodynamic
streamlining and multistays the deck structure
can be slim, having span-to-depth ratios of 150
to 400, and not convey a massive visual impression.
2.
Concrete deck structures, by virtue of their
mass and because concrete has inherently
favorable damping characteristics, are not as
susceptible to aerodynamic vibrations.
3.
The horizontal component of cable-stay force,
w hich causes compression w ith bending in the
deck structure, favors a concrete deck structure. The stay forces produce a prestress force
in the concrete, and concrete is at its best in
compression.
The amount of steel required in the stays is
comparatively small. A proper choice of height
of pylon with respect to span can yield an optimum solution.g
4.
5.
Live-load deflections are small because of the
live-load-to-dead-load ratio, and therefore
Introduction
403
concrete cable-stayed bridges are applicable to
railroad or mass-transit loadings.
6.
Erection of the superstructure and cable stays
is relatively easy with today’s technology of
prestressing, prefabrication; and seg&&tal
cantilever construction.
9.1.3
STRUCTURAL STYLE AND ARRANGEMENT
Many of the concrete cable-stayed bridges have
been designed by Morandi or have been strongly
influenced by his style. Commencing with the Lake
Maracaibo Bridge, of the 12 bridges constructed,
excluding pedestrian and pipe bridges (see Table
9.1), six have been designed by Morandi, Figures
9.1 through 9.6. A third prize winner in the 1967
Danish Great Belt Bridge Competition was the
Morandi-style design proposed by the English consulting firm of White Young and Partners, Figure
9.7. The ChacoKorrientes Bridge, Figure 9.8, very
much resembles the Morandi style.
FIGURE 9.3. Magliana Viaduct (courtesy of L’Industria Italiana de1 Cemento).
FIGURE 9.4. Wadi Kuf Bridge, general consrruction
’
view (courtesy of Prof. R. Morandi).
FIGURE 9.1. Lake Maracaibo Bridge, general view,
from reference 11 (courtesy of Julius Berger-Bauboag
Aktiengesellschaft).
FIGURE
9.2.
Polcevera
Creek Bridge, general view.
FIGURE 9.5. Barranquilla Bridge (courtesy of L. A
Garrido).
404
Concrete Segmental Cable-Stayed Bridges
FIGURE 9.6. Carpineto Viaduct (courtesy of L’Industria Italiana de1 Cemento).
FIGURE 9.7. Danish Great Belt Bridge, artist’s
ering (courtesy of White Young and Partners).
These structures, with the exception of the Magliana, Barranquilla, and Carpineto bridges, are
typified by the A-frame pylon positioned in the
plane of the stays and an auxiliary X frame or inclined struts to support the deck structure at the
pylon. They are statically determinate systems so as
to preclude any possible damage from differential settlements of the bridge piers and pylons or
from light seismic shocks.
A simple schematic of the structural scheme is
shown in Figure 9.9, which consists of a series of
independent balanced systems, each carried by an
individual pier and pylon. These systems are then
connected by drop-in girders, which are simple
span girders spanning between independent systems.‘O The cantilever girder is supported at two
points (C and D) by a pier system and elasticafly
supported at two points (B and E) by the cable
stays, thus producing a three-span girder with
cantilevers on each side. The stays are supported
by a pylon portal frame that is independent of the
pier system supporting the girder.
Another entry in the 1967 Danish Great Belt
Competition by Ulrich Finsterwalder, of the German firm Dyckerhoff 8c Widmann, deviated from
the Morandi style and was awarded a second prize.
Finsterwalder’s design proposed a multiple-span,
multistay system using Dywidag bars for the stays,
Figure 9.10. The deck was envisioned as being constructed by the cast-in-place balanced cantilever
rend-
FIGURE 9.9. Schematic of Morandi-style structural
scheme, from reference 10 (courtesy of the American
Concrete Institute).
FIGURE 9.8. ChacoiCorr ientes Bridge, general \iew,
from reference 13 (courtesy of Normer Gray).
FIGURE 9.10. Da&h Gre,lr Belt Bridge, .I1 list’\ rrndering (courtesy of Ulrich Finsterwalder).
Lake Maracaibo Bridge, Venezuela
segmental method, each segment being supported
by a set of stays. This concept was later to be consummated in the Main Bridge and in the design of
the Dame Point Bridge.
The choice of geometrical configuration and
number of stays in a cable-stayed bridge system is
subject to a wide variety of considerations. If cable
stays are few, they result in large stay forces, which
require massive anchorage systems. A relatively
deep girder is required to span the large distance
between stays, producing span-to-depth ratios varying from 45 to 100 (see Table 9.2). Depending
upon the location of the longitudinal main girders
with respect to the cable-stay planes, large transverse cross girders may be required to transfer the
stay force to the main girder.
A large number of cable stays, approaching a
continuous supporting elastic media, simplifies the
anchorage and distribution of forces to the girder
and permits the use of a shallower girder, with
span-to-depth ratio varying from 150 to 400 (see
Table 9.2). The construction of the deck can be
erected roadway-width by free cantilever methods
from stay to stay without auxiliary methods or
stays. If the depth of the roadway girder can be
kept at a minimum, the deck becomes, more or
less, the bottom chord of a large cantilevering
truss; it needs almost no bending stiffness because
the inclined stays do not allow any large deflections
under concentrated loads.”
In the 55 years since Torroja’s Tempul Aqueduct the concrete cable-stayed bridge has evolved
from basically a statically determinate structure
with one stay on each side of the pylon to a
highly indeterminate system with multistays. As
demonstrated by the Danish Great Belt Bridge
Competition, the Pasco-Kennewick Bridge, and
the Pont de Brotonne, spans of approximately
1000 ft (300 m) are practical and have been accomplished. The practicality of spans of 1300 ft
(400 m) is demonstrated by the Dame Point
Bridge, and spans approaching 1600 ft (500 m) are
considered technically feasible. Leonhardt” has
projected that with an aerodynamically shaped
composite concrete and steel deck a span of 2300 ft
(1500 m) can be achieved. With today’s technology
of prefabrication, prestressing, and segmental
cantilever construction, it is obvious that cablestayed bridges are extending the competitive span
range of concrete bridges to dimensions that had
previously been considered impossible and int.o a
range that had previously been the domain of
structural steel. This technological means exist;
they only require implementation.
405
9.2 Lake Maracaibo Bridge, Venezuela
This bridge, Figure 9.1, has a total length of 5.4
miles (8.7 km). Five main navigation openings consist of prestressed concrete cable-stayed structures
w ith suspended spans totaling 771 ft (235 m). The
cantilever span is supported on four parallel X
frames, while the cable stays are supported on two
A frames with a portal member at the top. There is
no connection anywhere between the X and A
frames, Figure 9.11. The continuous cantilever
girder is a three-cell box girder 16.4 ft deep by 46.7
ft wide (5 m by 14.22 m). An axial prestress force is
induced into the girder as a result of the horizontal
component of cable force, thus, for the most part,
only conventional reinforcement is required. Additional prestress tendons are required for negative moment above the X-frame support and the
transverse cable-stay anchorage beams. l1
The pier cap consists of the three-cell box girder
with the X frames continued up into the girder to
act as transverse diaphragms, Figures 9.12 and
9.13. After completion of the pier, service girders
were raised into position to be used in the construction of the cantilever arm. Owing to the additional moment, produced during this construction
stage by the service girder and weight of the cantilever arm, additional concentric prestressing was
required in the pier cap, Figure 9.13. To avoid
overstressing of the X frames during this operation, temporary horizontal ties were installed and
tensioned by hydraulic jacks, Figures 9.13 and
9.14.
FIGURE 9.11. Lake Maracaibo Bridge, pier cap with
X frames, from reference 11 (courtesy of Julius
Berger-Bauboag
Aktiengesellschaft).
IL
I
I
FIGURE 9.12.
II
Lake Maracaibo Bridge, main span tower and X-f’rames,
reference 11 (courtesy of Julius Berger-Bauboag Aktiengesellschaft).
I---,
Worklnq
Service qirder
’
\
\
/I I
from
I
I
for
FIGURE 9.13. Lake Maracaibo Bridge, pier cap of a main span and service
of Julius Berger-Bauboag Aktiengesellgirder, from reference 11 (courtesv
shaft).
406
407
Wadi Kuf Bridge, Libya
Bridge, brace mernFIGURE 9.14. Lake Maracaibo
bers bear against X frames after being tensioned by hy-
draulic ,jacks, from ref‘e rence 11 (courtesv of’ Julius
Berger-Bauboag
Aktiengesellschaft).
In the construction of the cantilever arms, special steel trusses (service girders) were used for
formwork. They were supported at one end by the
completed pier cap and at the other end by auxiliary piers and foundations, as show n in Figure
9.15.
The anchorages for the cable stays are located in
a 73.8 ft (22.5 m) long prestressed inclined transverse girder. The reinforcing cages for these
members were fabricated on shore in a position
corresponding to the inclination of the stays. They
FIGURE 9.15. Lake Maracaibo Bridge, placing service girder for forming cantilever girders, from referenc e 11 ( c o u rtesy o f Ju liu s Berg er- Bau b o ag Aktiengesellschaft).
w eighed 60 tons and contained 70 prestressing
tendons, Figure 9.16. The cable stays are housed in
thick-walled steel pipes, Figure 9.1’7, which were
welded to steel plates at their extremities and were
encased in the anchorage beam. A special steel
spreader beam was used to erect the fabricated
cage in its proper orientation. The suspended
spans are composed of four prestressed T sections.
9.3
Wadi
Kuf Bridge, Libya
The Wadi Kuf Bridge in Libya, designed by Prof.
Morandi, consists of two independent balanced
FIGURE 9.16. Lake >fal-acaibo Bridge, fabrication of anchorage beam, from reierence
11 (courtesy of Julius Berger-Bauboag Aktiengesellschaft).
408
Concrete
Segmental
Cable-Stayed
Bridges
9.4 ChacolCorrientes Bridge, Argentina
FIGURE 9.17. L&e hla~.rc&o
BI idge, bousing fo r
cable stays, from reference 11 (courtesy of Julius
Berger-Bauboag
Aktiengesellschaft).
cable-stay systems having their ends anchored to
the abutment by a short hinge strut. The cable-stay
systems are connected by a simply supported
drop-in span, Figure 9.4.
This structure consists of only three spans. The
center span is 925 ft (280 m) long and the two end
spans are each 320 ft (97.5 m), for a total length of
1565 ft (475 m). The simply supported drop-in
center portion of the main span consists of three
double-T beams 180 ft (55 m) in length; each beam
w eighs approximately 220 tons (200 mt).12
The A-frame towers are 459 ft and 400 ft (140
and 122 m) high and the roadw ay deck is 597 (182
m) above the lowest point of the valley beneath the
structure. l2 The superstructure is a single-cell box
girder that varies from 13 ft (4.0 m) to 23 ft (7.0 m)
at the pylons. The single-cell box is 24 ft (7.4 m)
wide and with cantilever flanges forms a 42.7 ft (13
m) deck.
The contractor made good use of traveling
forms to construct the box girder and deck, using
the balanced cantilever technique to build on both
sides of the pylons at the same time. Traveling
forms were used because extreme height and
difficult terrain made other conventional construction methods impossible or too costly. The
deck was constructed by progressive cast-in-place
segments, attached to the previously completed
segments by means of temporary prestress ties and
subsequent permanent post-tensioning Dywidag
bars. The procedure adopted required temporary
cable stays to support the cantilever arms during
the construction sequence as the superstructure
progressed in both directions from the pylon.
When the superstructure extended sufficiently, the
permanent stays were installed, and the structure
was completed in the same manner.
The ChacoKorrientes Bridge (also referred to as
the General Manuel Belgrano Bridge) crosses the
Parana River between the provinces of Chaco and
Corrientes in northeast Argentina and is an important link in one of the highways between Brazil
and Argentina, Figure 9.8. It has a center navigation span of 803 ft 10 in. (245 m), side spans of 537
ft (163.7 m), and a number of 271 ft (82.6 m) approach spans on both the Chaco and Corrientes
sides of the river. The vertical clearance in the
main spans above flood level is 115 ft (35 m).i3.14
The superstructure of this bridge consists of two
cast-in-place concrete A-frame pylons, which support a deck of precast segmental post-tensioned
concrete. The pylons are flanked by concrete
struts, which reduce the unsupported length of the
deck, Figure 9.18. Although the pier cap section of
the deck (between inclined struts) is cast in place,
the cantilever portion consists of precast segments.
The drop-in spans are cast in place.
The deck structure consists of two longitudinal
hollow boxes 8 ft 2% in. (2.5 m) w ide and w ith a
constant depth of 11 ft 6 in. (3.5 m), which support
precast roadway deck elements, Figure 9.19. The
precast girder elements were match-cast on the
river bank in lengths of 13 ft 1% in. (4.0 m), with
the exception of shorter units at the point of stay
attachment, w hich contain an inclined transverse
anchorage beam, Figure 9.20. Units were cast by
the long-line method on a concrete foundation
with the proper camber built in. Each unit was cast
with three alignment keys, one in each web and
one in the top flange. The units were erected as
balanced cantilevers with respect to the pylon to
minimize erection stresses. After a unit was hoisted,
an epoxy joint material was placed over all of the
butting area; then the unit was placed against the
already erected unit and tensioned.13
To eliminate the need for falsework, the tnclmed
struts and pylon legs were supported by horizontal
ties at successive levels as construction proceeded,
Figure 9.21. The legs were poured in segments by
cantilevering the formwork from previously constructed segments. When deck level was reached,
the girder section between the extremities of the
inclined ties was cast on formwork. To further stiffen the pylon structure, a slab was cast between box
girders at the level of the girder bottom flanges.
This slab is within the limits of the cast-in-place box
girders and inclined struts and serves as an additional element to accept the horizontal thrust from
the cable stays. The upper portion of the pylon was
409
ChacolCorrientes Bridge, Argentina
k-
Precast construction --+I- Cast-irt-place 4 f Precast construction +
369 ft 1 in. (112.50 ml -4
369ft 1 in.(112.50m)
~803 ft 10 in (245.00 m)
=I=
Center span
537 ft 0 in. (163.70 rn)M
Side span
FIGURE 9.18. Chaco/ Corrientes Bridge, longitudinal geometry, from reference 14 (courtesy of Civil Engineering-ASCE).
8; I”.
(22cm)
9 ft 2: in
n
’
(2.80 m)
8: i n .
I I
’ (22 cm1
nl
9 ft 2f in.
27 ft 3 in.
(8.30 m) Cast-w-place
concrete7
I
(2.8Om)
i in.
.’
; 11 ft 5; in.
Ii( 3 . 5 0’
/+8 ft 2f in.&-11
(2.50 m)
ft 3: in.
(3.45 ml
11 ft 3: in. -& 8 ft 2f in.4
(3.45 m)
ml
(25-30 c m )
(2.50 m)
FIGURE 9.19. Chaco/ Corrientes
Bridge, deck cross section, from reference
14 (courtesy of Civil Engineering-AXE).
1 each rade of cable
4 ,n. anchor bolt
,.
.,i.,...,.:,.. : ‘.. .
.
:.:..,;.;y.. ,;
,,. . Box girder -i: . . .! ‘.
.:.-:.
,.. . . ,?.:..:
.._ . ..*. .., . ,. :. 2, T.‘..
FIGURE 9.20. ChacoKorrientes Brid g e, c ab le anchorage at girder, from reference 14 (courtesy of Civil
Engineering-AXE).
then completed, using horizontal struts to brace
the legs until they were connected at the apex, Figure 9.21.r3*14
The precast box girder units, with the exception
of those at the cable-stay anchorage, were cast 13 ft
1% in. (4 m) in length by the long-line, match-cast
procedure. The soffit bed of the casting form had
the required camber built in. Alignment keys were
cast into both webs and the top flange. Match casting and alignment keys were required to ensure a
precise fit during erection. Each 44 ton (40 mt) unit
was transported by barge to the construction site
and erected by a traveling crane operating on the
erected portion of the deck. Since each box was
lifted by a balance beam, four heavy vertical bolts
had to be cast into the top flange of each box. The
lifting crane at deck level allowed longitudinal
410
Concrete Segmental Cable-Stayed Bridges
2.
Erect diaphragms between lines of boxes and
post-tension.
3.
Place temporary and permanent stays as erection proceeds.
4. Remove temporary stays.
5. Remove temporary post-tensioning in the cantilever sections.
6.
Place precast deck slabs between box girders.
7.
Concrete the three 65 ft 8 in. (20 m) drop-in
spans.
Place asphalt pavement, curbs, and railings.
8.
1
2
P---
FIGURE 9.21. ChacoiCorrientes
Bridge, erection sequence of pylon, front reference 14 (courtesy of Civil
Engineering-A SCE).
movement of the suspended box. Upon erection to
the proper elevation, the unit was held to within 6
in. (150 mm) of the mating unit while epoxy joint
material was’applied. Bearing surfaces of the unit
were sand-blasted and water-soaked before erection. The water film was removed before erection
and application of the epoxy joint material. The
traveling deck crane held the unit in position
against its mating unit until it could be post-tensioned into position. The crane was slacked off
without waiting for the joint material to cure.13*14
To minimize overturning forces and stresses in
the pylon, it was necessary to erect the precast box
units by a balanced cantilever method on both sides
of the centerline of the pylon. The erection schedule demanded simultaneous erection at each
pylon, although the pylons are independent of
each other. When four precast box units were
erected in the cantilever on each side of the pylon,
temporary stays were installed from the top of the
pylon to their respective connections at deck level.
After installation of the temporary stays, cantilever
erection proceeded to the positions of the permanent stays, and the procedure was repeated to
completion of the installation of the precast box
units.13
The erection sequence may be outlined as follows:
1. Erect precast boxes and post-tension successively.
9.5 Mainbriicke, Germany
The M ain Brid g e near H o ec hst, a su b u rb o f
Frankfort, constructed in 1971 is a prestressed,
cast-in-place, segmental, cable-stayed structure
that connects the Fabwerke Hoechst’s chemical industrial complex on both sides of the River Main
in West Germany, Figure 9.22. It carries two
three-lane roads separated by a railway track and
pipelines. This structure, a successor to Finsterwalder’s Danish Great Belt Bridge proposal, represents the first practical application of the Dywidag
bar stay. l5
The bridge spans the river at a skew of 70” from
the high northern bank to the southern bank,
which is 23 ft (7 m) lower. The center navigation
span is 486 ft (148.23 m) w ith a northern approach span of 86 ft (26.17m) and southern approach spans of 55, 84, 95, and 129 ft (16.91,
25.65, 29, and 39.35 m), Figure 9.23.
Railroad track and pipelines are in the median
between the two cantilever pylon shafts and are
supported on an 8.7 ft (2.66 m) deep torsionally
stiff box girder, Figure 9.24. The centerline of the
FIGURE 9.22. .1Ltinbt
nuke, from reference 16.
M ainbriicke,
zyxwvutsrqponmlkjih
. . . .\~;
_\ \\_ .LI~P .\\.~:
:
FIGURE
411
G ermany
9.23.
Mainbriicke, elevation and plan, from reference 16.
FIGURE 9.24. Ll;ti~~l~l-iicke,
cl-ass sections, f’rotl~
longitudinal webs of the box girder coincides with
the centerline of the individual cantilever pylons,
and they are 26.25 ft (8 m) apart. Transverse cross
beams at 9.8 ft (3 m) centers form diaphragms for
the box and cantilevers, which extend 39 ft (11.95
m) on one side and 36 ft (11 m) on the other side of
the central box to support the two roadways, Figure
9.25.
The cross section of the towers consists of an anchoring web in the center, sandwiched by two flatplate flange elements, Figure 9.26. In a transverse
elevation of the pylons, the width of the pylon increases from the top to just below the transverse
strut, where it decreases to accommodate clearance
requirements for both modes of traffic, Figure
9.26. The stay cables (Dywidag bars) are in pairs,
horizontal to each other in the main span and vertical in the side span, thus simplifying the anchorage detail at the pylon, Figure 9.26.1fi
FIGURE 9.25. Mainbriicke, view of deck at pylon
(courtesy of Richard Heinen).
rcf’evellcc\ l(i.
Ii]
VERANKERUNG DER
SCHR;LiGSEILE
IM PYLON
FIGURE 9.26. Mainbticke, pylon
and cable configuration, from refer_
ence 16.
Construction of the bridge superstructure was
by the cast-in-place segmental method, Figure
9.27. Segments in the river span were 20.7 ft (6.3 m)
in length, corresponding to the spacing of the
stays. Segments in the anchor span were 19 ft (5.8
m) in length. Segments in the anchor span were
concreted before the corresponding segment in
the river span to maintain stability. The pylon
segments were associated with the superstructure segments, and each pylon segment was slipformed.
Figure 9.28 shows the partially completed
structure and the falsework necessary to install the
stays. Each stay is composed of twenty-five 16 mm
(5/ s in.) diameter Dywidag bars encased in a metal
duct, which is grouted for corrosion protection
similar to post-tensioned prestressed concrete construction.
FIGURE 9.27, Mainbriicke, casting of deck segments
(co urtesy o f Dyckerho ff & Widmann).
9.6
Tie1 Bridge, The Netherlands
The Tie1 Bridge, l7 Fi gures 9.29 and 9.30, crosses
the Waal River, which, together with the Maas and
the Rhine, flowing east to west, divides the
parCall\
FIGURE 9.28. M ainb tic ke,
structure (courtesy of Richard Heinen).
complered
VIADUC
D’ACCF5
APPROACH VIADUCT
FIGURE
9.29.
‘l‘ie l
OUVRAGE
Brid g e ,
PRINCIPAL
g e ne ra l
la yo ut.
MAIN BRIDGE
Concrete Segmental Cable-Stayed Bridges
414
FIGURE
9.30.
l‘ie l Hric lg c ,
ma in
sp lls.
The ten-span 2648 ft (806 m) long access viaduct
is continuous over its entire length. The superstructure is supported on the piers by sliding
teflon bearings, except at the three center piers
where it is supported on neoprene bearings, having a thickness such that they ftx the viaduct at
these piers. Expansion joints are located at piers 1
and 11. The superstructure in the access viaduct
consists of two precast rectangular boxes of a constant depth of 11.5 ft (3.5 m) and width of 21 ft 8
in. (6.6 m). The top flange including cantilever
overhangs has a w idth of 44 ft (13.44 m). The
overall width of the approach viaduct deck is 89 ft
3 in. (27.2 m), including a longitudinal pour strip.
The viaduct was constructed by the precast balanced cantilever method with cast-in-place closure
pours at the midspans. To accommodate the cantilever compressive stresses in the bottom flange
over the piers, the thickness of the bottom flange is
linearly increased from a minimum of 8 in. (200
mm) to 24 in. (600 mm) over a length of 33 ft (10
m) on each side of the pier. Each pier segment
contains a diaphragm.
Because of the potential flooding of the river
from April through December and the consequent
loss or damage of falsework and loss of time, it was
decided to build the access viaduct utilizing precast
segments in the balanced or “ free” cantilever construction. The segments could be cast during
flooding and placed in storage. Erection of the
segments, which would take less time than the
casting, could be accomplished after the flood had
subsided.
The precast segments, weighing 132 tons (120
mt), were cast in movable forms on a casting bed
having the length of one span (by the long-line
method, see Section 11.6.2). Segments were stored
by and parallel to the casting bed and handled by a
130 ft (40 m) span gantry crane, Figure 9.32. They
were transported to the site (access viaduct abutment) by means of a 132 ton (120 mt) capacity
trolley and then placed in the structure by the same
gantry crane used in the precasting yard for handling, Figure 9.33; The trolley was used to transport the segments because the gantry was usually
engaged in the precasting yard or in placing segments in the viaduct. The gantry crane was such
that it spanned over the twin boxes in the superstructure and the trolleyway used to transport the
segments.
Segment joints are of the epoxy-bonded type
(see Section 11.5). Cantilever imbalance is accommodated by a temporary support ad.jacent to the
pier, Figure 9.33. Five temporary prestress bars
zyxwvu
Netherlands into northern and southern parts.
This structure provides a needed traffic link between the town of Tie1 and the south of the country and is a major north-south route.
The structure has an overall length of 4656 ft
(1419 m) and consists of a 2644 ft (806 m) curved
viaduct on a 19,685 ft (6000 m) radius, w hich includes ten continuous 258 ft (78.5 m) long spans
and a 2008 ft (612 m) straight main structure comprising three stayed spans of 3 12, 876, and 3 12 ft
(95, 267, and 95 m) and two 254 ft (77.5 m) side
spans.
The cross section consists of two precast concrete
boxes, each supporting two vehicular and one bicycle lane. The total width of the superstructure,
which is 89 ft (27.2 m) in the access viaduct, Figure
9.31, is enlarged to 103 ft (31.5 m) over the main
structure so as to accommodate the pylon supporting the stays.
The structure crosses not only the Waal River but
also a flood plain, w hich is under w ater during the
winter months. Navigation requirements dictate a
horizontal clearance of 853 ft (260 m) and a vertical clearance of 30 ft (9.1 m).
Tie1 Bridge, The Netherlands
415
FIGURE 9.32. Precasting plant. (1) Casting bed, (2)
I-e-bar storage, (3) segment storage, (4) concrete batch
plant, (5) office, (6) gantry crane, (7) bridge approach.
are used as provisional prestressing to hold the
segments in position until permanent prestress
tendons can be threaded into the ducts and
stressed.
The symmetrical box girder main structure
consists of a 254 ft (77.5 m) side span, a 312 ft (95
m) side sta ye d span, and a 33 1 ft (101 m) section of
sta ye d center span cantilevering toward the center
of the bridge. The center section between the
stayed cantilever ends is made up of four 213 ft (65
m) suspended lightweight concrete girders.
Two alternatives were considered for the cablestay pylons: a single pylon located on the longitudinal centerline of the bridge or a portal-type
pylon. To simplify the project, the portal-type
pylon was selected. The portal pylon is fixed to the
pier and passes freely through the superstructure,
Figure 9.34. The superstructure is fixed at the
pylon piers except for rotation. It is allowed to
move longitudinally at succeeding piers.
Two alternatives were also considered for the
stay system: a multiple stay system supporting the
deck almost continuously and a system consisting
of a few large stays. As prestressed concrete stays
had been selected, the second solution became
somewhat mandatory. Construction of prestressed
FIGURE
9.33.
FIGURE
9.34.
Free passage of pylon through deck.
concrete stays is a costly operation requiring extensive high scaffolding, Figure 9.35; thus it is advantageous to reduce the number of stays.
The short stays of the bridge have a slope of 1: 1
and the long stays a slope of 1:2. Their points of
anchorage to the deck are respectively at 156 ft
(47.5 m) and 3 12 ft (95 m) on both sides of a pylon.
The long stays have a cross section of 3 by 3.3 ft
(0.9 by 1 .O m) and are prestressed by 36 tendons on
the bank side and by 40 tendons on the river side,
because of the larger load on that side, Figure
9.36a. The effect of the different loads on the stays
introduces a flexural moment into the pylon. The
short stays have a cross section of 2.13 by 3.3 ft
(0.65 by 1.0 m) and are prestressed by 16 tendons
Placing of segments by gantry crane.
416
Concrete Segmental Cable-Stayed Bridges
Three loading conditions were considered for
the stays from a statics point of view:
FIGURE 9.35. Falsework fbr stay construction.
on the bank side and 20 tendons on the river side,
Figure 9.366.
The concrete of the stays has a 2%day strength
of approximately 8700 psi (60 MPa). Its function is
not only to protect the tendons, but also to increase
the rigidity of the stays, which is four times that of
the tendons alone.
Long stays
40/36 cables
E?!l
t
Short stays
20/16 cables
65
8,:
(b)
I
FIGURE 9.36. Cross section of stays.
1.
For the self-weight of the stays and dead load
of the superstructure, the deck is considered as
supported on nonyielding supports, which are
the stay anchorage points, and the load in the
stays results from the reactions at these points.
2.
For design live load, the deck is considered as
supported on yielding supports, the rigidity of
which is determined by the rigidity of the prestressed stays.
3.
The prestress of the stays was calculated with a
safety factor against cracking of 1 .l for dead
load and 1.3 for live load, without allowing any
tension in the concrete. The ultimate load
safety factor is 1.8. For the load condition between cracking and collapse the stay rigidity is
reduced to the rigidity of the tendons alone.
Their exc essiv e elo ng atio n, in c ase they
yielded, would lead to an excessive deflection
of the box girder and a premature collapse before the proposed safety limit. Therefore, it
was necessary to reduce the initial stress of the
tendons to 40 to 45% of their ultimate strength
in order to keep them in the elastic range up to
ultimate load determined by the safety factor
of the structure as a whole.
The sag of the long stay is 2.3 ft (0.70 m) in a
length of 328 ft (100 m) under dead load. Under
live load the sag is reduced to 1.8 ft (0.55 m). The
cross section of the stays at their extremities is increased slightly to resist bending stresses. These
stresses were calculated by the method of finite differences.
In the longitudinal direction the girders are
prestressed primarily by the horizontal components
of the stay forces. The unstayed end spans are prestressed w ith 54 tendons. In the other spans additional prestressing is provided by 10 tendons that
overlap each other at the supports. These tendons
were required until such time as the stay forces
were applied and, at completion, to provide safety
against cracking and collapse. The deck slab is prestressed transversely by tendons spaced at 12 to 17
in. (0.30 to 0.44 m).
The suspended 213 ft (65 m) span is composed
of four precast lightweight concrete girders with a
6500 psi (45 MPa) concrete. The cast-in-place deck
slab is increased from a thickness of 9.8 in. (250
mm) in the box girders to 12.6 in. (320 mm), owing
to the smaller restraint of the slab in the one web
girders.
417
Tie1 Bridge, The Netherlands
The following restraints and conditions were
considered in the determination of the construction procedure for the main spans of the structure:
1.
PH A SE 1
The exclusion of falsework from the river because of ndvigation requirements.
2. The potential for flooding.
3. The presence of the precasting plant on the
north bank.
4. The possibility of adjusting the attachment
points of the stay to the deck.
Construction was executed in increments limited
by the attachment points of the stays to the deck.
The stays were prestressed progressively, by increasing the number of stressed tendons as the
load in the stays increased. However, during certain construction phases when the load in the stays
decreased, some of the tendons were detensioned
or slacked off.
Using the north side (access viaduct side) as an
example, the construction was divided into the
following phases, Figure 9.37:
Phme 1:
Construction oj the outer spans-that is, the
stay-supported side span andjanking span
a.
b.
c.
Superstructure from pier 11 to pier
12 and a 72 ft (22 m) cantilever into
the next span
Extension up to temporary support
12A
Extension up to pier 13 with a 26 ft
(8 m) cantilever into the center span;
simultaneous construction of the
pylon
Phase 2:
Construction qf the$rst section over the river
and the shortfbrestay.
Phase 3: Construction of the second section ouer the
ri-iw a11d the long,fowstay.
The external spans on the north side were constructed on falsework during the dry season.
Utilizing the precast plant on the north side, precast segments 16.7 ft (5.10 m) long weighing 132
tons (120 mt) were assembled on the falsework.
Segments were joined by f in. (5 mm) cast-in-place
joints. Placing of the segments was carried out by
the same gantry crane as for the access viaduct. On
the south bank, where there was no precasting
plant, the external spans were cast in place on
falsew o rk.
The cantilever river spans were built on 157 ft
(48 m) long steel falsework, consisting of four 10 ft
PHASES DE
L’OUVRAGE
CONSTRUCTION
PRINCIPAL
DE
MAIN BRIDGE
LONSTRUCTION P H A S E S
FIGURE 9.37. Main bridge construction phases.
(3 m) deep girders on 23 ft (7.10 m) centers. This
falsework was suspended at one end by prestressing strands from the top of the pylons. At the
lower end, the temporary support strands were anchored in a cross beam that supported the steel
falsework by four 350 ton (315 mt) jacks. The 3 ft
(1 .O m) stroke of the jacks allowed adjustment of
the level of the suspension points, and the jacks
were used also to release the temporary prestress
suspension strands w hen the final stays w ere installed. At the opposite end, the steel falsework was
hinged. The horizontal force component on these
hinges was transmitted directly to the completed
part of the deck, and the vertical component was
taken by 1 in. (26 mm) bars.
In Phase 3, the temporary stays were deflected
by means of 95 ft (29 m) booms. This provided the
advantage of maintaining the angles at the lower
connection equal to that of Phase 2 and keeping
approximately the same force level in the temporary stay.
The falsew ork used in Phases 2 and 3 w as carried on a barge; it was positioned by two derricks
located on the completed part of the deck and by a
floating crane. After the box girders were cast, the
level of the falsework was adjusted, the last joint
418
Concrete
Segmental
was cast, and the concrete was prestressed. The
next steps were constructing the stays, prestressing
them, releasing the temporary stays, and removing
the falsework.
In order to reduce creep and shrinkage, the
stays were made of 17 ft (5.15 m) long segments
with protruding reinforcement and 16 in. (0.4 m)
cast-in-place joints. The building of the falsework
for the stays and the handling of the precast segments were carried out with the help of a 16 ton
(15 mt) tower crane 2 13 ft (65 m) high, running on
the deck.
The precast 213 ft (65 m) suspended span girders weighed 468 tons (425 mt) and were transported by barge.
9.7 Pasco-Kennewick Bridge, U.S.A.
The first cable-stayed bridge with a segmental concrete superstructure to be constructed in the
United States is the Pasco-Kennewick Intercity
Bridge crossing the Columbia River in the state of
Washington, Figure 9.38. Construction began in
August 1975 and was completed in May 1978. The
overall length of this structure is 2503 ft (763 m).
The center cable-stayed span is 981 ft (299 m), and
the stayed flanking spans are 406.5 ft (124 m). The
Pasco approach is a single span of 126 ft (38.4 m),
FIGURE 9.38. Pasco-Kennewick
(courtesy of Arvid Grant).
Intercity
Bridge
Cable-Stayed
Bridges
while the Kennewick approach is one span at 124 ft
(37.8 m) and three spans at 148 ft (45.1 m).4*15*18,1s
The girder is continuous without expansion
joints from abutment to abutment, being fixed at
the Pasco (north) end and having an expansion
joint at the Kennewick (south) abutment. The concrete bridge girder is of uniform cross section, of
constant 7 ft (2 m) depth along its entire length and
79 ft 10 in. (24.3 m) width. The shallow girder and
the long main spans are necessary in order to reduce roadway grades to a minimum, to provide the
greatest possible navigation clearance below, and
to reduce the number of piers in the 70 ft (21.3 m)
deep river.
The bridge is not symmetrical. The Pasco pylon
is approximately 6 ft (1.8 m) shorter than the Kennewick pylon, and the girder has a 2000 ft (610 m)
vertical curve that is not symmetrical with the main
span. Therefore, the cable-stay pairs are not of
equal length, the longest being 506.43 ft (154 m).‘s
There is no attachment of the girder at the pylons, except for vertical neoprene-teflon bearings
to accommodate transverse loads. The girder is
supported only by the stay cables. There are, of
course, vertical bearings at the approach piers and
abutments. It is estimated that the natural frequency of the girder, where it will respond to
dynamic acceleration (i.e., earthquake), is 2 cycles
per second. If the situation occurs where the longitudinal acceleration exceeds this value, the vertical restraint at the Pasco (north) abutment is designed to fail in direct shear, thus changing the
structure frequency to 0.1 cycles per second, which
renders the system insensitive to dynamic excitation. The three main spans were assembled from
precast, prestressed concrete segments, while the
approach spans were cast in place on falsework,
Figures 9.39 and 9.40.
Deck segments were precast about 2 miles (3.2
km) downstream from the bridge site. Each segment weighs about 300 tons (272 mt) and is 27 ft
(8.2 m) long, Figure 9.41. The segment has an 8 in.
(0.2 m) thick roadway slab, supported by 9 in. (0.22
m) thick transverse beams on 9 ft (2.7 m) centers,
and is joined along the exterior girder edges by a
triangular box which serves the function of cable
anchorage stress distribution through the girder
body, Figure 9.42. 6 Each match-cast segment required approximately 145 yd3 (11 lm3) of concrete,
continuously placed in a previously adopted sequence within six hours. After initial curing in the
forms, the girder segments were wet cured for two
weeks in the storage yard, air cured for an additional six months, prestressed transversely,
Brotonne
F I G U R E 9 . 3 9 . l’r~w~-K~~~~w~ I& Intu Lit\ HI ldge,
precast segments in main spans (courtesy of Arvid
Grant).
FIGURE 9.40. Paaco-Kenne\\ic 1, Intel city Bridge, appreach spans cast in place on falsework (courtesy of
Walter Bryant, FHWA Region 10).
Bridge,
France
419
cleaned, repaired, completed, loaded on a barge,
and transported to the structure site for installation in their final location. For possible unpredicted
developments a shimming process was held in reserve for maintaining the assembled girder
geometry correctness, but it was not used. There
are no shims in the segmentally assembled,
epoxy-joined prestressed concrete girder.‘“+18*‘”
The sections were barged directly beneath their
place in the bridge and hoisted into position, Figure
9.43. Fifty-eight precast bridge girder segments
were required for the project.
The stays are arranged in two parallel planes
with 72 stays in each plane-that is, 18 stays on
each side of a pylon in each plane. They are held at
each p y l o n top, 180 ft (55 m) above the bridge
roadway, in a steel weldment, Figure 9.44. Stay anchorages in the bridge deck are spaced at 27 ft (8.2
m) to correspond with the segment length. The
stays are composed of + in. (6 mm) diameter parallel
high-strength steel wires of the BBR type. The
prefabricated stays, manufactured by The Preston
Corporation, arrived on the job site on reels, Figure 9.45, and contained from 73 to 283 wires, depending upon their location in the structure. They
were covered with a # in. (10 mm) thick polyethylene pipe, and after installation and final
adjustment were protected against corrosion by
pressure-injected cement grout. The outside diameter of the pipe covering varies from 5 to 7 in.
(0.12 to 0.17 m). Design stress level for the stays is
109 ksi (751.5 MPa). Stay anchorages are of the
epoxy-steel ball (HiAmp) fatigue type produced by
The Preston Corporation.
This structure was designed by Arvid Grant and
Associates, Inc., of Olympia, Washington, in professional collaboration with Leonhardt and Andra
of Stuttgart, Germany.
9.8 Brotonne Bridge, France
FIGURE 9.41. Pasco-Kennewick Intercity Bridge,
precast segments in casting yard (courtesy of Arvid
Grant).
The Pont de Brotonne, designed and built by
Campenon Bernard of Paris, crosses the Seine
River downstream from Rouen in France. Because
of increased navigation traffic in the area, a second
crossing over the Seine River was urgently needed
between the two harbors of Le Havre and Rouen.
The first one, the steel suspension bridge of Tancarville, was opened to traffic in 1959. The second,
the Brotonne Bridge, the world’s largest cablestayed prestressed concrete bridge, was opened to
traffic in June 1977. 2o A model of the structure is
I
22.50 m
I-
cl
CROSS
-SECTION
OF
7
CONCRETE
BRIDGE
/
eoprene sleeve
SECTION. ELEVATION B - B
FIGURE 9.42. Pasco-Kennewick Intercity
cables (courtesy of Prof. Fritz Leonhardt).
FIGURE 9.43. Pasco-Kennewick Intercity Bridge,
erection of precast segments from barge (courtesy of
Arvid Grant).
SECTION A - A
Rridge, cross section and anchorage of sta?
FIGURE 9.44. I’asco-Kennewick
lntercity Bridge,
pylon and stay attachment steel weldment at top (courtesy of Arvid Grant).
Bro to nne Bridge, France
421
Bridge,
FIGURE 9.47. Artist’s rendering of the Pont de
Bro to nne.
shown in Figure 9.46 and the general layout in Figures 9.47 and 9.48. The box girder carries four
lanes and replaces ferry service between two major
highways that run north and south of the Seine.
Because large ships use this section of the river to
approach the inland port of Rouen 22 miles (35
km) to the east, vertical navigation clearance is
164 ft (50 m) above water level, which results in a
6.5% grade for its longer approach.15*21
Total length of structure is 4194 ft (1,278.4 m),
consisting of the main bridge and two approach
viaducts. The main crossing has a span of 1050 ft
(320 m). On the right bank, the transition between
the main span and the ground is quite short because of a favorable topography where limestone
strata slope upward to a relatively steep cliff. On
the left bank, the terrain is flat and occupied by
meadows. With an allowable maximum grade of
6.5% and a maximum height of fill of 50 ft (15 m),
a nine-span viaduct was required to reach the main
bridge. In a structural sense, the bridge is divided
into tw o sections separated by an expansion joint at
a point of contraflexure in the left-bank viaduct
span adjacent to the cable-stayed side span, Figure
9.48.‘O
The prestressed segmental concrete deck consists of a single-cell trapezoidal box girder with
interior stiffening struts, Figures 9.49 and 9.50. In
the approach spans, web thickness is increased
from 8 in. (200 mm) to 16 in. (400 mm) near the
piers, and the bottom flange thickness is increased
to a maximum thickness of 17 in. (430 mm). The
only portion of the segment that was precast is its
sloping webs, Figure 9.51, which were precast at
the site. The other portions of the cross section,
including top and bottom flanges, interior stiffening struts, and cable-stay anchorages (in the main
structure only), were cast in place. Each segment is
9.8 ft (3 m) long.
Extensive use of prestressing was made in the
deck to provide adequate strength to this light
structure. To resist the extreme shear stresses it
was decided to place vertical prestressing in the
webs. Pretensioned units were stressed on a casting
bed, Figure 9.52, and equipped with specially designed button heads, thus producing a combination of pretensioning and anchorage plates. This
system has the advantage of ensuring a perfect
centering of the prestressing force together with a
very rapid transfer of this force at both ends. Intensive rupture tests proved that an extremely
high resistance to shear was created by this system.20
Finally, prestressing w as also used as follow s,
Figure 9.53:20
FIGURE 9.45. l’asco-E;enne~~,ick
prefr~bricated cable stay on reel.
FIGURE 9.46.
Intcrcity
Model of the Pant dc Rrotonne.
1. Transversely in the top flange to provide
flexural strength to the thin 8 in. (200 mm)
slab.
2. In the inclined internal stiffeners, to accommodate tensile forces created by the transfer
of loads from the box girder to the stays.
-+s_s_sO,s8>~_+
5850
1
5850
1 5850
4
5850
1
5850
i
5850
(5850
i
32000
69750
127840
FIGURE 9.48. General layout of Brotonne Bridge.
_ _ _ _ _
sl
b
-’,, 1.50
(5’) .,
6.50
t
I
I- 1.60
(5’) . 1.60
(5’) _,
6.50
1 1.50
(5’) *
I
I
5.60
1
4.00
(18’)
I
4.00
(13’)
FIGURE
9.49.
1
5.60
II
(18’)
(13’)
Cross section of Brotonne Bridge.
FIGURE 9.50. Interior view of deck, Brotonne
Bridge.
FIGURE
coupler
for
9.51.
Precast webs, Hrotonnc
tensioning jack
36 am dia tension
Dyuidag -mat -..
FIGURE 9.52. Casting bed for pretensioned webs.
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFE
131 It lgC.
distribution beam
FIGURE
9.53.
Various prestressing systems in the box girder.
3. Transversely in the bottom flange, to counteract tensile forces created by the stiffeners.
4. Longitudinally near the center of the main
span, to allow for a reasonable margin of the
order of 300 psi (2 MPa) of compressive stress
in view of creep and secondary tensile stresses.
Befo re erectio n o f the sup erstructure, the
bridge’s 12 approach piers were slip-formed, nine
on the left bank and three on the right. The pier
shafts have an octagonal curvilinear cross section
inscribed inside a 13 by 29 ft (4.0 by 8.75 m) rectangle, Figure 9.54. The same section was used for
all the approach-span piers, whose height varied
from 40 to 160 ft (12 to 49 m). The shape of the
piers did not substantially increase costs but did increase the aesthetic appeal of the piers. The piers
bear through a reinforced concrete footing on four
rectangular slurry trench walls used as piles with a
maximum length of 60 ft (18 m), Figure 5.17.
The pylon pier shafts also have an octagonal
curvilinear shape inscribed inside a 30 ft (9.2 m)
square to produce equal bending resistance about
both principal axes. They are supported on foundation shafts having a diameter of 35 ft (10.86 m)
with a maximum wall thickness of 6 ft 8 in. (2.03
m). The foundation shafts transfer the loads to a
limestone stratum at a depth of 115 ft (35 m) below
ground level. Foundation shafts were built inside
a circular slurry trench w all, w hich w as used as a
cofferdam for dewatering.*”
When slip-forming of the piers reached deck
level, the piers were prestressed to their foundation so as to stabilize them for erection of the deck
segments. As the precast deck units were erected,
!I/
*J
.
FIGURE 9.54. Pier and foundation of approach
spans.
wl
c
1.50 ,,
*I
(5’)
6.50
,
T
_
1. 6 0
,.
(5’) -
I
1. 6 0
A
(5’)
6 .5 0
(21’)
ROADWAY
;
1
1. 5 0
(d
I
FIGURE
FIGURE
9.50.
In ter ior
view
of
9.49.
d eck ,
Cross section of Brotonne Bridge.
B r oton n e
FIGURE
9.51. Precast webs, Brotonnc
Hr~tlgc.zyxwvutsrqponmlkjihgfedcba
Bridge.
bulkhead
. . ___-_- - steel
i. forma
-.
_-.--
coupler for
., tcnaion rods
___i..
36
tensioning - jack
__.a-__--_
.--._
am dia
tendon .rodr;
_ ._--..
dlatributiw
,-- ___.i- :’
-
bean
adjustable
_-bracketa
11-36 IUS Rw idag
--...--tenrion bars
FIGURE
i preatrcrsin~
_ _ _ __tendons
__ ._-._
-- soffit
..__
-
9.52. Casting bed for pretensioned webs.
FIGURE 9.53.
Various prestressing systems in the box girder.
3. Transversely in the bottom flange, to counteract tensile forces created by the stiffeners.
4. Longitudinally near the center of the main
span, to allow for a reasonable margin of the
order of 300 psi (2 MPa) of compressive stress
in view of creep and secondary tensile stresses.
Befo re erectio n o f the sup erstructure, the
bridge’s 12 approach piers were slip-formed, nine
on the left bank and three on the right. The pier
shafts have an octagonal curvilinear cross section
inscribed inside a 13 by 29 ft (4.0 by 8.75 m) rectangle, Figure 9.54. The same section was used for
all the approach-span piers, whose height varied
from 40 to 160 ft ( 12 to 49 m). The shape of the
piers did not substantially increase costs but did increase the aesthetic appeal of the piers. The piers
bear through a reinforced concrete footing on four
rectangular slurry trench walls used as piles with a
maximum length of 60 ft (18 m), Figure 5.17.
The pylon pier shafts also have an octagonal
curvilinear shape inscribed inside a 30 ft (9.2 m)
square to produce equal bending resistance about
both principal axes. They are supported on foundation shafts having a diameter of 35 ft (10.86 m)
with a maximum wall thickness of 6 ft 8 in. (2.03
m). The foundation shafts transfer the loads to a
limestone stratum at a depth of 115 ft (35 m) below
ground level. Foundation shafts were built inside
a circular slurry trench w all, w hich w as used as a
cofferdam for dewatering.2u
When slip-forming of the piers reached deck
level, the piers were prestressed to their foundation so as to stabilize them for erection of the deck
segments. As the precast deck units were erected,
FIGURE 9.54. Pier and foundation of approach
spans.
Bccticm D-D
Lalgitullim1bctic0A-AzyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
FIGURE 9.55. Half center span and pylon.
longitudinal
cocaectioaktweea
rection
Pyloncrdl Plcr
------c
I
c
I
__ . . . _ - 2
*
l
w
O
.
N
i
._
” +
FIGURE 9.56.
.i.
-
.._..
Pzo.
Connection between pylon, deck, and pier.
_ . . -. . . . .
the pylon was constructed by conventional
methods.
Two single-shaft pylons carry a system of 21
stays located on the longitudinal axis of the structure, Figure 9.55. The reinforced concrete pylons
required limited cross-sectional dimensions to preclude an unnecessary increase of the deck width
while providing sufficient dimension to accommodate bending stresses from a transverse wind direction. Total pylon height above the deck is 23 1 ft
(70.5 m). Construction of the pylon required Ieapfrog forms with 10 ft (3 m) lifts. An interesting
feature is the total fixity of the pylon with the box
girder deck. Because the bending capacity of the
pylon pier and foundation had to be such as to accommodate unsymmetrical loads due to the cantilever construction, a decision was made to take
advantage of this requirement in the final structure
to reduce the effect of live load in the deck. Therefore, the pylon was constructed integral with the
deck at its base, both pylon and deck being separated from the pier by a ring of neoprene bearings,
Figure 9.56.20
FIGURE 9.57. C.;;rl,lc-SI;I>
mhor;1gc.
All deck loads are carried to the pylon piers by
21 stays on each pylon. Each stay consists of 39 to
60-0.6 in. (15 mm) strands encased in a steel pipe,
which is grouted after final tensioning. Stay length
varies from 275 to 1115 ft (84 to 340 m). Anchorage spacing of the stays at deck level is every 19.7 ft
(6 m), every other segment, where the inclined stiffeners in the deck segments converge, Figures 9.53
and 9.57. A special deck anchorage block was designed to accommodate the variable number of
strands in the stay as well as to allow full adjustment of the tension in the stays by a simple anchoring nut, Figure 9.58. The anchorage of the
stays is such that it is possible at any time during
the life of the structure to either readjust the tension in the stay or replace it without interrupting
traffic on the bridge. Permanent jacks are incorporated into the anchorage, Figure 9.59, such that by
tensioning the stay the adjusting nut can be sla c ke d
off. Stays are continuous through the pylon
where they transfer load to the pylon by a steel
saddle. The pipe wall thickness is increased near
the anchorage points and near the pylon so as to
improve fatigue resistance of the stays with regard
to bending reversaIs.20
In constructing the deck girder, the operation
was to extend the bottom flange form from a
traveling form at the completed segment, placing
the precast web units that form the basic shape and
act as a guide for the remaining traveling form.
After placement of the precast webs the interior
steel form was jacked forward to cast the bottom
flange struts and the top flange. Tower cranes at
the pylon placed, as far as they could reach in both
directions, the precast webs, Figure 9.60. Beyond
the range of the tower cranes, gantry cranes running on rails on the top flange and extending 9.8 ft
FIGURE 9.58. Jacking of stay.
427
Danube Canal Bridge, Austria
FIGURE 9.59. Permanent stay anchorage.
FIGURE 9.60. llZain pier, pylon, anti deck during
construction, from reference 20.
(3 m) beyond the end of the completed section
were used to place new elements.
The structure is shown at the start of main span
construction in Figure 9.61, before closure of the
main span in Figure 9.62, and completed in Figure
9.63.20
FIGURE 9.6 1.
reference 20.
Stnrt of mnill \pdn Loll$tructiotl.
FIGURE 9.62.
erence 20
&fore closure of mn111 SP;III, f 1oln Ed-
FIGURE 9.63. Aerial view of the Hroronne
from reference 20.
9.9
flom
Hrldge,
Danube Canal Bridge, Austria
This structure is located on the West Motorway
(Vienna Airport Motorway) and crosses the
Danube Canal at a skew of 45” . It has a 390 ft (119
m) center span and 182.7 ft (55.7 m) side spans,
Concrete Segmental Cable-Stayed Bridges
428
182.7 ft
*
55.7 m
FIGURE 9.64.
canal.
182.7 ft
110m
55.7 m
I
Elevation of the Danube Canal Bridge.
Figure 9.64. It is unique because of its construction
technique. Because construction was not allowed to
interfere with navigation on the canal, the structure was built in two 360.8 ft (110 m) halves on
each bank and parallel to the canal, Figure 9.65.
Upon completion the two halves were rotated into
FIGURE 9.65.
390 ft
Construction of half-bridge on bank of
final position and a cast-in-place closure joint w as
made, Figures 9.66 through 9.69. In other words,
each half w as constructed as a one-time sw ing span.
The bridge superstructure is a 5 1.8 ft (15.8 m)
wide trapezoidal three-cell box girder, Figure 9.70.
The central box was cast in 25 ft (7.6 m) long segments on falsework, Figure 9.7 1. After the precast
inclined web segments were placed, Figure 9.72,
the top slab w as cast.
Each half-structure has two cantilever pylons
fixed in a heavily prestressed trapezoidal crosshead
protruding under the deck with a two-point bearing on the pier, Figure 9.73. At the deck level the
stays attach to steel brackets connected to prestressed crossbeams, Figures 9.74 and 9.75.
Each stay consists of eight cables, two horizontal
by four vertical. At the top of the pylons each cable
is seated in a cast-iron saddle. The cable saddles are
stacked four high, Figure 9.76, and are fixed to
each other as well as to those in the adjacent plane.
The cables were first laid out on the deck, fixed to a
saddle, and then lifted by a crane for placement at
the top of the pylon. The cables were then pulled
FIGURE 9.66. Plan of Danube Canal Bridge during construction and final state.
Danube Canal Bridge, Austria
zy
429
FIGURE 9.67. Danube Canal Bridge during rotation.
FIGURE 9.69. Closure joint, Danube Canal Bridge.
FIGURE
9.68.
Ihnubc Canal Bridge during rotation.
FIGURE 9.70. Cross section, Danube Canal Bridge.
at each extremity by a winch rope to their attachment point at the deck level.
During rotation of the two half-bridges, the deck
and pylon sat on a bearing consisting of five
epoxy-glued circular steel plates. The top plate was
coated with teflon, sitting in turn on a reinforced
concrete block that sat on a sand box. After rota-
tion the structure was lowered to permanent
bearings by emptying the sand box.
At the canal-bank end the deck had a concrete
wall on its underside, bearing on a circular conCrete sliding track, Figure 9.77. The bearing between the wall and the track was effected by two
concrete blocks clad with steel plates, under which
430
Concrete Segmental Cable-Stayed Bridges
FIGURE 9.71. Con~tl
llc tio tl O II
htnk, I),~nubc
C,~nnl
Bridge.
FIGURE 9.73. ‘Trapezoidal crosshead, Danube Canal
Bridge.
FIGURE 9.72. P~t.c,~st
\jcbr, Danube Canal Bridge.
teflon-coated neoprene pads were introduced
during the rotation movement (similar to the incremental launching method). The pivoting was accomplished by means of a jack pulling on a cable
anchored in a block located near the sliding-track
end.
After rotation the two halves of the structure
were connected by a cast-in-place closure joint, and
continuity tendons were placed and stressed.22
The
final structure is shown in Figure 9.78.
9.10
9.10.1
Notable Examples of Concepts
PROPOSED GREAT BELT BRIDGE,
DENMARK
The competition for a suitable bridge design in
Denmark produced many new concepts and architectural styles. The design requirements specified three lanes for vehicular traffic in each direction and a single railway line in each direction.
FIGURE
9.74. Jacking of stays, Danube Canal Rridge.
The rail traffic was based on speeds of 100 mph
(161 km/hr).23 Navigational requirements stipulated that the bridge deck be 220 ft (67 m) above
water level, and the clear width of the channel was
to be 1130 ft (345 m).
A third prize winner in this competition was the
Morandi-style design proposed by the English consulting firm of White Young and Partners, Figure
Notable Examples of Concepts
431
FIGURE 9.75. Cable-stay attachment, Danube Canal
Bridge.
FIGURE 9.77. Circular concrete sliding track, Danube Canal Bridge.
FIGURE 9.78. Completed Danube Canal Bridge.
FIGURE 9.76. Stay saddles at pylon, Danube Canal
Bridge.
9.7. This design embodied the principles of a
cable-stayed bridge combined with conventional
approaches of girders and piers w ith normal spans.
The principal feature of this bridge design is the
three-plane alignment of cable stays. This feature
may become more important in urban areas, where
trends in the future may dictate multimodal transportation requirements and an increase in the
number of automobile traffic lanes. The deck consists of two parallel single-cell prestressed concrete
box’girder segments, Figure 9.79. The rail traffic is
supported within the box on the bottom flange
and the road traffic is carried on the surface of the
top flange.
The box girder contemplated a depth of 23.5 ft
(7.2 m) and width of 27.75 ft (8.45 m) with the
top flange cantilevered out 12 ft (3.7 m) on each
side. The piers and towers were to be cast-in-place
construction to support the deck segments, which
were to be precast at various locations on shore and
flo ated to the brid g e site fo r erectio n. The
maximum weight of a single box segment was estimated at 2200 tons (2000 mt). All segments of
the superstructure were to be of reinforced and
prestressed concrete.
Up to this point in time, when the competition
for this structt.re was conducted, all the concrete
cable-stayed bridges had been either designed by
432
Concrete Segmental Cable-Stayed Bridges
-e--- 3.675 m
8.45 m
b ’<
Morandi (Lake Maracaibo, Wadi Kuf, and so on)
o r stro ng ly influenc ed b y his sty le (Chaco/
Corrientes). They were typified, for the most part,
by the transverse A-frame pylon with auxiliary
X-frame support for the girder. However, an entry
in the Danish Great Belt Competition by Ulrich
Finsterwalder of the German firm of Dyckerhoff
& Widmann deviated from this style and was
awarded a second prize.
Finsterwalder proposed a multiple span, multistay system using Dywidag bars for the stays, Figure 9.10. This proposal contemplated a spacing
between pylons of 1148 ft (350 m) and a spacing of
the stays at deck level of 32.8 ft (10 m). Pylon
height above water level was 520 ft (158.5 m). In a
transverse cross section the deck was 146 (44.5
m) wide with two centrally located vertical stay
planes 39 ft 4 in. (12 m) apart to accommodate the
two rail traffic lanes, and three automobile traffic
lanes in each direction outboard of the stay planes,
Figure 9.80.
The solid concrete deck had a thickness of 3 ft
(0.9 m) in the transverse center portion, under the
rail traffic, and tapered to a 1.3 ft (0.4 m) thickness
-k! 58,M
22,50
f 0.00
i
175.w-j-350,oo / 175,oo
-J
‘,.
-_-_-_--------_-__- _______r____
/
r _ _ . r ____ - -----‘-‘-,
-------,
r ________,/__ ------
I
‘-----:- -.---7 --__ -_,_ _ _
Bahn
j
1
Ir;
lu~+~
!
n
1
L.
-q-i-’
(
~
‘C
sam&
..I’ !
:o,llo
’ ON ----______ _
- - - - - - - - - - ------_-k
15,25 ------&2,00?c--- 10,00------+2,00MO
15,25 -
-
-
FIGURE 9.80. Danish Great Belt Bridge, elevation and cross section (coultesv
Dyckerhof’f’ & Widmann).
of
;p2,50
Notable Examples of Concepts
at the edges. The deck was to be constructed by the
cast-in-place
balanced
cantilever
segmental
method, each segment being supported by a set of
S&F s.
The proposed Dame Point Bridge over the St.
Johns River in Jacksonville, Florida, as designed
b) the firm o f H o w ard N eed les Tammen &
Bergendoff, is a cable-staved structure with a concrete and a steel alternative. .4n artist’s rendering
of the concrete cable-staved bridge alternative is
show n in Figure 9.81. Navigation requirements
dictate a 1250 ft (381 m) minimum horizontal
opening and a vertical clearance of 152 ft (46.3 m)
above mean high water at the centerline of the
clear opening. I‘he proposed concrete cable-stayed
main structure w ill have a 1300 ft (396 m) central
FIGURE 9.81. lhmc Po int l3ritigc , artist’s rendering
(cotll-tc’sv of’ Ho\vxcl Needles I‘a~nrncn ,Y- Bergendoff).
433
span w ith 650 ft (198 m) flanking spans. The layout
of the main structure is shown in Figure 9.82.‘ 4
Structural arrangement of the bridge deck is
shown in Figure 9.83. The bridge deck, which will
carry three lanes of traffic in each direction, will
span between longitudinal edge girders on each
side. The longitudinal edge girder is in turn supported by a vertical plane of stays arranged in a
harp configuration. The concrete deck and edge
girders take local and overall bending from dead
and live load in addition to the horizontal thrust
from the stavs.25 The stav cables are anchored in
massive vertical concrete pvlons, two at each main
pier, which carry all loads to the foundations, Figure 9.84.
In the center span, at each edge of the deck, the
stavs are in a single plane spaced 30 in. (0.76 m)
vertically, Figures 9.84 and 9.85. Stavs in the side
spans, along each edge, are in tw o planes spaced 30
in. (0.76 m) transversely. Spacing of pairs of stavs
along the edge beam is approximately 30 ft (9.1 m).
Preliminary design contemplates 7 to 9 Dlwidag
bars per stay, li in. (31.75 mm) in diameter, the
number of bars per stay being a function of stress
in the stay. The Dywidag bars are to be encased in a
metal duct. During erection the fabricated length
of duct is left uncoupled. After final adjustment
the lengths of duct are coupled and pressuregrouted. Thus, the steel encasing tube will then be
composite for live load and secondary dead load.‘”
Construction proceeds bv conventional methods
from the top of the pier bases at elevation 15.0 ft
(4.6 m) to the level of the roadway at elevation
144.6 ft (44 m). At this point, a fixed formtable is
secured and the first elements of the pylon and
edge girders are cast. Erection of the deck is bv the
FIGURE 9.82. Dame Point Bridge, concrete cable-stayed alternative, from reference
23 (courtesy of Howard Needles ‘l‘ammen & Bergendoff).
105’-10” (32 3m)
.Pr ecast
Cast -in-sit u
zyxwvu
T Be a m
Beam
FIGURE 9.83. Dame Point Bridge, structural arrangement of bridge deck. from
reference 24 (cowtesy of Howard Needles I‘ammen & Bergendoff).
Tower -.,
SIDE V I E W
FR ON T
V I EW
FIGURE 9.84. Dame Point Bridge, pylon arrangment, from reference 24
(courtesv of Howard Needles Tammen & Bergendoff).
434
;OMETRIC VIEW OF ERECTION SEQUENCE
x
*’
,’
15,
/
/’
,?zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
/ zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
<,,de’,
15
PLATE 17
436
Co ncrete Segmental Cable- Stay ed Bridges
balanced cantilever method. Two pairs of traveling
forms are then used for sequential casting of 17.5
ft (5.3 m) lengths of edge girders on each side of
the pylon. The bridge deck consists of single-T
precast floor beams spanning between longitudinal edge girders and a cast-in-place topping. The
precast T’s are pretensioned for erection loads.
After erection the entire deck is post-tensioned to
provide positive precompression between edge
girders under all conditions of loading, Figure
9.85.24*25
A hinge expansion joint is provided at the centerline of the main span to allow for changes of
superstructure length due to temperature, creep,
and shrinkage. Similar joints are provided at the
end piers, and link connections are used to prevent
vertical movement of the superstructure.
9.10.3 PROPOSED RUCK- A- M UCKY
BRIDGE, U.S.A.
The site for the proposed Ruck-A-Chucky Bridge
designed by T. Y. Lin International, Figure 9.86, is
approximately 10 miles (16 km) north of the proposed Auburn Dam and about 35 miles (56 km)
northeast of Sacramento, California, crossing the
middle fork of the American River. The river at
this location is about 30 ft (9 m) deep and 100 ft
(30.5 m) wide; however, upon impounding of the
water behind the proposed dam, the river will become 450 ft (137 m) deep and 1100 ft (335 m)
wide.26
In order to provide a 50 ft (15 m) vertical clearance above high reservoir water level, a bridge
length of 1300 ft (396 m) will be required between
the hillsides, w hich rise at a 40” angle from the
horizontal. Two existing roads parallel the canyon
faces; a straight bridge across the river would require extensive cuts into the rock faces of the canyon to provide the necessary turning radius at the
bridge approaches. This would be not only expensive but would also be damaging to the environment.
Conventional piers in the river provide prohibitive
design constraints, not only because of the 450 ft
(137 m) water depth, but also because of the seismicity of the area. The hydroseismic (seiche effect)
forces provide a formidable design load.
After extensive studies, the proposed final solution w as that of a hanging arc, Figures 9.87 and
9.88. The geometric configuration of this structure
is such that the stays are tensioned to control the
stresses and strains, in order to balance all the dead
load with zero deflection; the curved girder carries
the traffic and absorbs the horizontal component
of the stays as axial compression. The stays are anchored on the slope according to the design formation to control the line of pressure in the girder.
Thus, an ideal stress condition is achieved with
almost no bending or torsional moments. After
numerous studies and trade-offs a final radius of
curvature was selected at 1500 ft (457 m).26
Two alternative designs have been prepared for
this structure, one with a steel box girder and one
with a lightweight concrete box girder. The concrete box girder, Figure 9.89, is fixed at the abutments and has no hinges or expansion joints in the
1300 ft (396 m) span. Depth of this box girder is
8.5 ft (2.6 m), so as to provide vertical stiffness and
to distribute live load and construction loads on the
deck to a sufficient number of adjoining cables.
Stay anchorage at the girder is at 30 ft (9 m) intervals, based on construction and aesthetic considerations. 26
FIGURE 9.86. Ruck-A-Chucky
ering (courtesy of T. Y. Lin).
Bridge, artiht‘s
rend-
/
) I LAY CAe‘ES
/
+
L\
TAY CANES
sourn 0U%5l0h
PEDESTALS
SOUTH INSlOE
FIGURE 9.87. Kuck-A-Chucky Bridge, plan of bridge with concrete alternate, from
reference 26.
.
ABt J WEN T
FIGURE 9.88. Ruck-A-Chucky
from reference 26.
Bridge, elevation of bridge with concrete alternative,
References
54’-0”
FIGURE 9. 89.
Ruck-A-Chuckv
f ro m ref erenc e 26
Bridge, cross section of concrete box girder alternative,
References
1. A . Feige, ”
The
Evolution
of
10.
German
Cable-Stayed
Bri d g e s- A n O v e ral l Su rv e y ,”
Acier-Stahl-Steel,
12,
in
December 1966
July 1967).
(reprinted
AISC
2.
H.
Thul,
‘Cable-Stayed
Bridges
in
C o n c re te Brid g e D e sig n , AC1 Pu b lic atio n SP23,
Paper 23-25, American Concrete Institute, Detroit,
No.
1969.
Engineering
Journal,
Germany,”
Pro-
11. .A non.. The Bridge Spanning Lake ,21aracn1bo
W i e s b a d e n . B e r l i n . Bauverlag
ezurla,
September 26 to 28, 1966, London.
3. W. Podolny, Jr., and
velopment
of
Structural
Division,
J. F. Fleming, “ Historical De-
Cable-Stayed
Bridges,”
Vol. 98,
ASCE,
Journal
No.
of
ST9,
13. N.
5.
M.
Inc.,
S.
New
J. B. Scalzi, ‘Construction and
York,
Troitskp,
Design,‘*
&
C o nc rete
Lockwood
“ Latest
Bridges-Theory
Staples,
London,
D ev elo p m ents
of
and
1977.
sphere,”
15.
Cable-
uf
and F. K. Chang, “ Longest Precast-
Bo x- G ird er
Civil
Brid g e
Engineering,
in
W estern
March
ASCE,
H em i-
1974.
W. Podolny, Jr., “ Concrete Cable-Stayed Bridges,”
Transp o rtatio n Re se arc h Re c o rd 665, Brid g e Engineering, Vol. 2, Proceedings, Transportation Research
Br i d g e s f o r L o n g Sp a n s , ”
Saetryk
Meddelelser, V o l. 45, N o . 4, 1974
Bygoningsstatzske
Board
Denmark).
M O . , N atio nal A c ad em y o f Sc ienc es, W ashing to n,
7. E. Torroja, Philosophy
by J. J.
California
of
Structures,
Structure
in
U.S.,”
Civil
Brid g e - Pio n e e r
Engineering,
ASCE,
January
F.
Leonhardt
Hangbrucken
w.eiten iiber
Bridge
16.
and
and
W.
Zellner,
“ Vergleiche
und Schragkabelbrucken
600
111,”
Structural
International
Engineering,
zwischen
fur SpannAssociation
Vol. 32, 1972.
for
September
25- 27,
1978,
St.
Louis,
H . Sc ham b e c k, “ The C o nstru c tio n o f the M ain
Brid g e- H o ec hst to the D esig n o f the 365 m Sp an
Rh e in Brid g e D u sse ld o rf - Fle h e ,”
Bridges, Structural Engineering Series
1978,
1958.
Conference,
D.C.
English version
Polivka and Milos Polivka, University of
Pt-ess, Bet-kelev and Los Angeles, 1958.
8. H . M . H ad l e y , “ Tie d - C an tile v e r
9.
Bri d g e i n .4rgentina,”
Paper No. 380, Vol. 59,
Fo u rth Q u arte r, 19i3.
14. H. B. Rothman,
1976.
“ Cable-Stayed
Crosby
6. F. Le o n h ard t,
Sta y e d
15, 1971.
Gray, “ ChacoiCorrientes
M unicipal Engineers Journal,
D e sig n o f C ab le - Stay e d Brid g e s,” Jo hn W ile y
Sons,
July
Record,
te m b e r 1972.
4. W. Podolny, Jr., and
(;mbH..
12. A non., “ Longest Concrete
Cable-Stayed
Span Cantile v e re d o v e r To u g h Te rrain ,” Engineering ‘Vews-
the
Sep-
in VU-
1963.
ceedings, Conzerence on Structural Steelw ork, Institution
oJ Civil Engineer.s,
R. Morandi, “ So m e Ty p e s o f Tie d Brid g e s in Prestressed
Concrete,”
First
International
Symposium,
Bridge
tration,
Division,
Washington,
Federal
18. A rv id
March-April
G rant,
Highway
Adminis-
D.C.
17. A n o n . , “ Tie1 Bridge,” Freyssinet
Bulletin,
C ab le - Stay e d
No. 4, June
International, STUP
1973.
“ Pasc o - Kennew ic k
Brid g e- The
Concrete Segmental Cable-Stayed Bridges
440
Freyssinet
Lo ng e st C ab le - Stay e d Brid g e in N o rth A m e ric a,”
Civil Engineting,
AXE, V o l. 47, N o . 8, A u g u st
1977.
19.
23.
Arvid Grant, “ Intercity Bridge: A Concrete Ribbon
o v e r th e C o lu m b ia Riv e r, W ash in g to n ,” CableStayed
Bridges, Structural Engineering Series No.
4, Ju n e 1978, Bri d g e D i v i si o n , Fe d e ral H i g h w ay
Administration,
20.
C.
Lenglet,
Washington,
“ Brotonne
C o nc rete
C ab le
Bridges, Structural
1978,
tration,
Bridge
Federal
Prestressed
C ab le - Stay e d
No. 4, June
Highway
Adminis-
D.C.
22. A n o n . ,
“ Th e
D an u b e
STUP
Bulletin,
May-June,
p end ed
Sp an s,”
Design Allows Constant Sus-
C’otr~wlting Enginret ( L o n d o n ) .
M arc h 1967.
24. H . J. G rah am , “ Dame
Point Bridge,”
Cable-Stayed
Bridges, Structural Engineering Series No. 4, June
tration,
Bridge
Division,
Washington,
Federal
Highway
Adminis-
D.C.
25. A n o n . , “ Dame Point Bridge,” Design Report, Howard N e e d le s Tammen
SC Be rg e nd o f f , N o v e m b e r
1976.
26. T. Y. Lin, Y. C. Yang, H. K. Lu, and C. M. Redfield,
2 1. Anon., “ Cable-Stayed Bridge Goes to a Record with
Hvbrid G i rd e r D e si g n , ”
O c to b er 28, 1976.
A non., “ Morandi-Stvle
1978,
Longest
Stay e d Brid g e ,”
Engineering Series
Division,
Washington,
D.C.
Bridge:
International,
1975.
Engznwring Nrw.+R~cod.
C an al
“ D e sig n o f Ruck-A-Chucky
Brid g e ,” C ab le - Stay e d
Bridges, Structural Engineering Series No. 4, June
1978,
Brid g e
( A u stria) ,”
tration,
Bridge
Division,
Washington,
Federal
D.C.
Highway
Adminis-
10
Segmental Railway Bridges
10.1
INTRODUCTION TO PARTICULAR ASPECTS OF
RAILWAY BRIDGES AND FIELD OF APPLICATION
10.8
10.2
LA VOULTE BRIDGE OVER THE RHONE RIVER,
FRANCE
10.9
10.10
10.3
MORAND BRIDGE IN LYONS, FRANCE
10.4
CERGY PONTOISE BRIDGE NEAR PARIS,
FRANCE
MARNE LA VALLEE AND TORCY BRIDGES FOR
THE NEW EXPRESS LINE NEAR PARIS, FRANCE
10.5
10.6
CLICHY
FRANCE
RAILWAY BRIDGE NEAR PARIS,
10.7
OLIFANT’S
10.1
Zntroduction
to Particular Aspects of Railway
Bridges and Field of Application
10.11
RIVER BRIDGE, SOUTH AFRICA
Construction of segmental post-tensioned bridges
for railway structures started in France in 1952
w ith a b rid g e c ro ssing the Rho ne Riv er at La
Voulte, Figure 10.1. It has been used extensively
since that time in many countries. Precast segm ental c o nstruc tio n w as intro d uc ed in railw ay
structures in France with the Marne la Vallee Viad uc t and in Jap an w ith the Kako g aw a Brid g e,
while incremental launching was adopted for several large railway crossings including the world’s
longest bridge of this type: the Olifant’s River
Bridge in South Africa (see Section 7.5).
The major characteristic distinguishing railway
bridges from highway bridges is the magnitude
and application of loading. Live loading on a railway structure is two to four times larger than that
applied to a highway bridge of comparable size.
Every time a train crosses a railway bridge, the actual load applied to the structure is much closer to
design live load than for a highway bridge, where
even dense truck traffic usually represents only a
INCREMENTALLY LAUNCHED RAILWAY
BRIDGES FOR THE HIGH-SPEED LINE, PARIS
TO LYONS, FRANCE
SEGMENTAL RAILWAY BRIDGES IN JAPAN
SPECIAL DESIGN ASPECTS OF SEGMENTAL
RAILWAY BRIDGES
10.10.1 Magnitude of Vertical Loads
10.10.2 Horizontal Forces
10.10.3
Bearings
10.10.4 Stray Currents
10.10.5 Durability of the Structure
10.10.6
Conclusion
PROPOSED CONCEPTS FOR FUTURE SEGMENTAL RAILWAY BRIDGES
moderate proportion of the design load. Fatigue
and durability of railway structures, therefore, are
essential problems and need careful consideration,
particularly in view of the fact that maintenance
and repair of railway structures under permanent
FIGURE 10.1. La Voulte Bridge. view of the completed structure.
441
Segmental
Railway
Bridges
traffic is a very critical operation that can lead to
unacceptable disturbance in a railway network.
10.2
La Voulte Bridge over the Rhone River,
France
This first segmental prestressed concrete railway
bridge is a notable structure and a landmark in the
development of prestressed concrete. Constructed
in 1952, it carries one railway track over the Rhone
River near la Voulte, 80 miles (128 km) south of
Lyons, in the southeastern part of France.
The structure has five spans, each 164 ft (50 m)
long. Each pier is made up of two inclined legs, and
each span is an independent frame supported by
an inclined leg at each end. Between the inclined
legs on each pier, the deck is supported by a small
beam resting on simple bearings.
Construction proceeded using the cantilever
scheme, with poured-in-place segments. The form
travelers were supported by a temporary steel truss
bridge, Figure 10.2. The cantilevers were built
sy m m etric ally in o ne sp an, the u nb alanc ed
m o m ents being taken care o f by tem p o rary
post-tensioning connecting the two inclined legs
and the independent beam on one pier. The segments were 9 ft (2.75 m) long. The bending moments of each completed frame were adjusted by
jacks placed at midspan and by continuity posttensioning tendons, Figure 10.3.
10.3
Morand Bridge in Lyons, France
This structure is a combined highway and masstransit bridge over the Rhone River in Lyons,
FIGURE 10.2. La Voulte Bridge, aerial view of the
deck
under
construction.
L,a Voulte Bridge, cantllcver- deck conprogress.
FIGURE 10.3.
struction
in
France’s third largest city. It is a three-span continuous structure w ith span lengths of 160, 292,
and 160 ft (49, 89, and 49 m), resting on two river
piers and two end abutments, which allow the transition of highway and railway traffic on both banks.
The deck is made up of two parallel box girders
carrying at the upper level three lanes of highway
traffic including sidewalks. Inside each box girder
is a railway track for the mass-transit system, Figure 10.4.
This final scheme proved to be significantly less
expensive and more efhcient in terms of the layout
of the railway system than did the initial proposal,
which contemplated a submerged tunnel for the
railway crossing and a separate bridge for the
highivay traffic.
Dimensions of the structure in cross section are
show n in Figure 10.5. The railw ay clearance of 13
ft 5 in. (4.12 m), including ballast and rail, calls for
a 15 ft (4.56 m) structural height in excess of the
normal requirements for a maximum span length
of 292 ft (89 m). A constant-depth girder could
thus be maintained throughout the river crossing
except in the vicinity of the river piers, where short
straight haunches allow the depth to be increased
to 22 ft 7 in. (6.90 m). Over the piers a strong
transverse diaphragm c o nnec ts the tw o b o x
girders, and the additional height over the pier
allows the continuity of the diaphragm over the
height of the haunch while the full clearance of
the trains is maintained inside the box girders.
The deck was built in balanced cantilever with 10
ft (3.0 m) long cast-in-place segments using one
pair of travelers on a typical one-week cycle, Figures 10.6 and 10.7.
Typical quantities of materials are as follows for
the deck alone:
Mot-and Bridge in Ly ons, France
FIGURE
ON PIER
FIGURE
10.5.
10.4.
Morand Bridge, perspective view of the structure.
AT MID-SPAN
Mm-and Bridge, typical cross section.
Deck area
Concrete
Reinforcing steel
Prestressing steel
(longitudinal and
transverse)
443
31,200 ft*
3,100 yd”
618,000 lb
256,000 lb
2,900
2,400
280,000
116,000
m2
m3
kg
kg
FIGURE 10.6. Mm-and Bridge, construction of the
superstructure.
Both concrete and reinforcing steel quantities far
exceed those required for a typical highway because of the very important increase of loads due
to the railway lines in the box girders.
The structure was completed and opened to
traffic in 1977.
Segmental Railway Bridges
single box carries the twin tracks, with the depth
varying between 13.6 ft (4.15 m) and 17.9 ft (5.45
m) for the maximum span length of 280 ft (85 m)
as show n in Figures 10.9 and 10.10. The segmental
deck was cast in place, with travelers working in the
conventional balanced cantilever fashion.
10.5
The extension of the Paris mass-transit system in
the highly populated southeastern suburbs was the
occasion for building a long elevated segmental
prestressed concrete railway structure in a sensitive
urban environment, Figure 10.11. This structure,
located in the city of Marne la Vallee, includes a
bridge over the Marne River and a long viaduct
carrying two parallel railway tracks. Near the transition between the river bridge and the viaduct a
passenger station is carried by the bridge structure.
Three major considerations guided the choice of
the structure:
FIGURE 10.7. MOI and Brid ge, (onstl uction o f
superstructure. Note pier segment for second parallel
box girder.
10.4
Mane la Vallee and Torcy Bridges for the
New Express Line near Paris, France
Cergy-Pontoise Bridge near Paris, France
A new railway line was completed in 1977 between
Paris and the new satellite town of Cergy-Pontoise.
A major prestressed concrete structure carries this
line over several obstacles, including an interchange between two expressways (A-86 and A-14)
and two branches of the Seine River.
The trestle structures have a solid slab deck with
spans varying between 65 ft (20 m) and 117 ft
(35.60 m). Typical dimensions of the two main
bridges over the Seine are shown in Figure 10.8. A
Maintain maximum clearance at ground level, not
only to reduce the visual disturbance to the
neighboring population, but also to allow all piers
of the new structure to be fully compatible with the
layout of all existing and future roads.
Elevation
(a)
Typical cross section
(b)
FIGURE 10.8. Cergy-Pontoise Bridge, dimensions. (a) Elevation. (b) Typical cross
section.
Marne la Vallee and Torcy Bridges near Paris, France
FIGURE 10.9. Ccrg!-Pontoise
str uctio n .
Bridge, cantilever con-
FIGURE 10.10. Ckrgy-Ponroise
closure.
445
FIGURE 10.11. Marne la Vallee Bridge, ae~i,~l Liew
of the completed structure.
Bridge, main span
FIGURE 10.12. Mane la Vailee Bridge, view of
finished structure from ground level.
Produce a structure that is aesthetically pleasing
when seen constantly from nearby.
In plan, the structure is laid out on a curve with a
minimum radius of curvature of 1640 ft (500 m),
Figure 10.11. Characteristic dimensions of the
Marne la Vallee Viaduct are shown in Figures
10.13 and 10.14 and are summarized as follow s:
Protect the neighboring population from unacceptable noise aggression.
Basically, the structure is a single box of constant
depth built of precast segments assembled by prestress into a continuous beam; the beam rests upon
vertical piers provided with an architectural shape
and regularly distributed at distances of 90 ft (27
m) to 120 ft (36 m), Figure 10.12.
Both parallel tracks are laid on the transversely
prestressed deck slab of the box girder and on a
crushed-stone bed retained sideways by three continuous reinforced concrete walls. A central noise
barrier separates the two opposite tracks and prevents the noise of a train riding one track to travel
across to the other. At the edge of the concrete box
girder, precast concrete panels manufactured with
special white cement improve the appearance of
the structure while providing the outside sound
barriers.
1.
Bridge over the Marne River:
a. Total length, 528 ft (161 m).
b.
Three-span continuous bridge with spans
of 157,246, and 125 ft (48,75, and 38 m).
c. Cross section: constant-depth box section
with depth of 12.8 ft (3.90 m), web thickness varying from 20 to 35 in. (0.50 to 0.90
m) and bottom flange thickness from 7 in.
(0.18 m) at midspan to 5 1 in. (1.30 m) over
the river piers. Length of precast segments
5.6 ft (1.71 m).
d . Tw o riv er p iers are fo und ed o n largediameter bored piles and support the
superstructure through special teflon
bearings.
n
I
t
ALLUVIUM
DEPOSITS
LIMESTONE
SAND
FIGURE
446
10.13.
Mar-ne la Vallee Bridge, typical sections of deck and piers.
Marne la Vallee and Torcy Bridges near Paris, France
447
f. A ll b earing s in the v iad u c t are stand ard
laminated elastomeric pads.
ck m --
-
4~70
I I 00
-
250
~~~
12
12 C R O S S
MID SPAN
CROSS
SECTION
SECTION ON PIER
g . Piers are m ad e o f tw in c o lu m ns lo c ated
under the webs of the box girder and connected at ground level by a common footing , w hic h transfers the lo ad s to d eep
slurry trenched walls anchored in limestone. The number and position of these
bearing walls under each pier has been
determined in relation to the magnitude
of the transverse and longitudinal horizontal loads transferred by the superstructure, particularly in the curved portion of
the viaduct.
(0)
(b)
FIGURE 10.14. Xlarne la Vallee and .Torcy Viaduct,
topical deck sections. (0) I\larne la Vallee trestle and
I‘orq Viaduct cross section. (h) klarne la Vallee Bridge
over the Llarne River.
2. Elev ated v iad uct:
a. Total length, 4482 ft (1367 m).
b. The v iad uct is d iv id ed into 11 sectio ns
separated by expansion joints, allowing
compatibility of thermal stress between the
continuous welded rails and the concrete
superstructure. The typical section is 412
ft (126 m) long with four spans of 88, 118,
118, and 88 ft (27, 36, 36, and 27 m).
c.
The
the
g er
and
d.
Typical cross section is a single box carrying the two tracks with two main vertical
webs 35 in. (0.90 m) thick and two sharply
inclined facia webs used essentially for architectural purposes to reduce the apparent structural depth of the box and focus
the eye on the high parapet wall.
e.
Average length of precast segments 7.5 ft
(2.30 m).
The entire project was predicated on the use of
precast segments with match casting and epoxy
joints. A precasting yard on the south bank of the
Marne, using four casting machines, produced the
690 segments with a maximum weight of 60 tons
(55 mt) in eleven months. Segments were transported with a tire-mounted self-propelled carrier
over the finished portion of the deck and placed in
the stru c tu re w ith a lau nc hing g antry , Fig u re
10.15, in balanced cantilever. The gantry used on
that project was that designed and built earlier for
the B-3 Viaducts project.
The gantry allowed all operations to be performed from the top in complete independence
from the ground and all its related constraints.
Placing of all segments was performed in a period
of nine months between March and December of
1976, including the three spans of the main bridge
and the fortv-four spans of the viaduct. The entire
p ro jec t w as’ c o m p leted in 24 m o nths (inc lud ing
preparation of the final design), for a total deck
area of 190,000 ft* (17,600 m’). Figure 10.16 shows
two south viaduct sections adjacent to
main river crossing carry the passenstatio n and hav e sho rter sp ans 69
92 ft (21 and 28 m).
FIGURE 10.15. Marne la Vallee Bridge, precast segments placed with the launching gantry.
FIGURE 10.18. Torcy Viaduct, segment transportation from Marne la Vallee to Torcy.
FIGURE 10.16. zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
l\larne la \‘allee Bridge, crossing the
Marne River and elevated passenger station.
The total length of 1870 ft (570 m) is divided into
three separate sections: one four-span unit, one
nine-span unit, and one four-span unit.
FIGURE 10.17. Marne la Vallee Bridge, aerial view of
the river crossing, a passenger station, and the elevated
viad uct.
Precast segments were placed in the structure
with an overhead launching gantry of a type
slightly different from the one used previously,
although calling on the same sequence of movements. Two parallel longitudinal trusses make
the track for a transverse overhead portal crane
carrying and placing the segments between the
trusses. Figures 10.19 and 10.20 show the general
view of the gantry and the detail of one segment
placing. The overall view of the finished bridge
appears in Figure 10.21.
the northern span of the river crossing and the elevated passenger station. Figure 10.17 is an aerial
view of the overall project.
In view of the success of this first application of
segmental construction in urban railway elevated
structures, the Paris Mass-Transit Authority decided to extend the same concept to construct another large structure a few miles eastward: the
Torcy Viaduct. Fortunately, the precasting yard
for the first bridge was still available and all segments could be manufactured there and trucked to
the second bridge site, Figure 10.18.
Dimensions of this new bridge are as follows:
Cross section: exactly the same as for the Marne la
Vallee elevated viaduct.
Distribution of spans: 17 spans w ith typical span
length of 115 ft (35 m).
FIGURE 10.19. ~l‘orcy Viaduct, precast segment placing with launching gantry.
Clichy
Railway
Bridge near Paris, France
FIGURE 10.20. I’orcy V i a d u c t , d e t a i l o f s e g m e n t
handling between twin trusses of launching gantry.
FIGURE 10.21.
structure.
10.6
449
vated metro. It crosses the Seine River adjacent to a
new highway bridge between the cities of Clichy
and Asnieres, as show n in Figure 10.22. Layout
and principal dimensions appear in Figures 10.23
and 10.24.
The prestressed concrete segmental structure is
1350 ft (412 m) long w ith a 280 ft (85 m) main span
over the river with a deck of variable depth. The
river piers of the two railway and highway bridges
match exactly to minimize water flow and barge
traffic disturbance. A provision is made for a second future highway bridge at the other side of the
railway bridge, as seen clearly in Figure 10.25~.
The restricted transverse clearances between the
three structures and their corresponding traffic
explains the special shape of the piers for the
center railway bridge, which was carefully studied
architecturally to enhance the appearance of the
project. Foundations were very close to one another but could be maintained structurally independent to better control settlement and avoid vibration interference between bridges and in the
ground.
To carry the two railway tracks, the deck has a
typical cross section consisting of three precast
webs connected by a bottom slab, which forms essentially the compression flange over the piers, and
an intermediate slab, which receives the ballast,
Figures 10.25b and 10.26. The depression thus
realized between the web top flange and the tracks
has several advantages, including providing full
safety against derailing on one track and reducing
the noise level.
Construction of the superstructure included
match casting of all webs in a yard near the project
site. The webs were placed in balanced cantilever
with a light portal crane carried by the finished
Torcy Viaduct, view of the completed
Clichy Railway Bridge near Paris, France
At about the same time the two structures described above were built, a large and innovative
railway bridge was constructed in the northeastern
suburb of Paris for another extension of the reno-
FIGURE 10.22. Clichy Railway Bridge, view of the
completed structure.
Segmental Railway Bridges
450
1 +o
L
t
3
9
!i
sur
4
1
1
.
7
0
Lot 4 SW
_ 76s~ .I
m
32
?i3
j+
rl,
FIGURE 10.23. Clichy Railway Bridge, layout and elevation of the structure. (n) Plan
view. (h) Elevation.
cf!mfy 4 .
~-~
(kg
FIGURE
C-7)
10.24.
~_~~~ ~~-~~ ~
(rps)
cps)
5ElNE
RIVER
P/a, (?&
6%fJ
m-g
Clichy Railway Bridge, main dimensions of segmental structure.
portion of the deck, Figure 10.27. Maximum
weight of precast webs was 19 tons (17 mt),
whereas segments that included the full three-web
box (or even a more conventional single box for
the equivalent span length) would have weighed in
excess of 66 tons (60 mt). After assembly of precast
webs with longitudinal post-tensioning, the two
twin slab sections were poured in place between the
webs in balanced cantilever on very simple travelers. Web segments were 7.3 ft (2.22 m) long for the
constant-depth part of the deck and 4.8 ft (1.48 m)
for the variable-depth part. In fact, the slabs were
cast in place between the three webs in two or three
increments of that length respectively (a length of
Clichy
Railway Bridge near Paris, France
451
LEG ENDE
(a)
FIGURE 10.25. Clichy Railway Bridge, typical sections of piers and deck. (a) Elevation of land and river
piers. (b) Dimensions of the deck cross section.
14.6 or 4.44 m) to reduce the number of site operations. A three-day cycle of operations could be
constantly maintained, including some overtime
work for the larger segments near the river piers.
Overall, construction in cantilever of the total
superstructure took one year between September
1977 and September 1978.
A special design aspect, specific to railw ay
bridges, was the transfer of horizontal loads (in-
FIGURE 10.26. Clichv Railway Bridge, pier segnrent
and cantilever constr&ion.
duced through braking or starting of the trains
over the bridge), to the piers and foundations. A
single fixed bearing was provided over pier P6, the
foundation of which was designed to transfer to
Segmental Railway Bridges
452
the need arise in the future. Two families of tendons could be added:
Above the lower slab in the two voids of the box
section, anchors being provided in blisters already
built in the structure.
Atop the center precast web and on the outside
face of the two facia webs, anchor blocks and deviation saddles being prestressed by high-strength
bolts to the precast webs.
FIGURE 10.27. Clichy Kailway Hrtdge,
cast webs for cantilever construction.
plar~ng
pre-
the limestone stratum the total maximum horizontal load of 660 tons (600 mt) applied to the
bridge. There are three pot bearings between the
deck and the pier shaft, each capable of safely
transferring half of the maximum horizontal load.
Each bearing can thus be changed under traffic
without reducing the capacity of the structure.
Special provisions were also included at the design stage, Figure 10.28, to allow additional prestressing to be incorporated in the structure should
The large precast architectural panels on both
sides of the deck could be temporarily removed to
allow this work of additional prestresstng to be performed. Upon completion, all additional tendons
would be fully protected and concealed behind the
panels.
The new line has been open to traffic since May
1980, and the first months of operation confirm
that the precautions taken to reduce noise and vibration disturbance through welded continuous
rails and sound-barrier panels make such elevated
structures an acceptable solution for mass-transit
lines in urbanized areas.
10.7
fb)
FIGURE 10.28. Clichy
Railway Bridge. (a) View of
ad.jacent highway and railway bridges crossing the Seine
River. (b) Provisions for future additional p&tress.
Olifant’s River Bridge, South
Africa
This structure is part of a line carrying iron ore on
special heavy trains 7500 ft (2300 m) long made up
of 200 cars with a total weight of 19,000 tons
(17,000 mt) to connect the Sishen mines to the harbour of Saldanha 110 miles north of Capetown.
Olifant’s River Viaduct is today the world’s longest
incrementally launched prestressed concrete
structure (refer to Chapter 7) with a total length of
3400 ft (1035 m) and 23 spans of 148 ft (45 m),
Figure 10.29.
Shown in cross section in Figure 10.29, the single
box girder deck accommodates only one track on
ballast. The equivalent uniform live load of the 33
ton (30 mt) axles is 7.1 kips/lineal ft (10.5 mt/lm),
which is increased by an impact factor of 1.29.
The 23 spans are divided into two 1 l-span sections, each anchored to the end abutment, and one
single transition span at the center. This scheme
allows all horizontal loads to be transferred to the
abutments. The maximum horizontal reaction including all thermal effects is in excess of 1200 tons
(1100 mt). The piers, which vary between 80 ft (25
m) and 150 ft (55 m) in height, are extremely flexible and do not, therefore, have an important effect
on the horizontal restraint of the structure, except
during construction. The pier shafts have an I-
Incrementally
I
51nl
--L
L
3.10
5 50m
;I
Launched
Railway
Bridges
453
zyxw
c -I
zyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
(b)
FIGURE 10.29. Olifant’s River Viaduct. (a) General
view of the structure. (b) ‘T‘ypical
cross section.
shaped cross section with longitudinally tapered
faces. Neoprene bearings are used for the piers
near the abutments and teflon sliding bearings in
the center of the structure.
The deck was entirely constructed behind one
abutment (see schematic view in Figure 7.29) and
incrementally launched in one direction. Construction time for the superstructure was nine months,
with a theoretical cycle of 10 working days for a
typical 148 ft (45 m) span realized after 10 spans; it
was further reduced to seven days with two shifts
toward the end of the project. The total weight of
the superstructure of 14,500 tons (13,000 mt)
called for two 200 ton jacks for the push-out operations in increments of 3.5 ft (1 .OO m). A 60 ft (18 m)
long launching nose was used in front of the first
span to reduce the variation of bending stresses in
the superstructure during the successive stages of
construction, Figure 10.30. The bridge nearing
completion is shown in Figure 10.3 1; it was opened
to iron ore trains in 1976.
10.8 Incrementally Launched Railway Bridges for
the High-Speed Line, Paris to Lyons, France
FIGURE 10.30. Olifant’s River Viaduct, launching
nose reaching beyond a high pier.
build some new very-high-speed train lines (safe
maximum speed of 200 mph or 320 km/hr) and
started the construction of the first such line between Paris and Lyons, which included an entirely
new structure over a distance of 250 miles (400
km) with proper connections to the existing metropolitan track and station system.
The new project required 400 bridges including
nine large viaducts, such as the structure shown in
Figure 10.32. A very comprehensive optimization
study followed, and a set of guidelines and structural standards were prepared for the French National Railways by a team of engineers headed by
one of the authors. Results of the preliminary investigations and of this optimization study can be
summarized as follows:
1.
To meet increased competition by domestic airlines, the French National Railways decided to
Track alignment is chosen to keep the curvature in plan more than 10,500 ft (3200 m) and
preferably more than 13,000 ft (4000 m). The
4.
5.
6.
FIGURE 10.31. Olifant’s River Viaduct, view of the
structure nearing completion.
FIGURE 10.32. Railwq
Viaducts for Paris-Lyons
high-speed line, view of the viaduct over the Saone
River.
corresponding cross fall between rails is 7 in.
(180 mm).
2. All rails are to be continuously welded and
placed on a ballast bed with a minimum thickness of 14 in. (0.35 m).
3. Maximum rigidity of the structures is obtained
by using a continuous box section with slenderness ratio of l/14. The corresponding
maximum deflection under design load is
therefore l/2000 of the span, whereas conventional specifications for normal-speed lines
allow up to l/800.
Adopt as much as possible single box girder
decks for the two parallel tracks with minimum
web thickness of 14 in. (0.35 m) and a
minimum top slab thickness of 10 in. (0.25 m).
The optimum span length is between 150 and
170 ft (45 to 50 m), which leaves the construction method open to various solutions (cantilever, span-by-span or incremental launching).
The horizontal loads should be transferred to
one abutment equipped with a special fixed
bearing, allowing all piers to be relieved of any
appreciable longitudinal bending. A typical
H-section was adopted as the most appropriate except for certain specific locations
where a box section might be required.
Because many of the viaducts were located in environmentally sensitive areas, an overall architectural study was also conducted to establish a unity
of appearance for all bridges in terms of the shapes
of deck and piers, parapet or guard rails, abutments and approach fills.
Of the nine viaducts, two were finally constructed with conventional methods and the remaining seven were incrementally launched. This
method proved economical in view of the moderate span lengths, the depth of the box section available, and especially because the superimposed
dead and live loads were so much more important
than for a highway bridge that the increased
dead-load moments during construction were in
proportion of much less significance.
Table 10.1 gives the essential characteristics of
the seven segmental viaducts, including principal
quantities of materials for the superstructure. Elevations of five bridges appear in Figure 10.33.
As an example of the construction method, some
details are given for the bridge over the Saone
River, where a launching nose 93 ft (28.50 m) long
and weighing 71 tons (65 mt) was used in front of
the first span to reduce the stress variations in the
superstructure during launching, Figure 10.34.
The bridge superstructure was cast in successive
increments in a fixed location behind the right
bank abutment in the length of a half-span,
Figure 10.35. A typical sequence of operations is
shown schematically in Figure 10.36. Each
superstructure section is in fact cast in two stages
T able 10.1.
Charact erist ics of Segment al Viaduct s on t he Paris-Ly ons High-Speed
Line
Quantities of
Deck
Dimensions of Deck
[ft or in. (m))
zyxwvutsrqponmlkjihgf
Bridge
Location
and Span Lengths
[ft @)I"
@J Saulieu
115-3@ 144- 115
(35 - 3 @ 44 - 35)
Q Serein
115-3@ 1 4 4 - 1 1 5
(35 - 3 @ 44 - 35)
0 SamRium
155 ~ 5 (a: 164 - 137
(47.2 - 5 6 50 - 41.8)
@ figtine
109-8@ 144- 1 0 9
(33.4 - 8 @ 44 - 33.4)
@I Rochc
108-7@149-108
(33.1 - 7 I@ 45.5 - 33.1)
0 Seine River
114-ZOl- 114
(34.8 - 61.4 - 34.8)
Q cmler Cad (length)
85 - 105 - 85
(26 - 32 - 27)
Flange Thick.
Bridge
Layout
hong. grade: 1.3%
Radius in plan:
20,000 ft (6000 m)
Lrmg. grade: 0.95%
Radius in plan:
26,000 ft (8000 m)
Circular profile
in elevation:
R = 130,000 ft
(40,000 m)
Long.
grade: 2.5%
Straight in plan
Long. grade: 3.5%
Radius in plan:
10,600 ft (3250 m)
Long. grade: 0.55%
Straight in plan
Long. grade: 0.2%
Straight in plan
Bridge
Length
Total
Width
(ft)
Height
BOX
Width
(W
(f0
Web
Thick.
(in.)
662
(202)
41.0
(12.50)
10.8
(3.30)
18.0
(5.50)
662
w-w
41.0
(12.50)
10.8
(3.30)
1112
(339)
40.3
(12.30)
1370
C0ncr.
H.T.
Steel
Reinf.
Steel
(in.)
Bottom
(in.)
Pier
Year
Height
[ftvf?
[Ib/yd3
[Ib/yd3
Cornpleted
lft WIzyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJ
WWI
(k g/ m?1
(kg/&)1
18/49
(0.45/1.25)
11
(0.275)
IO
(0.25)
461121
(14137)
2.52
(0.77)
(46)
18.0
(5.50)
18/49
(0.45/1.25)
11
(0.275)
IO
(0.25)
661148
(20/45)
2.52
(0.77)
78
240
(46)
(140)
1979
11.5
(3.51)
18.0
(5.50)
20
(0.50)
12.5
(0.32)
12.5
(0.32)
46
(14)
2.46
(0.75)
84
(50)
210
(125)
1979
39.0
(11.90)
10.8
(3.30)
18.0
(5.50)
24
(0.60)
12.5
(0.32)
14
(0.35)
43/105
( 13/32)
2.30
(0.70)
84
(50)
250
(150)
1978
39.0
(11.90)
10.8
(3.30)
18.0
(5.50)
24
(0.60)
12.5
(0.32)
14
(0.35)
43/l 15
(I 3135)
2.30
(0.70)
84
(385)
(50)
250
(150)
1978
429
(131)
41.0
(12.50)
13.1
(4.00)
19.0
(5.80)
24135
(0.60/0.90)
11
(0.28)
12120
(0.30/0.50)
36
(11)
84
(50)
190
(110)
1980
279
40.0
(12.10)
7.8
(2.37)
19.0
(5.80)
12
(0.30)
10
(0.25)
8
(0.20)
43
(13)
-
(419)
1260
(85)
“Structures are numbered with increasing numbers from Paris to Lyons.
TOP
78
240
(140)
1978
1978
zyxwvutsr
Viaduc du Serrein
FIGURE
456
10.33.
Elevation of five segmental bridges for Paris-Lyons line.
Segmental Railway Bridges in Japan
457
(bottom slab during the first stage, webs and top
slab during the second stage). The typical construction cycle allowed casting a half-span every
week-that is, constructing two spans per month.
The launching operation proper called for a
very efficient system, developed and perfected
previously in Germany, including under each web
of the box section:
One vertical jack with sliding plate
Two coupled horizontal jacks for actual launching,
allowing movements in 3 ft increments
FIGURE 10.34. S;IO~C Kivcl- Bridge,
approachtng pier.
launching nose
Typically, launching of an 80 ft section took three
to three and a half hours, despite the large weight
of the concrete superstructure, reaching 9000 tons
(8000 mt) at the end of construction.
Figure 10.37 shows a completed structure, and
Figure 10.38 shows another aspect of the construction of these seven viaducts.
10.9
Segmental Railway Bridges in Japan
Many railroad bridges have been built in Japan
using the segmental construction technique. The
sketches shown in Figures 10.39 through 10.42 depict the elevation and the cross section of the following cast-in-place segmental bridges:
FIGURE 10.35. Saone Kivcr Bridge, xrial 1 ie\\ with
casting vard in behind abutment in foreground.
SLIDING
.~~
Natorigawa Bridge
zyxwvutsrqp
TEMPORARY
SUPPORT
F O R PUSHN;
PACE
*
Kyobashigawa Bridge
-
-
-
-
APPROACH
_^
-wAN~PRECAST Y A R D
e
-~
SITUATION DURING FABRICATION OF SEGMENT 7
,2r550~25~0
7Q5 0 0-%!!L
SITUATION AFTER PUSHING OF SEGMENT 7
GENERAL PRINCIPLE OF THE CONSTRUCTION METHOD BY PUSHING
FIGURE 16.36. Saone River Bridge, typical construction stages of incremental launching.
FIGURE
FIGURE 10.38. Digoine
ing over high piers.
Kisogawa
Ashidagawa
10.37.
Saone River Bridge, view of‘ the completed structure.
Bridge, incremental launch-
Bridge
Bridge
Figure 10.43 shows the Kakogawa Bridge during
construction. The superstructure is made of twin
constant-depth box girders, one box girder carrying one railway track. The total length of the
bridge is 1640 ft (500 m), with typical span length
of 180 ft (55 m). Each box is 13 ft (4 m) wide and
11.5 ft (.3.50 m) deep. The precast segments were
handled by a launching gantry and assembled by
longitudinal post-tensioning tendons. The erection
used the balanced cantilever system.
The most outstanding prestressed concrete railway structure, however, is the Akayagawa arch
bridge shown in Figure 10.44. Total length is 980
ft (298 m) and the center arch span is 410 ft (126
m). The 13 ft (4.00 m) deep box girder carrying
two railway tracks is continued throughout between abutments and rests over the center gorge
on a very flat arch rib through ten spandrel columns. The respective proportions are such that the
deck carries all bending moments and the arch rib
carries the normal load induced by its curvature.
The erection scheme was unique and called for
cantilever construction starting from both sides.
A very strong back stay made up of a prestressed
concrete member with a prestress force of 5300
tons (4800 mt) was installed diagonally between the
top of the main transition piers between the arch
structure and the approaches on one hand, and the
foundation of the adjacent piers in the approach
structures on the other hand.
While erection progressed, high-strength steel
bars were placed diagonally between the vertical
members, forming a temporary truss structure
until the crown was reached from both ends. Control of tensioning of those steel bars was very critical and complicated. Finally, all steel bars and the
two temporary back stays were removed after closure of the arch at midspan.
10.10
Special Design Aspects of Segmental
Railway Bridges
10.10.1 MAGNITUDE OF VERTICAL LOADS
Most bridges carry tracks laid on ballast with a
minimum thickness of 10 to 14 in. (0.25 to 0.35 m).
OSAKA
- L 33.90
SHIN
199
8Om
HAKATA
66.00
66.00
33.90
ELEVA TION
p-s- 61 51
/ j ---+I
2 so 2.70 m
IO'
90592
24,
i
CROSS SECTION
FIGURE 10.39. Hyobashigalva
Live loading used in design of railway bridges
varies between countries-Cooper loading for
Anglo-Saxon countries, new UIC loading for most
European countries-and also according to the nature of the structure: mass-transit lines are usually
designed for much lighter loads than normal train
lines. The heaviest loadings are for ore freight
trains.
To exemplify the basic difference between a
highway and a railway bridge, Figure 10.45 compares a typical 150 ft span and a 36 ft w ide deck
normally designed for three highway lanes of
traffic or two railway tracks. The total superimposed dead and live load is 3.6 times greater for
the railway bridge. In addition, the weight of ballast (representing 40% of the total load) must be
considered as a live load to cover the cases where
the ballast is removed from the deck or has not yet
been placed on a new bridge.
zyx
459
Special Design Aspects of Segmental Railway Bridges
Bridge, Japan
10.10.2 HORIZONTAL FORCES
Railway bridges have to carry very important horizontal forces, between five and ten times the horizontal forces carried by a highway bridge of similar
size. The standard current practice for long viaducts is to have a fixed bearing on one abutment if
the bridge length is less than 1500 ft (450 m), and
on both abutments and on intermediate special
bents if it is greater. The order of magnitude of
this horizontal force on the abutments carrying the
fixed bearings is often 1000 tons for a two-track
viaduct.
The various forces involved are described below:
Longitudinal Forces
Braking and acceleration forces
Segmental Railway Bridges
460
/
TOKYO
MORIOKA
,A
524.90 m
z
ELEVATION
CROSS SECTION
FIGURE 10.40. Natorigawa Bridge, Japan.
Forces due to box girder deformations: creep,
shrinkage, and temperature variations
Loads induced by the length variations of long
welded rails under temperature variations
Longitudinal component of wind forces
Braking and ac c eleratio n f o rc es are oneseventh of the total weight of live loads, with a
ceiling of 285 tons for braking and 53 tons for acceleration (French regulations).
Forces due to longitudinal deformations of the
box girder vary because of creep, shrinkage, and
temperature variations. The bearing displacements induce horizontal loads by distortion or
friction.
Length variations of the long welded rails due to
temperature variations create a horizontal force
parallel to the rail. This force can be estimated at
50 tons per rail (length of rail more than 100 meters). For a two-track bridge it is 2 x 2 x 50 = 200
tons.
Longitudinal component of wind forces are described in the AASHTO specifications for bridges.
Transuerse Ho riz o ntal Fo rces
Centrifugal horizorltal
force can be very important f-or high-speed trains.
For the 200 mph train from Paris to Lyons this
force is more than 400 tons for some viaducts 1200
ft (380 m) long with two tracks and radius of curvature of 10,500 ft (3200 m). The lateral acceleration is more than 20% of that of gravity.
Trans v ers e w ind fo rce is d esc rib ed in the
AASHTO standards (50 lb/ f?).
10.10.3
BEARINGS
In order to gain complete control of these ver)
large horizontal forces, the bearings are specially
designed to take care of the vertical loads and rotation of the box girder and simultaneously to provide all possible horizontal restraints (fixed bearing, bearing free lengthwise or crosswise, or both).
Special Design Aspects of Segmental Railway Bridges
461
NAGOYA
-
ELEVA TION
CROSS SECTION
FIGURE 10.41. Kisogawa Bridge, Japan.
These bearings are specially manufactured for this
tvpe of structure, Figures 10.46 and 10.47. The
sliding parts consist of a teflon-coated plate resting
on a stainless steel plate, and the restraints are provided by steel keys.
ference of potential with the ground may be measured at regular intervals, and a permanent connection with the ground may be decided on as a
result.
10.10.~
10.10.4 STRAY CURRENTS
For structures carrying electrified railways there is
some uncertainty about the long-term effect of
stray currents generated near the power lines. In
order to preclude electrolytic corrosion of reinforcing steel and prestressing steel, the following
precautions are now taken in prestressed concrete
structures:
DURABILITY OF THE STRUCTURE
Because very difficult problems of train traffic
would arise during repairs to these bridges, their
durability needs special attention. The following
provisions were established for the high-speed
bridges between Paris and Lyons:
Under the worst service loads the concrete must
remain under a 140 psi minimum compression.
The deck is electrically isolated from the ground,
piers, and abutments by elastomeric plates.
For continuous bridges, the design shall be
checked by weighing the dead-load vertical force
on the bearings.
The reinforcing and prestressing steel systems of
the entire deck are interconnected by mild steel
bars to equalize the electric potential. The dif-
The stressing force of the post-tensioning tendons
shall be less than 80% of the ultimate strength of
the tendons.
Segmental Railway Bridges
462
317.00 m
ELEVATION
fy
FJ
L.
‘i
r’.
---e-J
I
53
2.61
‘!I
2.30
1
./.
2.70
5.90
/
1
L
I
2.41
10.92 m
CROSS SECTION
FIGURE 10.42. Ashidagawa Bridge, Japan.
10.10.6 CONCLUSION
FIGURE 10.43. Kakogawa Railway Bridge, placing
precast segments with launching gantry.
The ultimate strength of the structure should be
capable of supporting the service loads increased
by 30%, if 30% of the post-tensioning steel were
missing.
Provisions shall be made for installing additional
tendons while the structure is under traffic. The
additional post-tensioning force shall be 15% of
the designed force minimum.
It shall be possible to replace all the bearings.
This review of specific design problems of railway
bridges should raise no doubts whatsoever about
the advantages of prestressed concrete and segmental construction’in this field. Prestressed concrete is the safest material known today to resist
indefinitely the large variations of loads such as
those applied to a railway bridge.
The problem of fatigue has been covered briefly
in Chapter 4, and the results mentioned there
apply particularly well to railway bridges. The
main objective in the design and construction of
prestressed concrete bridges should be to minimize
and even eliminate concrete cracking, which is always a source of weakness in a structure subject to
cyclic lo ad ing.
The use of the provisions laid down in Section
10.10.5 should result in practically crack-free
structures with an expected life free of major
maintenance.
978
I
184’
,
98’
,
,
I
414’
98’_,_
I
184
I
I
(b)
FIGURE 10.44. Akayagalva Rail\\av Bridge, general dimensions. ((0 Elevation. (1~) l‘vpical cross section .4-A.
55 k
(25tl
5.4 k/cF
(8 t/ml)
55 k
125t1
Y
5.2’
Span length
Deck width
Number of lanes or tracks
Superimposed dead load:
Ballast
Curb, pavement, etc.
Total S.L.
Live loads:
Equivalent uniform
load
Impact factor
Total L.L.
Total (S.L. + L.L.)
55 k
(25d
(8 t/ml1
V
Description
55 k
(25d
V
5.2’
v
5.2’
2.6’
+I----
Highway Bridge
Railway Bridge
150 ft (45 m)
36ft (11 m)
Three lanes
150 ft (45 m)
36ft(ll m)
Two tracks
1.5 kips/ft
-
6.5 kips/ft
0.5 kips/ft
7.0 kips/ft
1.5 kips/ft
6.8 kips/ft
2.4 kips/ft
30%
18%
2.8 kips/ft
8.8 kips/ft
4.3 kips/ft
15.8 kips/ft
FIGURE 10.45. Vertical loading on railway bridges. (a) Typical UIC - track loading.
(b) Comparison of superimposed dead and live loading on highway and railway bridges.
463
464
Segmental Railway Bridges
10.11
Proposed Concepts for Future Segmental
Railway Bridges
We should note that many types of structures described for highway bridges are equally appropriate
for railway bridges: the structures described in this
chapter were essentially girder or arch bridges
built in cantilever or incrementally launched.
Today, many design projects are based on stayed
bridges. As an example, Figure 10.48 shows a proposed crossing of the Caroni River in Venzuela for
heavy iron ore freight trains.
FIGURE 10.46. Detail of pot bearing with unidirectional horizontal movement.
FIGURE 10.48. Proposed crossing of Rio Caroni for
iron ore railway line.
FIGURE 10.47. Detail of fixed bearing.
11
Technology and Construction
of Segmental Bridges
11.1 SCOPE AND INTRODUCTION
11.2 CONCRETE AND FORMWORK
FOR SEGMENTAL
CONSTRUCTION
11.2.1
11.2.2
11.2.3
11.2.4
11. 3
Concrete Design and Properties
Concrete Heat Curing
Dimensional Tolerances
Formwork for Segmental Construction
POST-TENSIONING
MATERIALS
AND
11.6
11.6.1
11.6.2
11.6.3
11.6.4
11.6.5
OPERATIONS
11.3.1 General
1 1 . 3 . 2 Ducts
11.3.3 Tendon Anchors
11.3.4 Tendon Layout
11.3.5 Friction Losses in Prestressing
11.3.6 Grouting
11.4
11.3.7 Unbended Tendons
SEGMENT FABRICATION FOR
CANTILEVER CONSTRUCIION
Tendons
CAST-IN-PLACE
11.7
11.8
SEGMENTS
Launching Girders
Launching Girders Slightly Longer Than the Span
hn%h
Launching Girders Slightly Longer Than Twice
the Typical Span
CHARACTERISTICS OF PRECAST SEGMENTS AND
MATCH-CAST EPOXY JOINTS
Certain problems are common to all types of segm ental c o nstru c tio n- f o r exam p le, the selec tio n
and
control of materials, prestressing operations,
and choice of bearings, joints, and wearing surface.
Other- problems are specific to a particular construction method. The use of form travelers in
cast-in-place cantilever construction and the casting and handling of segments in precast cantilever
construction are two such examples. This chapter
covers these various topics in the following order:
CAST SEGMENTS
PLACING PRECAST
11.8.3
Flanges
11.4.5 Practical Problems in Cast-in-Place Construction
Camber Control
Scope and Introduction
Survey Control During Construction
Conclusion
11.6.6 Precasting Yard and Factories
HANDLING AND TEMPORARY ASSEMBLY OF PRE-
11.8.1 Independent Lifting Equipment
11.8.2 The Beam-and-Winch Method
11.4.2 Self-Supporting Mobile Formwork
11.4.3 Tw&tage Casting
1 1 . 4 . 4 Combination of Precast Webs with Cast-in-Place
1 I .l
Introduction
Long-Line Casting
Short-Line Horizontal Casting
Short-Line Vertical Casting
Geometry and Survey Control
Segment Precasting in a Casting Machine
Segment Casting Parameters
Survey Control During Precasting Operations
11.4.1 Conventional Travelers
11. 5
11.5.1 First-Generation Segments
11.5.2 Second-Generation Segments
11.5.3 Epoxy for Joints
MANUFACTURE OF PRECAST SEGMENTS
REFERENCES
1.
Problems common to all segmental bridges
2. Pro b lem s sp ecific to cast-in-p lace cantilev er
construction
3. Pro b lem s sp ecific to m atc h- c ast seg m ental
bridges with particular emphasis on cantilever
construction, which is the most widely used
method.
In designing segmental bridges, it is important
to pay attention to certain details at the time of
conception, in order to keep the project as simple
as possible and thereby achieve economy and effi465
466
Technology and Construction of Segmental Bridges
ciency during construction. The following guidelines apply to both cast-in-place and precast construction:
1.
Keep the length of the segments equal, and
keep the segments straight even for curved
structures (chord elements).
2.
Maintain constant cross-section dimensions as
much as possible. Variations of cross-section
d im ensio ns sho u ld b e lim ited to c hang e o f
depth of webs and thickness of bottom slab.
3.
4.
Corners should be beveled to facilitate casting.
5.
A v o id as m uch as p o ssib le surface discontinuities on webs and flanges caused by anchor
blocks, inserts, and so on.
6.
Use a repetitive layout for tendons and anchors, if possible.
7.
Minimize the number of diaphragms and stiffeners.
8.
Avoid dowels passing
possible.
Segment proportions (shear keys, for example)
should be such as to allow easy form stripping.
through
formwork, if
11.2
Concrete and Formwork for Segmental
Construction
11.2.1
CONCRETE DESIGLV
A,VD PROPERTIES
Uniform quality of concrete is essential for segmental
construction. Pro c ed u res f o r o b taining
high-quality concrete are covered in PC1 and AC1
p ub lic atio ns.‘ ** Bo th no rm al w eig ht and lig htweight concrete can be made consistent and uniform by means of proper mix proportioning and
production controls.
Ideal concrete will have a slump as low as practicable, notwithstanding the possible use of special
placing equipment such as pumps, and a 28-day
strength greater than the minimum specified by
structural design. It is recommended that statistical
methods be used to evaluate uniformity of concrete mixes.
The methods and procedures used to obtain the
concrete characteristics required by the design may
vary somewhat, depending on whether the segments are cast in the field or in a plant. The results
will be affected by curing temperature and type of
c uring . Liq uid o r steam c uring o r elec tric heat
curing may be used.
In temperate climates and where curing is carried out in an isothermal enclosure, only small ad-
ditions of heat are required to maintain the curing
temperature, full advantage being taken of the
heat of hydration generated by the fresh concrete.
In this case heat demand will be a function of the
ambient temperature, more heat being required in
winter and little or no additional heat during hot
summer weather.
Where segment production rate is not critical, it
may be possible to do without accelerated curing
and simply use a normal curing period of a few
days, during which the concrete is well protected
against excessive temperature variations and all
exposed surfaces are kept moist.
A sufficient number of trial mixes must be made
to assure uniformity of strength and modulus of
elasticity at all important phases of construction.
Careful selection of aggregates, cement, admixtures, and water will improve strength and modulus of elasticity and will also reduce shrinkage and
creep. Soft aggregates and poor sands must be
avoided. Creep and shrinkage data for the concrete mixes should be determined bv tests.
Corrosive admixtures such as calcium chloride
should never be used, since they can have a detrim ental ef f ec t o n hard ened c o nc rete and c an
cause corrosion of reinforcement and prestressing
steel. W ater-red u c ing ad m ixtu res and also airentraining admixtures that improve concrete resistance to environmental effects, such as de-icing
salts and freeze and thaw actions, are highly desirable. Very careful control at the batching stage is
required, however, since the advantages of airentrained concrete cannot be relied upon unless
the quantity of entrained air is within specified
limits.
The cement, fine aggregate, coarse aggregate,
water, and admixture should be combined to produce a homogeneous concrete mixture of a qualitv
that will conform to the minimum held-test and
structural design requirements. Care is necessary
in proportioning concrete mixes to insure that the\
meet specified criteria. Reliable data on the potential of the mix in terms of strength gain, creep, and
shrinkage performance should be developed to
serve as the basis for improved design parameters.
Proper vibration should be used to permit the
use of low-slump concrete and to allow for the optimum consolidation of the concrete.
11.2.2 CONCRETE HEAT ClJRI,\‘G
An early concrete strength usually is required to
reduce the cycle of operations and to maintain the
467
Concrete and Formwork for Segmental Construction
efficiency of the special equipment used either in
cast-in-place or in precast construction. Two
methods may be used for this purpose, either separately or together: (a) preheating the fresh concrete, before placing it in the forms or in the casting machines, (b) heat curing the concrete after
consolidation in the forms.
In the first case the concrete is preheated to
about 85 to 90°F (30 to 35°C). This operation is
achieved in several ways:
1.
Steam heating the aggregates-a simple solution that presents the disadvantage of changing the aggregate water content
2.
Heating the water-a solution that has limited
efficiency, owing to the small proportion of
water in comparison with the other components (water at 140°F raises the concrete temperature by only 20°F).
3.
Direct heating of the concrete mix by injecting
steam into the mixer itself-the best solution
and the one most easily controlled.
To avoid heat loss, the forms are generally insulated and some source of radiant heat is installed
inside the segment (radiators or infrared elements).
In the second case, the concrete is heated in its
mold inside a container in which low-pressure
steam is circulated. In this way it is relatively easy to
obtain the strength required for prestressing operations [3500 to 4000 psi (25 to 28 MPa)] after one
or two days, even in winter. If however, tensioning
operations are to be performed earlier, after 24
hours for example, modifications must be made to
the concrete in the anchorage zone.
Electrical resistances may be embedded in the
concrete, or precast end-blocks may be used. Precast end-blocks were used notably for the Issyles-Moulineaux, Clichy, and Gennevilliers Bridges.
For the Gennevilliers Bridge, despite the exceptional dimensions of the box girder deck, two
segments were cast each week through an early
stressing of the prestress tendons.
In the case of precast segments, the accelerated
curing of the concrete must attain two apparently
contradictory objectives:
1.
Accelerated curing to permit rapid stripping.
2.
Final compressive strength as near as possible
to that of the design concrete.
Several curing systems may be considered:
1. Conventional kilns.
2. Direct heating of forms with electric resistances.
3. Direct heating of forms with low-pressure
steam.
The use of a conventional kiln entails several
precautions. First, a constant temperature must be
maintained in the kiln. Second, the segment sections of varying thickness are all heated to the same
temperature, which may produce unacceptable
local thermal gradients and cracking if heat curing
is excessive. Finally, the heated segment may be
subjected to a thermal shock when removed from
the kiln, if the difference between the ambient
temperature and the kiln temperature is greater
than 60°F. However, kiln curing is a simple solution and is acceptable for long curing cycles-for
example, of 10 to 14 hours.
Form heating by means of electrical resistances is
perfectly adapted to long curing cycles. This system permits a wide range of adjustment per zone,
varying the temperature between the thick and
thin sections of the segment and thereby minimizing thermal gradients and eliminating the risk of
permanent damage to the concrete at the beginning of its solidifying phase.
The heating of forms with low-pressure steam is
preferable for short curing cycles lasting less than
five hours, as it permits the distribution of a large
quantity of calories over a short period, causing a
rise in the internal temperature of the concrete of
the order of 20 to 30°F (10 to 15°C) per hour. This
system, however, requires a complex regulator to
ensure an equal temperature in all the form panel
enclosures, at all times during the treatment, whatev er their therm al inertia and the external
influences to which they are subjected, Figure 11.1.
FIGURE 11.1.
Viaducts).
Heat-curing
control system
(B-3 South
Technology and Construction of Segmental Bridges
468
The different systems (kiln, electrical resistances,
and low-pressure steam) have all been applied successfully to segmental bridges. The segments for
the Choisy-le-Roi and Courbevoie bridges were
kiln cured. Electric heating was adopted for the
c o nstru c tio n o f the u p stream and d o w nstream
b rid g es o n the Paris Ring Ro ad and the Blo is
Bridge, among others. Form heating using lowp ressu re steam w as u sed f o r the Pierre Benite
Bridges, the Oleron Viaduct, and the B-3 South
Viaducts.
Whether forms are heated by electricity or b)
steam, it is relatively easy to produce a long curing
cycle, and the desired final concrete strength is
easily obtained. A short curing cycle, on the other
hand, requires a great deal of caution and meticulous preliminary calculations. Particular attention
must be given to:
In order to avoid a drop in the long-term mechanical properties of the concrete, the temperature curve during the heat curing must necessaril\
include, see Figure 11.2:
1.
If the initial curing period is short
An initial curing period of two to three hours,
during which the concrete is kept at the ambient
temperature
,411 increase in temperature at a low rate of less
than 36°F (20°C) per hour
A period (depending upon the concrete strength
to be attained) during which the temperature is
held constant and below 150°F (65T)
A period during which the concrete is cooled at a
rate similar to that used for the temperature increase
‘The lo ss o f streng th in the lo ng term w ill b e
greater:
Choosing a cement, the performance of which
is adapted to the accelerated curing of concrete
(preferred is artificial Portland cement with:
C,,A s 11% and C,,SIC,S 2 3).
If the temperature increase is rapid
If the maximum temperature is high
2. C o nsistently m anu f ac tu ring c o nc rete w ith a
minimum water content and a maximum temperature of 95°F (35°C) at the time of pouring.
As an example, the short-cycle treatment used
for the B-3 Viaduct segments was the follo\~ing,
see Figure 11.3:
3. Using suffic iently rig id fo rm s to resist the
thermal expansion of the concrete in its plastic
state while heating.
Initial period of 14 hour at 95°F (35°C) (mixing
temperature)
PREHEATING
212
i
ALTERNATIVE
I
FORM
I
I
I
I
I
2 to 3H
-Y-Jw\COOLING
piNI;;
TEMPERATURE
CONSTANT
TEMPERATURE
INCREASE
PERIOD
FIGURE 11.2. Heat-treatlnent c-~c-le.
ti
STRIPPING
Concrete and Formwork for Segmental Construction
1 I .2.4
SHORT
LONG
CYCLE
CYCLE
Temperature increase of 27°F (15°C) per hour for
2 hours
.-\ c o nstant tem p eratu re o f 150° F ( 65° C ) f o r lf
hours
Figure 11.3 shows an example of- tong-cycle heat
treatment, the Conflans Bridge, which had a total
heat-treatment duration of 19 hours.
11.2.3 DI,~lESSIO,\-‘4L TOLERrllVCES
Formw ork that produces tvpical bridge box girder
segments within the following tolerances is considered to be of good quality3.4:
FORMWORK FOR SEGMENTAL
CO,\5 TR UCTIO,V
Formw ork along with its supports and foundations
must be designed to safely support all loads that
might be applied without undesired deformations
or settlements. Soil stabilization of the foundation
mav be required.
Since economical production of cast-in-place or
precast segments is based on repetitive use of the
same forms as much as possible, the formw ork
must be sturdy and special attention must be given
to construction details. Where formw ork is to be
assembled by persons other than the manufacturer
o r his rep resentativ es, p artic u lar c are m u st b e
taken with erection details and assembly instructions. All elements of the formw ork must be easy to
handle. 1-Z
Formw ork for structures of variable geometry
will need to be relatively flexible in order to allow
adaptation at the various joints. Both external and
internal forms are usually retractable in order to
leave a free working space for placing reinforcing
steel and prestressing ducts3
Special consideration must be given to those
parts of the forms that have variable dimensions.
To fac ilitate alig nm ent o r ad justm ent, sp ec ial
equipment such as turnbuckles, prefitted wedges,
screws, or hydraulic jacks should be provided.
Tendon anchors and inserts must be designed in
such a way that they remain rigidly in position
d u ring c asting . Pro jec ting anc ho rag e b lo c ks o r
o ther su c h irreg u larities sho u ld b e d etailed to
permit easy form stripping.3
If accelerated steam curing with temperatures of
the order of 130°F (55°C) is to be used, then the
deformations of the forms caused by heating and
cooling must be considered in order to prevent
cracking of the young concrete.
In general, internal vibration using needle vibrators should always be applied. External vibrators, if used, must be attached at locations that will
Width of web
2; in.
(+ 10 mm)
Depth of bottom slab
+f in. to 0 in.
(+lO mm to 0 mm)
Depth of top slab
k+ in.
(55 mm)
Overall depth of segment
?& of depth
(5 mm min.)
with f in. min.
Overall width of segment
?h of width
(5 mm min.)
with f in. min.
Length of match-cast segment
*a in.
(25 mm)
Diaphragm
ki in.
(210 mm)
dimensions
469
Technology and Construction of Segmental Bridges
470
achieve maximum efficiency of consolidation and
permit easy replacement in the case of a breakdown during casting operations. External vibration
may lead to fatigue failure in welded joints, and
regular inspection should be made to help prevent
any sudden failure of this kind.3
Paste leakage through formw ork joints must be
prevented by suitable design of joint seals. Normally this can be achieved by using a flexible sealing material. This is particularly important at the
joint face with the matching segment, where loss of
cement paste can lead to poorly formed joint surfaces and subsequent spalling and loss of matching,
requiring repair. Special attention must be given to
the junction of tendon sheathing with the forms.3l4
All form surfaces, especially welded joints in contact with the concrete, must be perfectly smooth
and free from reentrant areas, pitting, or other
discontinuities, which could entrap small volumes
of concrete and lead to spalling during form stripping.3
I I .3
Post-Tensioning
Materials
and
Ducts must have sufficient grouting inlets, shutoff valves, and drains to allow proper grouting and
to avoid accumulation of water during storage.
Vent pipes should not be spaced more than approximately 400 ft (120 m) apart.’ This spacing
may have to be reduced, depending upon the expertise of the personnel performing the grouting.
Particular attention must be paid to the qualit!
of duct connections at the joints between segments.
At the joints, accurate placing is mandator\. ~I‘he
method of duct connection depends on the type of
joint3:
Telescopic sleeves
-w id e jo ints
Screw-on
type
pushed
over
sleeves-wide
projecting
ducts
joints
Internal ru b b er o r p lastic sleev es- m atc h- c ast
joints
G a ske ts
or
other
special
seals-match-cast
joints
No special provisions: clean ducts with a torpedo
af ter jo inting to rem o v e p enetrated ep o xv if
anv - m atc h- c ast jo ints
Operations
1 1 3 . 1 GE,\‘ERAL
Technical details relating to the different methods
av ailab le are d esc rib ed in the v ario u s posttensio ning manuals5g6
and in the sp ecific d o cuments issued by suppliers.
113.2 DUCTS
Ducts are used to form the holes or enclose the
space in which the prestressing tendons are located. The ducts may be located inside or outside
the concrete section.
A ltho u g h in so m e instanc es the tend o ns are
placed in the ducts before concreting (cast-in-place
and
span-by-span
construction),
post-tensioning
tendons will normally be threaded into the ducts
af-ter erec tio n o f the seg m ents. The d u c t c ro ss
sec tio n m u st, theref o re, b e ad eq u ate to allo w
proper threading; and in general it will be about f
in. (5 mm) larger in any direction than for ducts in
which the tendons are placed before concreting.
The duct dimension must allow not only the installation of the tendons but also free passage of
grout materials after stressing. The ratio or proportion of cross-sectional area of the duct with respect to the net area of prestressing steel should
conform to appropriate specifications or codes.4 A
minimum value of 2 usually leads to satisfactory results.
Connection tightness is essential in order to prevent penetration of joint material, water, or other
liquids or solids into the ducts, which would introduce a risk of blockage, and also to prevent leakage
at the joint during tendon grouting operations.3
1 1 . 3 3 TE,VDO.V
ASCHORS
Tendon anchors usually consist of a bearing plate
and an anchorage device either in combination or
as separate units. Shape and dimensions of the anchors must conform with the applicable specifications, particularly insofar as bearing stresses are
concerned.
C ho ic e o f anc ho r p o sitio ns in the seg m ents
should ta ke into account the following considerations3:
Tendon layout
quences.
requirements
and
installation
se-
Stresses generated around the anchors.
Ease of tendon threading and stressing.
Ease of formw ork
crete placing.
preparation, stripping and con-
Certain anchorage positions, such as the anchorage
block on a thin slab shown in Figure 11.4, should
be avoided. If this type of detail cannot be avoided,
then particular care must be taken in design and
construction of the zone concerned.3
471
Post-Tensioning Materials and Operations
Tendonanchorageblocks
FIG U RE 11.4. .Anchor-qr
block
p o sitio n to he
;I\ aided.
Bearing plates are usually embedded in the segment at the time of casting. In certain cases they
are installed against the hardened surface of the
concrete with a dry mortar bed or a suitable cushioning material such as asbestos cement or synthetic resin.
struction under ordinary working conditions and
supervision. The actual results obtained in a segmental bridge built in Europe are given below by
way of example for the benefit of future project
designers.
Cantilever tendons were placed along a straight
profile in the roadway slab and anchored either on
the segment face or tn a block-out inside the box
girder. Continuity tendons were either anchored
in a block-out at the bottom slab level or draped
upward in the webs and anchored in the same
block-out of the cantilever tendons. All tendons
were made up of twelve 0.6 in. diameter strands.
Soluble oil for reducing friction in the ducts was
not allowed by the consultant. The calculations
were carried out using the following values for
curvature and wobble friction coefficients:
/ .l = 0.20,
This subject has been covered in Chapter 4 relating
to design. The choice of tendon layout must be
treated carefully, with special attention paid to the
following factors:
C o nstru c tio n seq u enc e w ith resp ec t to tend o n
placing, segment casting (or erection), and other
construction
imperatives
Standardization and repetition of essential features, especialla duct and anchor positions at joints
(in order to facilitate formw ork design)
Various loading conditions throughout
struction period and in service
the
con-
When using large tendons, it is not advisable to
use couplers or crossed splices, for reasons of congestion and formw ork complication. Also, couplers
and splices should not be located in areas where
vielding mav occur und er ultim ate lo ad conditions.3
In order to limit friction losses, and to facilitate
tend o n thread ing , exc essiv ely c u rv ed tend o ns
should be avoided if possible.
11.3.5 FRICTIO S LO SSES IS PRESTRESSISG
TESDO,VS
Segmental construction usually calls for prestressing tendons to be installed through a succession of
short duct lengths coupled to one another at the
joints between segments, these being at approximately 8 to 30 ft (2.5 to 10 m) intervals.
The friction factors (for curvature and wobble)
usually accepted for long tendons in cast-in-place
structures may not be realistic for this type of con-
K = O.OOf/ ft
= 0.0021/ m
The Young’s modulus of the tendon samples
tested in the factory or in the laboratory varied
between 28,000 and 29,000 ksi, and the variation
between various heats over the whole structure was
very low. According to direct tests carried out on
site, and a systematic analysis of all results of tendon elongations recorded during the stressing operation, the actual Young’s modulus of a (twelve
0.6 in. diameter strand) tendon at first tensioning
varied between 25,000 and 26,000 ksi, which is
only 90% of the value recorded during factory and
laboratory tests.
Figures 11.5 and 11.6 show values of the wobble
friction coefficient K measured for all the tendons
in the structure’s 18 cantilevers. All the tendons are
shown in Figure 11.5, while Figure 11.6 shows only
those tendons in the spans without hinges, and
separates the tendons anchored on the segment
face from those anchored in block-outs (the tendons had the same layout except a rather severe
curvature at the end). It is obvious that:
As construction proceeded and the quality of
manufacture and supervision improved, the
results got better.
At the beginning of the job, the effect of the
curved ends of certain tendons was lost in the
generally mediocre results. As these results got
better (value of K equal to that used in calculation from cantilever 11 on), this effect became
preponderant, counteracting that of the improved standard of work.
As the site staff became accustomed to the
work and the effort and supervision dropped,
the results became gradually worse (compare
cantilevers 13 and 17, for example).
Technology and Construction of Segmental Bridges
472
dK,
2
3
k
5
6
7
8
9
10
11
12
13
lk
15
16
17
18
M
CANTILEVER
FIG U RE 11.5. Prestressing in a cantilever bridge. Variation of uobble 1.1.ic
tion
zy
n ”
coefficient for cantilever tendons in each of the structure’s 18 spans.
1O’K
ANCHORED
IN
BLOCK - OUTS
TENDONS
THE
2
ANCHORED
AT
SEGMENT
3
5
7
9
11
13
is
17
18
w
CANTILEVER
FIGURE 11.6.
Prestressing in a cantilever bridge. Wobble friction coefficient for C;LIItilever tendons in the 10 spans without hinges.
As an example, a straight tendon in the top slab
fillet between slab and web was isolated. The wobble friction coefficient depends on the care exercised in fastening the duct to the reinforcing steel
cage as the concrete is poured (when the tendon is
in the slab rather than in the fillet, the accidental
deviations are much smaller). For the first seven
cantilevers (see Figure 11.7) the wobble coefficient
473
Post-Tensioning Materials and Operations
F L E X I B L E
1
2
3
4
5
6
+
8
L
9
zyxw
L::!!::!:
:-
10
11
12
0
U
15
16
17
M
FIGURE 11.7. Prestressing in a cantilever bridge. Wobble friction coefficient for a
straight tendon located in the upper fillet.
reached up to six times the assumed value used in
the calculations, and yet very careful construction
will enable this assumed value to be reached or at
least approached closely to obtain the desired prestress with little room for uncertainty.
The presence of hinged segments not only complicates the tendon profile and the construction
phases, but introduces uncertainty about obtaining
the required prestress force. Owing to the technical restrictions imposed by the consultant, the traditional prestress layout employed in earlier
bridges could not be used. Consequently, long tendons stressed only at the opposite end had to be
accepted. It was thought that a realistic value of the
final force for each of the tendons (twelve 0.6 in.
diameter strand) would be 350 kips (160 mt). It is
Technology and Construction of Segmental Bridges
474
fortunate that a direct check was made at the site,
which revealed the actual initial load at transfer to
be the following for the four tendons under consideration: 130 kips (60 mt), 210 kips (96 mt), 130
kips (60 mt), and 200 kips (90 mt). The average
initial prestress load per tendon was therefore 170
kips (78 mt), and the probable final force would
have been 150 kips (70 mt) as compared to the assumed value of 350 kips (160 mt). Fortunately, the
situation could be easily corrected and remedial
measures put into effect as follows:
1. The reinfo rc ing steel and lo c al p restressing
tendons allowed for a certain margin of safety.
2.
It was possible to restress two of the four cables
in the first cantilever and then to change the
p ro file and m etho d o f p lac ing seg m ents in
order to stress all the tendons at both ends for
the rest of- the cantilevers.
The above results, quoted rigorously so as to illustrate several important aspects of friction losses,
must not lead the reader to suppose that the safetv
of the structure was at any time compromised. Thk
force in a tendon varies much more slowly than
anv changes in the friction coefficients for ordinar)
tendon profiles. For example, in a 270 ft (80 m)
tend o n stressed at b o th end s, if the f ric tio n
coefficients are multiplied by 4, the minimum force
in the tendon is reduced bv only 16%.
It is interesting to examine the results for the
actual prestress obtained in cantilevers 2 and 3 (the
ones having the worst results) shown in Figure 11.8
for each section, compared with the prestress used
in the calculations. The lac k o f p restress, m o st
m arked at midspan, w as c o m p ensated by ad d itional tendons to bring the force back up to that
required by the calculations in the first two spans.
Afterward, the originally calculated prestress was
alwavs sufficient.
To summarize, the authors wish to underline the
following points:
1. Benc h tests sho u ld b e p erf o rm ed on site to
determine a realistic value of the modulus of
elasticity of the tendons to be used to compute
the theoretical tendon elongations.
2.
Realistic values of curvature and wobble friction coefficients should be used in the design
and further controlled on site. Direct friction
tests should be made together with a statistical
analvsis of the measured elongations for all
tendons.
SUPPORT
5UPPORT
MID - 5PAN
%f-
l.ax,
-
EFECTIVE
FIGURE
PRESTRESS IN
11.8.
5PAN5
2AND3
h
-
Prestressing in a cantilever bridge. Effective prestress in spans 2 and 3
1.W
Segment Fabrication for Cast-in-Place Cantilever Construction
3.
Provisions should be made at the design stage
for additional prestress to compensate for any
unexpected reduction in the design prestress
force due to excessive friction. This may be
done as follows:
a.
By adding additional ducts over and above
the number required by design calculations; if this method is used, the unused
ducts at the end of construction must be
grouted to prevent water from seeping inside and subsequently freezing with disastrous effects on the structure.
b.
By using larger than required sizes for
some of the ducts, so as to allow the use of
larger-capacity tendons if required.
By providing anchor blocks and possible
deviation saddles so as to allow the installation of external tendons located inside
the box girder but outside the concrete
section.
C.
If the correct approach is taken at the conception stage, perf-ect control of this aspect of prestress
mav be obtained and verv satisfactorv structures
can be built that give maintenance-free long-term
performance.
I I .3.h
used because it increases the moisture content of
the air and reduces the natural corrosion protection.
Another important and sometimes acute problem relates to potential grout leakage at segment
joints, which can lead to the passing of grout from
one duct to another. For this reason ducts must be
well connected and sealed at joints. To check the
grout tightness of the joints and to avoid blockages,
it is advisable to flush the ducts with water under
pressure before grouting. Any leakage points thus
detected may then be sealed. If communication is
d isc o v ered b etw een tend o n d u c ts, the tend o n
groups affected should be grouted in one operation after threading and stressing of all the tendons
involved. 3
If couplers are being used (notably for single-bar
tendons), precautions tnust be taken to limit the
risk o f g ro u t b lo c kag e at the c o u p ling p o ints.
Couplers must be housed in special enlarged enclosures with two essential features3:
I. Clear cross-sectional area for the passage of
grout equal to or greater than that for the rest
of the tendon.
2.
Independent grout inlets and vent pipes.
GROC’TI,\‘G
As in conventional post-tensioned structures, segmental construction requires the grouting of prestressing tendons after tensioning to provide corrosion protection and to develop bond between the
tend o n and the su rro u nd ing c o nc rete. C u rrent
recommendations and provisions of good practice
are theref o re ap p lic ab le to seg m ental b rid g es.
However, several important points need to be examined.
Grouting must not be carried out if the temperature in the ducts is less than 35°F (2°C) or if the
su rro u nd ing c o nc rete tem p eratu re is less than
32°F (2°C). This requirement virtually precludes
grouting operations during the winter months in
the northern and middle western United States,
unless very special winter precautions are used. It
is preferable to postpone all grouting operations
until the following spring, even though some tendons may be left tensioned and ungrouted for a
long period. Attention must then be given to corrosion protection of the high-tensile steel bars or
strands. Satisfactory protection is obtained by sealing all tendon ducts at both ends after blowing out
with cool compressed air. Hot air should not be
475
I1 3.7 UNBONDED TENDONS
Unbonded tendons may be used in segmental construction provided that the performance requirements of the post-tensioning steel are also met by
the tend o n anc ho rag e, notably with respect to
fatigue
characteristics. In u nb o nd ed p o st- tensioning a corrosion protection system must be
provided to guarantee at least the same degree of
corrosion protection as for bonded tensioning.
This may be achieved by enclosing the tendons in
flexible ducts (such as polyethylene pipes) and by
cement grouting after tensioning.
I I .4
Segment Fabrication for Cast-in-Place
Cantilever Construction
11.4.1
CONVENTIONAL
TRAVELERS
The c o nv entio nal f o rm trav eler su p p o rts the
weight of fresh concrete of the new segment by
means of longitudinal beams extending out in cantilever from the last segment in order to support
the forms and service walkways.
Form Trav elers with Top M ain Beam (Fig ure
2.83) The longitudinal main beams or girders are
usually located above the segment to be concreted,
in line with the webs. The outside forms, the bottom forms, the work floor, and the service walkways are hung from the main beams with the help
of cross beams. The inside forms are supported on
a trolley, which travels inside the deck.
The main beams are anchored to the previous
segment. In order to maintain stability during the
pouring operation a counterweight is sometimes
used to reduce the uplift forces applied to the concrete section. When the traveler is transported to
its new position ready for the next segment, the
counterweight keeps it in balance between two successive anchoring positions. The main beams that
support the load due to concrete, forms, walkways,
and so on are often subject to large deflections,
which can give rise to transverse cracking along the
joints between segments. These cracks appear at
the upper face of the bottom slab and at the connection between web and top slab. This undesirable condition can be avoided by using a rigid
structure; the weight of the traveler is increased
together with the prestress required in the cantilevers. The form traveler used for the Oissel
Bridge weighed 120 tons (110 mt) and may be considered as a heavy form traveler.
If the travelers are light, care must be taken to
compensate deflections during concreting by adjusting jacks. This type of traveler weighs (excluding counterweight) a little less than half the
maximum concrete segment weight. An example
of a light form traveler is shown in Figure 11.9 for
the Tourville Bridge. Each traveler weighs 33 tons
(30 mt).
Form Travelers with Lateral Main Beams (Figure
11.10)
Travelers with their main beams above the
bridge deck present the disadvantage of hindering
the construction operation concerning the upper
part of the segment. For this reason certain form
travelers have their main beams disposed laterally
parallel to the outside webs, underneath the bridge
deck. This solution leaves a clear working surface
and allows easy access to all surfaces to be formed,
reinforced, and concreted. In this way, the technology originally developed for precast segmental
construction can be applied to cast-in-place cantilever methods, resulting in shorter construction
cycles. The Moulin-les-Metz Bridge in eastern
France, Figure 11.11, was constructed using this
type of form traveler.
FIGURE 11.9.
eler.
1 I .4.2
I‘ourville-la-Kwiere
Bridge form trav-
SELF-SUPPORTING M OBILE FORMWORK
In the case of traditional form travelers, the resulting deflections seen during construction are
alm o st entirely d ue to the m ain b eam s. The
formwork as such usually acts only as a mold and
does not support any part of the total load, even
though it comprises very stiff walls.
In several recent bridges the traveler concept has
been modified so as to use the rigid formwork as
the weight-carrying member, thus producing a
self-supporting rigid mold. Several advantages are
gained with this concept:
Surveying control and correction of bridge deck
geometry are easily obtained.
Cracking near the joints caused by the deflection of
conventional travelers is completely eliminated.
The work area is maintained completely free and
allows prefabricated reinforcing steel cages to be
used as in precast segmental construction.
This type of mobile formwork was first used for
constant-inertia bridge decks such as the Kennedy
Bridge, Dijon, and the Canadians Interchange in
Paris, Figure 11.12.
During the concreting operations, the mobile
formwork is prestressed to the existing deck. The
exact positioning of the formwork is obtained by
Segment
Fabrication
for
Cast-in-Place
Cantilever
Construction
1 a CONCRETING PHASE
zyx
2, LAUNCHING PtlASE
FIGURE 11.10. Typical for.rn traveler with later .a1 main beams.
FIGURE 11.11. Moulins-les-Metz
form traveler.
means of adjusting pins located at the rear in reservations provided in the previously poured segments. The formwork is transported to its new position, ready for the next segment, on an overhead
trolley, which travels along short steel girders cantilevered out from the existing hardened concrete
in line w ith the w ebs.
A further refinement was to use pretensioned
reinforcing to add to the stability of the traveler
FIGURE 11.12. Canadians Viaduct (Paris), view of
form traveler in operation.
while pouring the segment. Figure I 1.13 shows the
arrangement for the Canadians Viaduct in Paris,
France. Monostrands located in the webs are provisionally anchored to the front of the traveler and
embedded in the webs of the concrete segments to
be incorporated in the reinforcement of the permanent structure.
The use of the self-supporting mobile formwork
was later extended to variable-depth bridge decks
Technology and Construction of Segmental Bridges
478
S&le S t r a n d s 1
(4x0 6”)
tic R e a r Fixation
’
iA
’
i.E
SECTION B.B
SECTION A.A
\
FIGURE 11.13.
U er Fixations
/-=-7
Sinqle S t r a n d s
Canadians
I
\
Viaduct (Paris), details of
as well as three-web cross sections, as in the Clichy,
Orleans, and Gennevilliers Bridges.
The structural members of the mobile formwork
are therefore the side forms of the exterior face of
the outside webs and the bottom forms of the
underside of the bottom slab, both of which are
stiffened transversely by front and rear frames
braced together for additional rigidity, Figure
11.14. In this manner a rigid box is formed, which
is prestressed to the existing deck. The change of
section height is achieved by vertical displacement
of the bottom forms, which are fastened to the
front stiffening framework and bottom slab of the
last segment.
The stability of the self-supporting mobile forms
of the Gennevilliers Bridge was ensured by (Figure
11.15):
1.
Two steel pins fixed to the top of the outside
forms and matching imprints provided on the
outside face of the previous segment, the connection being assured by high-strength bars
going through each web.
2.
Two steel pins fixed to the upper surface of the
bottom forms and matching the corresponding
imprints provided in the last segment bottom
slab, again held by prestress bars.
The self-weight of the mobile forms and the fresh
concrete creates an overturning moment, which is
the
Lower Fixations
’
self-supporting
form
traveler.
balanced by two forces F sustained by the previously described locating pins. Practically all the
shear force is taken by the upper pins. Because of
the large forces transmitted through the top pins
to the concrete, precast concrete elements are used
to avoid the transmission of high stresses to young
concrete, Figure 11.16. These forces are transmitted by friction between pin and concrete, and this
determines the necessary prestress force.
11.4.3 TWO-STAGE CASTING
The method of two-stage casting involves, first, the
fabrication of the bottom slab and the webs together with a small part of the top slab in order to
create a flange in which all or some of the cantilever tendons can be located. This operation, carried out using a conventional form traveler, produces either a U-shaped or a W-shaped section,
depending on the number of webs, Figure 11.17.
After the cantilever tendons are stressed the form
traveler is moved to the next position, the top is
poured using a mobile formwork of relatively simple design. This second stage usually follows the
first with a minimum interval of two or three segments, and concrete can be placed in a simple pour
over the length of several segments.
This method has the advantage of reducing the
concrete volume to be supported by the form
traveler, thus reducing the weight of the traveler.
Segment Fabrication for Cast-in-Place Cantilever Construction
bunt
a
n
d
r r ar
479
st if f r ncr s
!
‘B0 t t 0 m
Mob/ h
\ Bot t om
f 0r ms
t
r
’
u s s
f or ms
FIGURE 11.14. Self-supporting mobile forms for variable-depth bridge decks. (a)
Co ncreting. (b) Mo ving f&w ard .-
In addition, the second stage is independent with
respect to the first and so is no longer on the critical
path of concreting operations.
The bridge decks of the Saint Isidore and Magnan Viaducts on the Nizza A -8 bypass w ere constructed using this method. All of the 130 ft (40 m)
spans of the Saint Isidore Viaduct were completed
for stage one only, including closure to the preceding span, before the second stage was completed, using mobile formwork w hich rolled along
the bottom slab from one abutment to the other.
As regards the Magnan Viaduct, the second stage
followed the first with an interval of three seg-
ments, because of the long spans in this structure.
The same procedure was used for the Clichy, Joinville, and Woippy Bridges, Figure 11.18.
11.4.4
COMBINATION OF PRECAST WEBS WITH
CAST-IN-PLACE FLANGES
The preceding methods allowed a considerable reduction in the construction cycle. Two pairs of
segments could thus be completed every week, corresponding to an average rate of construction of 7
to 10 ft (2 to 3 m) per working day.
MOBILE
FORM
STABILITY
Pr est r essina
FIGURE 11.15.
b a r s
zyxw
Stability of the Genne~illiers Bridge self-supporting mobile forms.
_
PRECAST
JOINT
_
Pin
FIGURE 11.16. Precast
FIGURE 11.17.
g usse t
for Genne\illiers Bridge
Two-stage construction of a two-web bridge deck.
Segment Fabricatio n fo r Cast- in- Place Cantilev er Co nstructio n
y’
. I.
.1. .(r+.“.r.,
--,..
. ” ” .,
x -,___-_ i. j
‘1. .‘.W *&e “,~..,‘?C~~:r^~.y’~.~~
**v.,; ;a -3 - _,
_ ,
FIGURE 11.18. \Voippy Viaduct, France. Detail of the
self supporting form traveler and two-stage casting.
zyxwvuts
The main obstacle preventing further reduction
in the construction cycle and therefore a closer approach to the speed of precast segmental construction is the lack of strength of young concrete and
the consequent interference with stressing operations. Apart from several other methods already
discussed, the problem can be partially overcome
by using precast end blocks or precast webs or
both. This was first tried for the construction of the
Brotonne Viaduct approach spans, Figure 11.19.
The webs, which were rather thin and heavily inclined, were precast in pairs and pretensioned,
Figure 11.20.
The deck Was cantilevered out from the piers
using 10 ft (3 m) long segments assembled in two
phases. In the first phase, the precast webs weighing up to 18 tons ( 16 mt) were placed inside the
form traveler, previously adjusted to the bridge
profile including the desired camber. The webs
were then prestressed to the preceding segment
with provisional prestress bars, the joint being
TENSION
FIGURE 11.19. Brotonne Bridge, mobile form carrier.
match-cast or cast in place. The second phase consisted of casting the rest of the segment inside the
form traveler, which was now suspended from the
new ly stressed w ebs.
This procedure, which requires partial prefabrication of the segments using light casting equipment, enables a considerable simplification of the
form traveling equipment, the limitation of total
weight to 39 tons (35 mt), and a reduction in the
construction cycle such as to produce, even for a
cable-stayed bridge, as many as four segments per
week for each pair of form travelers.
COUPLER
TENSION
RODS
STEEL
FOR
RODS
TENSIONING
JACK
36 mm dia TENSION RODS
FORMS
ADJUSTABLE
ADJUSTABLE
BRACKETS
11-36
mm dia TENSION RODS
TENDONS
BED FOR PRETENSIONED WEBS
FIGURE 11.20. Brotonne Bridge, precasting of webs.
CASTING
BED
482
Techno lo gy and Co nstructio n
During construction of the Brotonne cablestayed bridge, the precast webs were placed by
tower crane traveling parallel to the bridge deck
above the river banks and by an overhead gantry
crane above the Seine River.
Another example of the use of precast webs is
found in the Clichy Bridge carrying the metropolitan line over the Seine in the northwest of
Paris. The bridge deck with a 280 ft (85 m)
maximum span consists of a three-web box girder
without cantilever flanges and with the deck supporting the live loads as low as possible in order to
reduce the length of the access ramps to the structure. The 8 ft (2.5 m) long segments were also constructed in two stages, Figure 11.2 1.
The precast webs, with epoxy match-cast joints,
are placed with the aid of a mobile handling system
rolling along the webs of the previously placed
segments. They are then prestressed to the existing
structure before the top and bottom slabs are
poured in place on the length of two segments.
11.4.5 PRACTICAL PROBLEM S IN CAST- IN- PLACE
C O N STR U C TI O N C A M B ER C O N TR O L
Before proceeding with the cantilever construction
proper, a starting base must first be completed on
the various piers. This first special segment, called
a pier segment or a pier table, is generally constructed on a temporary platform anchored by
FIGURE 11.21. Precast web placing equipment for
Clichy Bridge carrying the metropolitan line over the
Seine River.
of Segmental Bridges
ELEVATION
SECTION
Ro tre ssinq
b a n /
I
I
FIGURE 11.22. Construction of the pier segment for
a cast-in-place cantilever deck.
prestressing the pier top, Figure 11.22. This special segment may either be given the minimum
length to insure adequate connection to the pier
for the stability of the future cantilever or else be of
such length as to allow both travelers to be installed
simultaneously, Figure 11.23.
Another important problem relates to the safety
of the travelers during construction. Chapter 4 described the difficulties of ensuring pier safety in
the event a form traveler fell during transfer from
one position to the next. The difficulties would
even be greater in the event of an accident during
the casting operation. Consequently, all precautions must be taken both at the design stage and
during construction to eliminate this potential
hazard. The load-carrying members of the traveler
must be carefully inspected and Ray even be load
tested before use so as to practically eliminate the
danger of structural failure.
The most critical areas are in the safety of the
suspension rods and the transfer of the traveler
reactions to the concrete. Preferably all suspension
rods and anchor bars should be doubled. Also, the
prestressing tendons must have an adequate margin of safety. Use of a single strand or a single bar
in each web of the box should be avoided. Rather a
multistrand tendon with individual anchors for
each strand or two prestress bars should be used.
Worldwide use of cast-in-place cantilever construction has established an extremely good safety
record, much better than that for cast-in-place construction on fabework. Accidents are very few and
far between; however, designers and constructors
must always be safety conscious.
Segment Fabrication for Cast-in-Place Cantilever Construction
483
stress forces impose upon the cantilever a new
deflection curve.
3. Deflections of the various cantilever arms after
construction and after removal of the travelers
before continuity is achieved with the other
parts of the deck.
4. Short- and long-term deflections of the continuous structure, includ ing the effect o f
superimposed dead loads (curbs, railings,
pavements, utilities, and so on) and live loads.
5. Short- and long-term pier shortenings and
foundation settlements.
FIGURE 11.23. Stal-t of cantilever construction from
the piel- segment. (n) Short pier segment - successive
inst;illation
of travelers. (6) Long pier segment
-simultaneo us installatio n of travelers.
The most critical practical problem of cast-inplace construction is deflection control, particularly for long-span structures. There are five
categories of deflections (or space geometrical
movements of the structure) during construction
and after completion:
1.
Deflection of the travelers under the weight of
the concrete segment. This value is given by
the manufacturer or may be computed and
checked at the site during the first operations.
2. Deflection of the concrete cantilever arms
during construction. For each casting of a pair
of segments, the weight of the concrete segments and the corresponding cantilever pre-
Using the data available on concrete properties
and foundation conditions, the designer should
compute the various deflections mentioned under
items 3, 4, and 5 above, assuming the bridge unloaded for foundation settlements and long-term
concrete deflections and half the design live load
for computation of the short-term concrete deflections.
The sum of the various deflection values obtained in the successive sections of the deck allows
the construction of a camber diagram, which
should be added to the theoretical longitudinal
profile of the bridge to determine for each cantilever arm an adequate casting curve. This casting
curve is the goal toward which construction proceeds during cantilever casting. The essential
difficulty is that no absolute coordinates are available in a system where everything changes at each
construction stage (transfer of traveler, concrete
casting, or cantilever prestressing).
A very simple example may illustrate the solution of the problem of accommodating the deflections described under item 2 above. For simplicity,
assume only a four-segment cantilever arm, for
which a horizontal longitudinal profile is required,
Figure 11.24.
As outlined in Chapter 4 and summarized briefly
above, the designer analyzes the various deflection
curves for each construction step (casting segment
and precasting). The typical results are shown in
Figure 11.24. The cumulative deflection curve is
immediately obtained together with the camber diagram, Figure 11.25. The use of the camber diagram
for determining the adequate deflection at each construction stage is simple; however, it is much less
simple to use in a proper manner in the field, and
experienced surveyors have often made mistakes.
When properly used, the camber diagram allows
the determination at each joint, of offset values
such as yle2, yzm3, and y3.4 at each point, w hich w ill
484
Technology and Construction of Segmental Bridges
I
ELEVATION OF TYPICAL CANTILEVER
Downward deflection IS posltlve
I
CASTING AND
PRESTRESSING
SEGMENT
0
0
0
0
I
VERTICAL DEFLECTIONS (in mm)
&
&J
@
(-11)
(-17)
(-23)
1
-5
2
1
5
(9)
1131
3
5
10
20
(30)
4
8
16
29
49
+9
+22
41
69
TOTAL
DEFLECTION
FIGURE 11.24.
structio n stag e.
Partial deflections due to girder weight and prestressing
bring the traveler in the proper position to realize
the desired final geometry. The sketch and table in
Figure 11.26 show how to use the camber diagram
properly. It is very important to realize that at no
construction stage does the profile of the cantilever
coincide with either the final deflection curve or
the camber diagram.
The natural tendency would be to build up the
traveler to the required offset to make its nose fall
z
at each ~011.
exactly on the camber diagram. The results of this
improper procedure are shown in detail in Figure
11.27. The bridge is built with an undesired double
curvature, particularly undesirable toward the end
of the cantilever. When the mistake is discovered, it
is usually too late to put into effect any remedial
measures, because the final shape of a cantilever
d ep end s essentially u p o n the ac c u rac y o f the
geometry near the piers, where the deck is sub-
Characteristics of Precast Segments and Match-Cast Epoxy Joints
zyxwvuts
-69
/
Camber
485
curve
41
/’
,’
.' /‘
Yr 9
6 0
.LY,-;-~
zyxwvut
( a s s u m e polygonaizyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
lme 1
FIGURE 11.25. Cumulati\-e
deflection curve and choice of
camber.
b.
jetted to the highest moments and where its
deflections have the greatest effect at midspan.
2.
I I .5
Characteristics of Precast Segments
Match-Cast Epoxy Joints
and
Developed originally to allow a rapid and safe assembly of precast segments at the construction site,
the technique of match casting was progressively
refined as experience was gained. We shall describe the characteristics of segments in the early
structures to further highlight the latest improvements and variations of the original concept.
11.5.1 FIRST-GE~VERATIOS
SEGME.VTS
In those early structures the epoxy resin played
several important roles:
1. During assembly before hardening:
a. To lubricate the mating surfaces while
final positioning took place.
To compensate for minor imperfections in
the match-cast surfaces.
In the finished structure after hardening:
a. To ensure the watertightness of the joints,
especially in the top slab.
b. To participate in the structural resistance
by transmitting compression and shear
forces. However, before hardening of the
epoxy resin, the joints present no shear resistance whatsoever, because the epoxy
behaves like a perfect lubricant. It was
therefore necessary to provide shear keys
in each web in order to ensure the shearforce transfer between segments. These
keys, as well as those situated in the top
slab, also allowed a very accurate assembly
of one segment with respect to another.
During assembly of the deck, some sort of temporary fixation, either mechanical or by means of
prestress bars, allowed the placing equipment
(launching girder, crane, and so on) to be quickly
486
Technology and Construction of Segmental Bridges
zyxwvutsrq
I
1
FIGURE 11.26. Follow-up of deflections with proper
camber diagram.
unloaded without waiting for the cantilever tendons to be stressed.
Figure 11.28 shows how a typical first-generation
segment can be assembled to the existing structure
using a temporary apparatus located on the top
and bottom slabs, which is used to create forces F,
and F, which ensure the equilibrium of the new
segment at the joint.
These two forces, combined with the weight W
of the segment, give the resultant force R, w hich is
inclined with respect to the joint. Because of the
very small coefficient of friction of the epoxy, the
shearing component of R produced by W can
be balanced only by the vertical component of the
reaction C, which exists normal to the bottom face
of the web shear keys, Figure 11.28. The resultant
R is composed, therefore, of the oblique reaction C
supported by the shear keys and a horizontal reaction N, which is responsible for securing the joint.
The axial stress distribution at the joint cross
section differs in this case from what would be ob-
use
of
tained by ordinary calculations. It is obvious that N
is smaller than F (the sum of forces F, and FJ. Let
(Y be the angle of the key support faces with respect
to the horizontal; then F - N = W tan (Y, and for a
typical case of tan (Y = 0.50, F - N = W/2. Consider
a segment weighing 50 tons (45 mt), temporarily
assembled by a prestress force of 100 tons (90 mt)
located in the top slab; the axial force reduction is
25 tons (23 mt)-that is, 25% of the total applied
prestress force.
If the rate of erection of the precast segments is
sufficient to ensure the positioning of four segments before the resin in the first joint has set, then
the reduction in the effective axial force in this
joint will be 100 tons (90 mt), which more or less
corresponds to one tendon of twelve f in. diameter
strands. The same conclusion would be valid when
the permanent prestressing was used to ensure the
temporary stability of the cantilever.
In conclusion, it is recommended that this reduction of the effective prestress force be taken
Characteristics of Precast Segments and Match-Cast Epoxy Joints
.-~*-~
i-21)
487
.~~.-~~( 13) C-25)
FIGURE 11.27. Follow-up of deflections with improper use of
camber diagram.
TEMPORARY
SEGMENT
ASSEMBLY
JOINT EPUILIBRIUM
Fl
(J)
lb)
(a)
FIGURE 11.28.
equilibrium.
Temporary assembly. (a) Elevation of temporary assembly. (6) Joint
into account while verifying the cantilever resistance and stability. Failure to do so may result in
temporary joint opening, which is undesirable although not dangerous for stability.
It is also preferable to choose the intensity and
the point of application of the forces F, and F, such
as to allow the axial force N to be as close as possible to the section centroidal axis, thus ensuring a
488
Technology and Construction of Segmental Bridges
nearly uniform axial stress distribution over the
total height and hence a resin film of constant
thickness.
Permanent Assembly: Structural Importance of Epoxy
Resins As regards the final prestress tendon profiles, it was shown in Chapter 4 how the resistance
of the different cantilevers is ensured by a first
group of tendons, known as cantilever tendons,
which may be straight or curved in profile and anchored on the various segment faces. The stressing
operations remain in the critical path of construction because a new pair of segments cannot be
placed before the last pair has been stressed to the
existing cantilever, Figure 11.29.
The second group of tendons joins the different
cantilevers together and makes the structure COIItinuous. They are anchored either in block-outs in
the bottom slab or in the fillets at the junction between the top slab and the webs after upward deviation to top slab level.
The service shear forces that act upon the joints
vary according to the type and characteristics of
the structure. In variable-depth bridge decks with
draped prestressing tendons the shear stress across
the joints is usually low. In a long-span, constantdepth bridge deck with straight tendons, however,
the shear stresses at the joints can exceed 600 psi (4
MPa), as was the case in several structures mentioned in Chapter 4. A bad choice, or improper
use, of the epoxy resin can be a critical factor concerning the shear resistance of the joints, and for
this reason joints of this type require strict quality
control.
In general, the different types of epoxy resins
available have final strengths substantially exceeding that of concrete, so they do not constitute a
weak point in themselves. Several conditions must
be satisfied, however, in order that the resin cure
properly.
1. Mixing the constituents in their correct proportions.
2.
Eliminating any solvents that have a fatal effect
the propertles of the resin.
on
3. A v o id ing any flexible ad d itiv es, such as
thiokol, that greatly increase the deformabilit)
of the epoxy.
4.
Mixing
and
applying
carefully.
With respect to the last point, the surfaces to be
joined must be specially treated if the best results
are to be obtained. Comparative tests have shown
that sand blasting gives the most satisfactorv I-esuits, the surfaces being kept clean. dry, and free
from g rease d uring p lac ing . I n d am p o r rain!
weather alcohol is burnt on the joint surfaces to
eliminate surface moisture. The water present in
the concrete itself has no detrimental effect on the
performance of the resin.
It has also been established that rapid placing ot
successive segments has a favorable effect on the
properties of the resin. The additional compressive
stress applied to an epoxy joint under polymeriLation when the next segment is prestressed improves the resin’s ultimate mechanical properties.
Finally, note that in variable-height structures
the joint detailing is such that the joint plane is
not normal to the principal stress, especially at the
bottom slab level. The epoxy joint is then subjected
to shear forces that may be quite large and that can
cause failure of the bottom slab in the event of
nonpolymerization of the epoxy resin.
In addition to the precautions taken to ensure
correct curing, one may provide against the risk of
bad results by including shear keys in the bottom
slab.
FINAL SEGMENT ASSEMBLY
1 I S.2
DETAIL
A
FIGURE 11.29. Final segment assembly.
SECOND-GE,\‘EKA
TI0.V SEC;.LlE.VTS
Although the characteristics and performance of
the first structures built with match-cast joints are
not in doubt, it seems a good idea to investigate
new types of joints allowing the transmission of
shear forces without relying on the strength of
epoxy resins.
Second-generation segments do just this, being
equipped with interlocking keys in the top and
bottom slabs and in most of the height of the webs.
This configuration of shear keys at regular intervals, which improves the behavior of joints bv relieving the epoxy of its structural role, has the
Characteristics of Precast Segments and Match-Cast Epoxy Joints
advantages of simplicity and safety. This type of
segment has been used with success in several
bridges, notably the Alpine Motorways, the Saint
Andre de Cubzac Bridge, and the Sallingsund
Bridge, and more recently in several structures
in the United States such as the Long Key and
Seven Mile bridges in Florida.
Anchorage
blocks (blisters) or stiffening ribs are currently
used inside the segments for the final longitudinal
prestress anchors. The tendons, ensuring the stabilit\, and resistance of the cantilever and placed
progressively as construction proceeds, can be anchored away from the joint faces, thereby rendering the stressing operations and the segmentplacing operations independent of one another.
The ribs and anchorage blocks are generally used
to house the temporary prestress that ensures the
provisional stability of the cantilever, thus leaving
the top slab completely free.
with properties that depend upon the type of resin
and hardener used. Three grades of epoxy resin
are commonly used, depending upon the ambient
temperature range under which the resin is to be
applied:
40 to 60°F (5 to 15°C)
Fast-reacting epoxy
60 to 75°F (15 to 25°C)
Medium-fast-reacting
75 to 105°F (25 to 40°C)
epoxy
Slow -reacting
Ribs and Interior Anchorage B1ock.s
Bolted Ribs Despite the tensile strength of the
epoxy resin at a glued joint, no tensile resistance is
usually considered, as precast segmental structures
are nearly always totally prestressed and so no tensile stresses can develop across the joint. However,
we can further improve epoxied match-cast joints
by giving them a certain resistance to tension by
using bolted ribs, which ensure the continuity of
the longitudinal reinforcing steel, Figure 11.30.
11.5.3
EPOXY FOR JOAVTS
The structural importance of the thin layer of
epoxy resin forming the joint between two adjacent
precast segments was discussed in Section 11.5.1.
We now take a closer look at the physical and mechanical properties of these resins and the various
precautions to be taken to ensure satisfactory and
consistent results.
-.
Epoxy Ty pes Epoxy resin glues are made up
from two components: the epoxy resin and the
hardener. Mixing these two components in the
correct proportions gives a thermostable product
489
epoxy
1. Color The resin and the hardener must be
of clearly contrasting colors thus avoiding any confusion. When properly mixed, the final product is
to be a homogeneous gray color similar to that of
concrete.
2. Shelf life of components Both components
may be stored for up to one year, provided that the
storage temperature is kept between 50 and 70°F
(10 and 20°C). After three months’ storage it is
necessary to check that the epoxy resin shows no
sign of becoming crystalline. If it does, then special
treatment must be given to the resin, followed by
tests, before use.
3. Pot Life of the M ixed Glue The pot life of an
epoxy resin is a measure of the time interval between the mixing of the components together and
the moment when the glue becomes no longer
workable. The workability of the glue is determined by its internal temperature, depending
upon the grade of epoxy resin employed. For a 10
lb (5 kg) mix used on site, mixed under isothermic
conditions until an even color of mix is obtained,
the following results are required:
Epoxy
Workability
Limit
Temperature
Grade
5 to 15°C
40°C ( 104°F)
40°C ( 104°F)
15 to 25°C
25 to 40°C
55 to 60°C
(131 to 140°F)
The pot life must be approximately:
A mbient Temperature
Epoxy
Grade
5 to 15°C
1 5 t o 25°C
25 t o 40°C
41°F
(5°C)
50°F
( 10%)
40 min.
15 min.
59°F
( 15°C)
(20°C)
20 min.
1 5 min.
68°F
86°F
(30°C)
95°F
(35°C)
25 min.
1 8 min.
490
Technology and Construction of Segmental Bridges
BOLTED RIB JOINTS
I
FIGURE 11.30.
On site, each 10 lb (5 kg) mix of epoxy resin must
be applied to the concrete surface within the potlife period as specified above.
4. Open Time of the Applied Epoxy Glue The
open time of the glue is defined as the period between its application to the concrete surface and
the moment when it reaches its workability limit
temperature. Because of the much greater heat
dissipation from the thin layer [& to a in. (1 to 3
mm)] on the concrete surface, the applied glue
takes much longer to reach the workability limit
temperature than the mix in the pot.
The open time must never be less than one
hour, regardless of the grade used. One measuring device used to determine open time is the
Vicat’s needle shown in Figure 11.3 1. A 1 mm layer
of epoxy glue is spread onto a steel plate, and the
stopwatch is started. The time lapsed before the
needle will penetrate only 0.5 trim into the glue
layer is defined as the open time.
5. Thixotropy This characteristic gives an indication of the epoxy resin’s ability to be applied to
vertical surfaces with relative ease and yet with subsequent running. Thixotropy may be measured
using Daniel’s gauge, Figure 11.32. The gauge is
placed on a level surface with the gutter section
horizontal. The gutter is then filled with freshly
mixed resin and hardener and abruptly turned to
the upright position, as shown in the diagram. The
flow time relationship is recorded. The test should
be carried out at the maximum temperature for
which the resin is specified. A resin that flows less
than 30 mm in 10 minutes-is suitable for application to vertical concrete surfaces. Other testing
methods are available such as the sag flow apparatus according to ASTM D2730-68.
Bolted rib joints.
Other characteristics of the epoxy glue that may
be tested on site are:
The angle of internalfriction: The ease w ith w hich
the excess resin may be squeezed out of the joint
when subject to uniform pressure.
FIGURE 11.31. Open-time testing-Vicat’s needle.
Characteristics of Precast Segments and Match-Cast Epoxy Joints
491
,4X4cm,
s’
100
FIGURE 11.32.
Shrinkngp:
/
mm
.I‘hisotropy
testing-Daniel’s gauge.
Must be practically nil.
Water absorption rate and solubility in water:
Maximum permissible true water absorption12%. Maximum permissible quantity of epoxy soluble in water at 25°C (i7” F)--4%.
FIGURE 11.33.
Shear-resistance
test
ceptable ultimate shear stress at the interface is
1400 psi (10 MPa).
2. Shear Modulus The instantaneous shear
modulus (Ci) must be greater than 220,000 psi
(1500 MPa) at:
Hen t resista rice:
Minimum required value according to Mostens (DIN 53458) on week-old 10 x 15
x 120 mm test rods is 50°C (122°F).
25°C (77°F) for grade 15 to 25°C
,Mechu rricnl p)-opertie.)
40°C (104°F) for grade 25 to 40°C
1. Shenr resistance The shear resistance of the
mixed epoxy glue is determined on rectangular
concrete test specimens with the following dimensions: 1.6 x 1.6 x 6.3 in. (4 x 4 x 16 cm) w ith a
resin interface at 17” to the vertical, Figure 11.33.
The concrete test pieces are made from a highquality concrete comparable to that used in precast
segment manufacture and are c ure d under water
seven days f-rom time of casting.
After removal from the water the pieces are dabdried and the surfaces to be assembled are prepared by shot blasting, wire brushing, or other
similar methods to remove laitance. The test pieces
are then resubmerged in water for three hours,
after which they are removed and dabbed dry with
a clean cloth. The resin is then applied in a layer of
& in. (2 mm) on one surface and the test beam
clamped in an assembly that maintains a normal
pressure on the interface of 2 1 psi (0.15 MPa). The
assembly is stored for seven d a ys at a temperature
representative of the desired working conditions,
and then the test is carried out. The minimum ac-
The long-term shear modulus must be greater
than 14,500 psi (1000 MPa) after 28 days at the
same temperatures as above. Solid cylindrical test
pieces are used for measuring these values in conjunction with the easily made test apparatus shown
in Figure 11.34.
Certain epoxy resins show an excessive sensitivity to high temperatures that makes them unacceptable in warm climates. Figure 11.35 shows
comparative results of ten different resins tested
for the Rio Niteroi Bridge. It is obvious that a
product that becomes practically plastic with no
shear modulus at 60°C is completely unacceptable.
15°C (59°F) for grade 5 to 15°C
3. Tensile Bending Strength A three-p o int
bending test is carried out on a pair of glued concrete cubes with a compressive strength of 5700 psi
(400 kg/ cm2), Figure 11.36. The faces to be glued
are shot blasted, or bush hammered, so as to remove laitance. The cubes are then submerged in
water for 72 hours. When taken out of the water
the surfaces to be glued are dried simply by dabbing with a clean cloth. Immediately after the dab
492
Technology and Construction of Segmental Bridges
Dial gauge,
I
I
x1
View X-X
View from one side
FIGURE 11.34. Shear=nioclulus
test.
G (M.&)
2500
2000
1500
1000
500
0
20
30
FIGURE 11.35.
40
50
Variation of’ shear ~nodulus
drying the glue is applied in a layer of & in. (1.5
m m ) to o ne o f the p rep ared f ac es. The c o rresponding face of the other cube is placed against
the glue layer, and the two cubes are clamped together with a clamping force of 300 lb (150 kg).
The assembly is then wrapped in a damp cloth,
w hic h m u st b e kep t w et u ntil the three- p o int
bending test is carried out.
4. Compressive
S t r e n g t h The
compressive
strength is determined according to DIN 1164 on 4
cm (14 in.) cubes of cured epoxy glue. After 24
hours (from the time of preparing the samples) at
the maximum temperatures for each grade the
compressive strength must be not less than 12,000
psi (80 MPa). The loading rate is to be approximately 3600 psi (25 MPa) per minute.
5. Elastic modulus in compression The instantaneous modulus (Ei) is determined on cubes of
p u re ep o xy af ter c u ring f o r sev en d ay s at the
60
m
tzyxwvutsrqponml
(‘Cl
G with temperature.
maximum group temperature. These cubes are the
sam e siz e as tho se u sed f o r the compressivestrength determinations. The modulus must not be
less than 1,140,OOO
psi (7850 MPa)
Practical Use of Epoxy in Match-Cast Joint.\
In
regard to the use of the resin, the two components should be mixed carefully and quickly as
near as possible to the surfaces to be coated. Under
no circumstances should oil or grease be allowed to
c o m e into c o ntac t w ith su rf ac es that are to b e
glued. Most standard demolding agents are suitable for use, but care should be taken to ensure that
no o il- b ased d em o ld ers are u sed . Exp o su re to
w eather d u ring the sto rag e p erio d is o f ten
sufficient to remove the demolding agent. For best
results, surface laitance should be removed by shot
blasting or bush hammering. This treatment is
normally carried out in the storage yard. With the
use of multiple keys, the structural role of the
Manufacture of Precast Segments
,L o a d
amlied
here
,
,
zyxw
p--j
-~ -___
FIGURE 11.36. Tensile bending-strength test.
epoxy is considerably reduced and a special preparation of the surface is not a mandatory feature.
Immediately before the glue is applied, the surfaces are to be cleaned to remove traces of dirt,
grease or oil, and dust.
Under normal climatic conditions it will not alwavs be possible to avoid dampness on the surfaces
to be glued. If the surfaces do show signs of moisture, they must be dab dried with a clean cloth, and
no gluing may proceed until all free water has been
eliminated.
The thickness of the glue layer should be about
h in. (1.5 mm). As soon as possible after the resin
has been applied, the surfaces must be brought together. Pressure must be applied before the open
time o f the epox) resin expires. The pressure
applied by either temporary or final prestress
should not be less than 30 psi (0.2 MPa).
11.6
Manufacture
of
Precast
ing along the bed for the successive casting operations.
2. Short-line casting (with either horizontal or
vertical casting), where segments are manufactured in a step-by-step procedure with the
forms maintained at a stationary position.
For match-cast joint structures, the accuracy of
the segment geometry is an absolute priority. Adequate surveying methods and equipment must be
used to ensure an accurate follow-up of the
geometry and an independent verification of all
measurements and adjustments.
Immediately after the manufacture of a segment
the as-cast geometry should be controlled and
compared to the theoretical geometry to allow any
necessary adjustment to be incorporated in subsequent casting operations. This aspect of match
casting is particularly important for the short-line
method and will be covered later in this chapter.
Segments
11 h.2 LONG-LINE CASTLX’G
11.6.1 1,VTRODUCTION
The various methods used until now for precasting
segments fall into two basic categories:
1.
Long-line casting, where all segments to make
up either half or a full cantilever are manufactured on a fixed bed with the formwork mov-
In this method all the segments are cast, in their
correct relative position, on a casting bed that
exactly reproduces the profile of the structure with
allowance for camber. One or more formwork
units travel along this line and are guided by a
preadjusted soffit. With this method the joint surfaces are invariably cast in a vertical position.
494
Technology and Construction of Segmental Bridges
Figure 11.37 shows the casting sequence.3 The
pier segment (3) is cast first, then the segments on
either side of the pier segment (1) and (2). If a pair
of forms is used, then the symmetrical segments on
each side of the pier segment can be cast simultaneously, thus saving casting time. As segment
casting progresses, the initial segments may be removed for storage, leaving the center portion of
the casting bed free. If enough forms are available,
then the casting of a second pair of cantilevers may
proceed even though the first pair is not completely cast.
Figure 11.38 shows the typical cross section of a
long-line casting bed with the formwork in operation. The method was initially developed for
constant-depth box girders (Choisy-le-Roi and
Courbevoie Bridges). It was later extended to the
case of variable-depth decks such as the Oleron
Viaduct (the two sketches of Figures 11.37 and
11.38 refer to this structure) and also adopted in
other countries (Hartel Bridge in Holland).
The important advantages of the long-line casting method are:
It is easy to set out and control the deck geometry.
After form stripping, it is not necessary to immediately transfer the segments to the storage area
in order to continue casting.
The disadvantages are:
Substantial space may be required. The minimum
length is usually slightly more than half the length
of the longest span of the structure, but it depends
upon the geometry and the svmmetrv of the
structure.
The
that
the
line
casting bed must be built on a firm foundation
will not settle or deflect under the weight of
segments. If the structure is curved, the long
must accommodate this curvature.
All equipment necessary for casting, curing, and so
on must be mobile.
11.6.3 SHORT-LIIVE
HORIZOh’TAL
The short-line casting method requires all segments to be cast in the same place, using stationary
forms, and against the previously cast segment in
order to obtain a match-cast joint. After casting
and initial curing, the previously cast segment is
Segments completed.
Segments being cast
FIGURE 11.37. Typical long-line precasting bed.
Travelling
crane leg
Mobile outside
r formwork
II
FIGURE
11.38.
CASTI.YG
Telescopic inside
f-form work
Typical cross section of long-line casting bed with formwork.
495
Manufacture of Precast Segments
removed for storage and the freshly cast segment is
moved into its place. The casting cycle is then repeated. This operation is illustrated in Figures
11.39 and 1 1.40.“*4
It is important that the reader fully comprehend
the principle of the method insofar as building a
deck of a given geometry is concerned. When a
straight box is desired, Figure 11.41, the match
marking mate segment (n - 1) is moved from the
casting position to the match-cast position along a
straight line, and this is usually verified by taking
measurements on four elevation bolts (a) embedded in the concrete roadway slab and two
alignment stirrups (b) located along the box centerline. A pure translation of each segment between the cast and match-cast positions therefore
results in the construction of a perfectly straight
bridge (both in elevation and in plan view), within
the accuracy of the measurements made at the
casting site.
To obtain a bridge with a vertical curve, the
match-cast segment (n - 1) must first be translated
from its original position and then give a small rotation in the vertical plane (angle CY shown in Figure 11.42). Usually the bulkhead is left in a fixed
position, and all segments have in elevation the
shape of a rectangular trapezoid with the tapered
face along the match-catch segment. It is therefore
only necessary to adjust the soffit of the cast seg-
z
BLANK END
TO
STORAGE
/
ment during the adjustment operations.
FIGURE 11.39.
t i on.
‘I‘ypical short-line precasting opera-
A curve in the horizontal plane is obtained in the
same fashion, Figure 11.43, by first moving the
match-cast segment (n - 1) to its position by a pure
translation followed by a rotation of a small angle p
in plan to realize the desired curvature.
TO
STORAGE
4
/,
FIGURE 11.40. Formwork used in casting segments.
ELEVATION
TRANSVERSE
STRAIGHT
SECTION
BRIDGE
PLAN VIEW
FIGURE 11.41. Straight bridge.
ELEVATION
_------e-------A-
TRANSVERSE SECTION
m-e----1-------;
BRIDGE WITH
VERTICAL CURVE
PLAN VIEW
FIGURE
496
11.42.
Bridge with vel-tical curve.
Manufacture of Precast Segments
497
TRANSVERSE SECTION
ELEVATION
BRIDGE WITH
HORIZONTAL CURVE
PLAN VIEW
FIGURE 11.43. Bridge with horizontal curve.
Change in the superelevation of the bridge may
also be achieved with a short-line casting; however,
the principle is a little more difficult to properly
grasp, Figures 11.44 and 11.45. A constant transverse fall of the bridge does not need to be repeated in the casting machine. Segments may be
cast with soffit and roadwav slab both horizontal
and placed at their proper attitude in the bridge by
offsetting the bearing elevation under the webs to
o b tain the d esired cro ss fall. O nly a v ariab le
superelevation must be accounted for in the casting operation, and this is the no rm al case in
bridges with reverse curves and in transition areas
between curves and straight alignments. In such a
case match-cast segment (n - 1) needs to be rotated by a small angle such as y around the bridge
centerline. Because the bridge geometry is usually
defined at roadway level and not at soffit level, the
rotation given to the match-cast segment results in
a slight horizontal displacement of the soffit in the
casting
machine, which must be accounted for.
Also all surfaces of the box segment (top slab,
soffit, and webs) are no longer true planes but are
slightly warped. To allow the formw ork panels to
adjust to this change of shape, it is absolutely mandatory to eliminate all restraints such as closed torsionally stiff members.
The basic advantages of the short-line casting
method are therefore the relatively small space req uired and the fac t that all eq uip m ent and
formw ork rem ain at a statio nary p o sitio n. The
mobility of equipment necessary for the long-line
method is no longer needed. Also, horizontal and
vertical curves as well as variable superelevation
are obtained with short-line casting without the
major change in soffit configuration that would be
required in the long-line casting method. However, success will depend upon the accuracy of adjustment of the match-cast segments, and precise
survey and control procedures must be initiated
(Section 11.6.5). This last aspect represents the
major potential disadvantage as a direct consequence of the intrinsic potential of the method.
11.6.4
SHORT-LINE
VERTICAL
CASTING
N o rm ally , f o r b o th the lo ng - and sho rt- line
methods, the segments are cast in a horizontal position. A variation in the short-line method is that
u sed f o r the A lp ine M o to rw ay s near Ly o ns,
France, where the segments were cast in a vertical
position (cast on end) as shown in Figure 3.100.
The procedure is as follows: after the first segment
is cast, the forms are removed and moved upward
498
Technology and Construction of Segmental Bridges
ELEVATION
TRANSVERSE
SECTION
BRIDGE WITH VARIABLE
SUPER ELEVATION
PLAN VIEW
FIGURE 11.44. Short-line casting-bridge with variable super-elevation.
\
END BULKHEAD
FIGURE 11.45. Short-line casting-isometric view of segment casting with variable
sup erelev atio n.
so that each succeeding segment can be cast above
the previous one. A f ter a seg m ent is c ast and
cured, the segment beneath it is transferred to
storage and the one removed from the forms is
moved down, to rest on the floor. The advantages
claimed for vertical match casting include easier
placing and vibration of the concrete. However,
special handling equipment and procedures are
required to rotate the segment from the vertical to
its final horizontal position.
Manufacture of Precast Segments
11.6.5
499
GEOMETRY AND SURVEY CONTROL
Segment Precasting in a Casting M achine
The principles described in this section apply to
short-line horizontal casting but may be easily extended to vertical casting. The apparatus used to
form the concrete segment is usually referred to as
a casting machine and is made up essentially of five
components:
1.
The bulkhead that forms the front section of
the segment.
2. The match-cast segment, properly coated at
the front end section with a suitable demolding
agent and used to form the back end section of
the newly cast segment.
3.
The mold bottom (or soffit).
4.
The side forms, properly hinged for stripping
and firmly sealed to the bulkhead and the
match-cast segment during casting. The inside
forms, which pivot and retract for stripping.
5.
The inside forms, which piv.ot and retract for
stripping.
The relationship between an individual segment
and the finished structure is established by means
of three different systems of reference:
1.
The final sy stem of reference, which is the refer-
ence for the finished geometry of the structure. In this system each segment is described
by its basic geometry.
2. The auxilia? system of reference, which corresponds to the precasting machine and is attached thereto.
3.
The elementa reference system, w hich is attached
to each segment and would be the equivalent
of intrinsic coordinates in space geometry.
The principle of the precasting method is as
follows. During the casting of segment A (segment
B being in the match-cast position) the elementary
reference system of A is identical with the auxiliary
reference system, that of the casting machine.
To position B with respect to A becomes simply a
matter of positioning B with respect to the precasting machine. It is the task of the design office to
provide the theoretical geometric information necessary for positioning. The values are computed
from the basic geometry with the addition of the
relevant compensatory values for deflections. The
definitions of these reference systems are presented below.
FIGURE 11.46. Auxiliary reference system (castingmachine
reference).
The auxiliary reference system refers to the
casting machine and is defined in Figure 11.46.
The plane of the bulkhead is perfectly vertical.
The upper edge of the bulkhead is a horizontal in
this plane except when segments do not have planar top surfaces. The x, y and z axes refer to the
casting-machine reference system, whereas XA, yA9
and z,., refer to the elementary system of reference.
The elementary system of reference is materialized
on each segment in the following manner:
1
2.
3.
This axis is represented by marks
(such as saw cuts) made on two steel stirrups
anchored in the top slab as near as possible to
the joints.
The origin 0,:
The origin o, is located at the
point where the x, axis intersects the plane of
the joint at the bulkhead.
The x, axis:
This plane may be defined
by three fixed leveling points, the position of
each point with respect to the plane x, o, y being
arbitrary but invariable. For practical reasons,
four leveling points are used and materialized
by bolts anchored in the top surface of the
segment above the webs and as close as possible
to the joints.
Theplunex,,o,,y ,:
Now that the elementary system of reference has
been established (all measurements and readings
being made while the segment is in the casting
machine before the forms are removed), the segment can be positioned with respect to the auxiliary reference system, so that it can be placed in
the correct countercasting position according to
the calculations supplied by the design office.
500
Technology and Construction of Segmental Bridges
In order to correctly position the countercasting
segment, information is needed about the final
geometry of the structure. The overall geometry of
a bridge structure is normally defined by the
geometry of the roadway. From this roadway
geometry it is necessary to determine the geometry
of the concrete structure itself.
The longitudinal reference line to which all the
necessary parameters are related is known as the
box girder line (BGL). This line may coincide with
the top concrete surface of the box girder, but it
may also be a fictitious line of reference if the box
girder top slab shape is not regular.
The box girder line is usually described using
two curves, Figure 11.47:
zontal plane and follow the curvilinear abscissas.
The segment lengths chosen on this basis may be
retained, but in calculating the real lengths of
cast-in-place closure joints and three-dimensional s
curve must be used.
Because of the way a casting machine works, the
segment joint at the bulkhead end is invariably
perpendicular to the axis of the segment. Therefore, in plan view, the segments are generally of
trapezoidal shape, except for segments over the
piers which are rectangular in order to provide a
constant starting point for each cantilever, Figure
11.48.
One curve (a) in a horizontal plane, w hich gives y
as a function of x for each point where the box
girder line intersects a joint plane between segments and also the center points of supports (abutments or piers); this curve is simply the projection
of the true space box girder line onto a horizontal
plane and is sometimes referred to as the “ bgl”
(small letters).
All measurements on a segment are made when
the seg m ent is still in the c asting m ac hine.
Readings must be taken when the concrete has
hardened and before formwork stripping, Figure
11.49. Horizontal alignment readings give the distance of the segment axes as marked on the stirrups from the casting-machine reference line.
Longitudinal profile level readings are given by the
four bolt elevations relative to the horizontal reference plane.
Readings must be taken on the segment just cast
and also on the match-cast segment. Corrections
are applied to allow for the geometric defects in
the preceding segment, Figure 11.50, and are used
as “ theoretical values for adjustment.”
One curve (5) in a developed vertical plane giving z
as a function of c for the same points mentioned
above. Thiss curve is the real box girder line, BGL.
To complete the definition of the segment position in space-at each joint line and at support
centers-we must define the transverse slope of
the theoretical extrados line.
It is important for both the bgl and the BGL to
calculate the m and s parameters, respectively, in
order to obtain an accurate determination of projected and real span lengths.
The calculations and structural drawings refer to
nominal segment lengths and span lengths. Usually these lengths refer to the projection on a hori-
Segment Casting Parameters
zyxwvu
Survqr
Control During Precasting Operations
The surveyor in charge of the operations must
complete a data sheet for each segment containing
essentially:
1. Theoretical basic data supplied by the design
office, allowing the preparation of the horizontal alignment and the two parallel bolt
lines.
2.
Bo x g irde r line
Corrected values defined either graphically or
by computer.
3. Survey control readings.
4. Linear measurements on the segments.
5. Schematic representation of the segment to
rapidly verify the relative positions of the segment axes.
Ho rizo nta l pqc c tio n
o f b o x g irde r line
bgl’
6.
A level check to pick up any gross error in level
readings on the same segment.
7.
Comments on the casting operations.
l
FIGURE 11.47. Bo x girder line curves.
501
Manufacture of Precast Segments
( Segment ax is
Segment
over
/
Hj+vi*3
pier
Pier
f
\
- BOX girder line “bgl*
or Q ( sigma ) curve
FIGURE 11.48.
Short-line casting-position of segment joints in plan view.
Ew
i’ d w
!Esw
zyxwvu
L
1dsw
. A!
?-r
L3
FIGURE 11.49. Casting-machine orientation and segment
measurements.
As an example, Figure 11.5 1 shows the typical
survey control made on the first four segments of a
typical cantilever. Control of alignment and levels
may be followed graphically or numerically by
computer, using the basic geometric data obtained
in the casting machine and show n in Figure 11.52.
In order to avoid any significant deviation from
the theoretical geometry, it is necessary to provide
for corrections when casting the next segment.
Figure 11.53 shows how this would be done for the
plan alignment. Similar corrections are made for
the longitudinal profile on the two parallel bolt
lines. It is essential not only to follow carefully the
trajectory of the two bolt lines separately but also to
check for each segment that the superelevation
(given by the crosswise difference in level between
Technology and Construction of Segmental Bridges
502
FIGURE 11.50. Plan view of’ casting operationreadings using survev
instruments.
FIGURE
11.51.
Casting
operation-topical
the two bolt lines) varies regularly according to the
theoretical geometry. Failure to do so has resulted
in important geometric imperfections on certain
projects.
Suruey
Control During Construction
The nature of match-cast segmental construction is
such that the structure is really “ built” in the precasting yard. Although corrections can be made in
the field, such corrections are undesirable and alw ays a source of additional expense and delays.
Close control of precasting is far more efficient. It
is nevertheless important to check the evolution of
the structural geometry during segment placing:
1.
To compare actual deflections with computed
values,
2.
survey
control
To ensure that no major errors have escaped
the control in the precast yard or factorv.
Such checks at the site should include:
1.
Pier positions, height and in plan.
2.
Bearing positions, level and orientation.
3.
Pier segments, level and orientation.
4. Cantilevers proper, every third segment, including levels, superelevation, and orientation.
5. Overall geometry of the structure after continuity is achieved between the individual cantilevers.
Conclukon
The principles of geometry and survey control are
more complicated to explain than to use, once the
503
Manufacture of Precast Segments
THEORETICAL AX15
OFO,
THEORETICALAXIS
OFO,
cAsTI Nt M~CI~NE
IHG
SEGMCNTO
’
BULKHEAD
QGMENT~
t
_ _
MACHINE
TUEORETltnL
-;--- -.--~~Z?-~
LEVEL FUR 1
+L;W’: ;R yyoE;;
= .
THEORETICnL,
LEVEL
FOR
0
R E A L LEVELFOR
SEGMENT 0
055TING M A C H I N E EUJLKHMD
sEGMENTI
FIGURE 11.52. Survey control-horizontal alignment and longitudinal profile results.
((I) Horizontal alignment. (h) Longitudinal profile.
REAL AXIS OF 1
TUE ORETICAL
AXIS
OF 0 (2 REAL AXIS)
FIGURE
11.53.
Typical alignment corrections during casting operations.
basic principles of a casting machine are thoroughly understood. The short-line method has
great potential to construct segments for bridges,
even those with very complicated trajectories,
rapidly and economically. Outstanding examples
are the Chillon and St. Cloud Viaducts in Europe
and Linn Cove Viaduct in the United States. At
Saint Cloud, 120- to 140-ton segments were cast
on a one-day cycle, and the final geometry of the
bridge was obtained with no on-site adjustment.
Technology and Construction of Segmental Bridges
504
O n the o ther hand , a lo o se ap p ro ac h to
geometry control at the casting yard may lead to
serious difficulties at the project site.
11.66 PRECASTING YARD AND FACTORIES
The precasting operations are usually carried out
in a yard or even a factory if the size of the project
allows the corresponding investment. All operations, such as:
Prep aratio n o f the reinf o rc ing steel c ag es and
ducts for post-tensioning tendons
Manufacture
of
concrete
Manufacture of segments including heat curing
Storage of segments including finishing and quality control are performed in a repetitive fashion
under factory conditions.
As an example of typical precasting-yard layouts, Figures 11.54 and 11.55 show views of:
The Saint Clo ud V iad uct p recasting y ard w ith
short-line casting
The Oleron Viaduct precasting yard with long-line
casting
The ty p ic al p rec asting c y c le (w ith either the
long-line or the short-line method) is of one segment per formw ork per day with a one-day work
shift, concrete hardening taking place during the
night (at least 14 hours between the completion of
concrete placing in the evening and the stripping
of forms the next morning). Shorter construction
cycles may be obtained by reducing the time of
concrete hardening, but quality may decline if all
the operations are not kept under very strict control.
Heat curing of the concrete to reduce the construction cycle and accelerate the rotation of the
casting machines is perfectly acceptable. Its improper use, however, may alter the accuracy of
joint matching between segments, as shown in Figure 11.56. This effect w o uld be p articularly
significant for wide but short segments.
Typical segments usually have the following dimensions:
Width
30 to 40 ft (9 to 12 m)
Length
10 to 12 ft (3 to 3.6 m)
Ratio
3 to 3.5
width/ length
In the case of wide decks or long spans, where the
seg m ent leng th is red u c ed to red u c e the u nit
weight, the usual geometric proportions may vary
c o nsid erab ly ; su c h is the c ase fo r tw o no tab le
structures:
St. Cloud
width 70 ft. length 7 ft,
ratio 10
St. Andre de Cubzac
width 58 ft, length 5.8 ft.
ratio 10
For such segments, heat curing is more likely to
create small changes in the segment shape, which
may build up progressively and so alter the effectiveness of joint matching. This is due to the
d ev elo p m ent o f a tem p eratu re g rad ient in the
match-cast segment, which is in contact on one
side with the newly cast heated segment and on
the other side with the lower outside temperature.
The problem may be completely eliminated by
always heat curing both segments simultaneously
so as to avoid any temperature gradient. Experience has proved the method totally efficient.
When the project involving segment precasting
is of sufficient magnitude or where climatic conditions are adverse, precasting factories are a logical
extrap o latio n f ro m the sho rt- line m etho d p erformed in an open precasting yard. Segment manu fac tu re takes p lac e in a c o m p letely enc lo sed
building with a better use of personnel and a more
consistent quality of products.
An interesting example is afforded by the B-3
South Viaducts, requiring production of 2200 precast segments weighing between 28 and 58 tons (25
to 53 mt). The precasting site was installed close to
the project and included four main areas:
1.
An assembly workshop, where the reinforcing
steel cages were prepared and the prestressing
ducts
positioned. The finished c ag es w ere
handled by a 5 ton tower crane.
2. A
3.
concrete mixing
plant.
A precasting factory where the segments were
cast and cured.
4. A storage area where the finished segments
were left to cure adequately. These segments
were handled by a traveling portal crane.
The precasting factory was equipped with four
precasting
machines, all of which were entirely
p ro tec ted f ro m the o u tsid e env iro nm ent. Tw o
machines were reserved for the manufacture of 15
to 20 ft (4.5 to 6 m) segments and two for the 20 to
PRECASTING
YARD
S c a l e 11500
Launching
1.
track
for
tr,irder
a n d trolle!y
10
Rplnforcernent
3. Loading
4.
\
I/
_ 27
“\ ‘‘\
zyx
,’ ’ _/’
,’ ,/’’ /’
,A’
zyxwvutsrqponmlkjih
Access ramp.
2
\
asspmbty
point
Launching
zone
/
0%
0
J
for s e g m e n t s
a sse m bly
g\ rdetzyxwvutsrqponmlkjihgfedcbaZYXWVUTSRQPONMLKJIHGFEDCBA
5
Segment
storage.
6.
Travelllng
I.
hlauld
8
r’rPstress1ng
crane t r a c k
bottom
steel sturaqe
17 F u t u r e carriageway
alignment
FIGURE 11.54. St. Cloud Viaduct, precasting yard layout. (1) Launching track for
girder and trolley. (2) Access ramp. (3) Loading point for segments. (4) Launchinggirder assembly zone. (5) Segment storage. (6) Traveling crane track. (7) Mold bottom.
(8) Prestressing steel storage. (9) Tower crane track. (10) Reinforcement assembly. (11)
Concrete plant. (12) Precast elements. (13) Prestress tendon manufacture. (14) Offices.
(15) General services. (16) Toll gate position. (17). Future carriageway alignment.
:‘i
Technology and Construction of Segmental Bedges
506
Staff
quarters
Launching
girder
assembly
Retnforcement
i Duct storage area
FIGURE 11.55. Oberon
L-l
L
FIGURE
L
OffIce
precasting yard layout.
zyxwv
SEGMENT
LENGTH
\‘iatiuct,
11.56.
CONJUGATE
E&CT OF IMPROPER CURING OF SEGMENTS
IN SHORT LINE CASTING
Effect of improper curing of segments in short-line casting.
31 ft. (6 to 9.5 m) segments, Figures 11.57 and 11.58.
Each casting machine was made up of a mobile
form, an end form or bulkhead, two hinged outside forms, and a telescopic inside form, Figure
11.59. Handling of concrete and reinforcing steel
inside the factory was performed by two 10 ton
travel cranes.
The production of the different segments involved the following operations:
1.
Assembly of the steel cages in a template.
2. Steel-cage storage.
3. Final steel-cage preparation and duct installation.
Handling and Temporary Assembly of Precast Segments
FIGURE 11.57.
Casting
R-3 Sot~rh
507
Viaducts, inside view of the precasting f’acto~~~.
machine
Concrete
Plant
Control
System
/
FIGURE 11.58.
Inside
B-3 South Viaducts, plan view of the precasting factory.
formwork
t
of the newly cast segment to the match-cast position by means of an independent motorized
trolley.
1 I .7 Handling and Tempera y Assembly
of Precast Segments
O u t s i d e formwork
FIGURE 11.59.
Bottom
formwork
B-3 South Viaducts, detail of a casting
machine.
4. Positioning of steel cage inside the formwork.
5. Adjustment of casting machine, including
alignment of match-cast segment and sealing
of all form panels.
6. Concrete casting and finishing.
7. Steam curing.
8. Formwork stripping; followed by transfer of
the match-cast segment to the storage yard and
In either long- or short-line casting, segments cannot be handled before the concrete has reached a
sufficient strength to prevent:
Spalling of edges and keys
Cracking of the parts of the segment subjected to
appreciable bending stresses due to self-weight
Inelastic deformations that would ultimately impair proper matching of the segments
Critical sections in a typical single-cell box segment
are, Figure 11.60:
508
Technology and Construction of Segmental Bridges
I
FIGURE 11.60. Critical sections in a typical segment at time of formwork stripping.
Section A where the side cantilevers are attached to
the webs
Sections B and C at midspan of the top and bottom
slab
Section A is almost always the most critical. Section
B is usually subjected to moderate tensile stress because the top slab is built-in on the web when the
inner formwork is stripped. Section C is critical
only on long-line casting when the casting bed does
not have a continuous soffit and when the span of
the bottom slab is larger than 16 to 20 ft (5 to 6 m).
Experience has shown that at the time of form
stripping and before any handling of the segment
is allowed, the tensile cracking strength of the concrete should be at least equal to the bending stress
due to the segment weight in the most critical sections (A, B, and C). Practically, the corresponding
compressive strength is:
f:i = 3000 to 4000 psi (21 to 28 MPa)
In the casting yard, segments are usually handled
by a portal crane traveling on rails or on steering
wheels for added mobility. A typical portal crane in
the Oleron Viaduct precasting yard is shown in
Figure 11.61.
Proper handling of the segment requires proper
pick-up points to keep the stresses in the section
within the allowable limits. A typical example of
handling three different shapes of box girders is
shown in Figure 11.62.
For the conventional single box, inserts or through
holes are provided near the web in the roadway
slab, allowing lifting to be accomplished by a simple
spreader beam.
FIGURE 11.61. Oleron Viaduct, portal crane in precasting yard.
For the twin-box, three-web section, a four-point
pick-up is usually necessary to eliminate excessive
transverse bending of the top and bottom slab. A
triple spreader-beam arrangement allows the load
transfer from the four pick-up points to the single
lifting hook.
For a triple-box, four-web section (such as used in
the Saint Cloud Bridge), temporary ties are provided in the outer cells to transfer the reaction of
the outside webs to the center webs. A simple
spreader beam is then sufficient to lift the segment.
Segments must be stored in a manner designed
to eliminate warping or secondary stresses. Concrete beams installed at ground level provide a
good bearing for the segments, which must be
supported under the web or very close thereto. If
stacking is required to save storage space, precautions must be taken to transfer weight from the
Placing Precast Segments
Slmnle
509
spreader beam
Handling precast segments. (a) Two-web segment. (6) Three-web segment. (c) Four-web segment.
FIGURE 11.62.
upper to the lower layers of segments without excessive bending of the slab.
11.8 Placing Precast Segments
Transportation and placement of segments may be
performed by one of several methods, depending
on the site location and the general characteristics
of the structure. These methods can be divided
into three main categories:
1. Transportation by land or water and placetnent by an independent lifting apparatus.
2.
Transportation by land or water and placement with the help of a beam and winch carried
by the bridge de c k itself.
3. Transportation by land, water, or along the
bridge deck already constructed and placement with the help of a launching girder.
There are tnethods that fall into none of these categories, such as the use of a cableway, but their use
is limited.
11.8.1 INDEPENDENT LIFTING EQUIPMENT
This method, where feasible, is the simplest and
least expensive. It was used for the Choisy-le-Roi,
Courbevoie, Juvisy, and Conflans bridges, where
the navigable stretch of water lent itself to the use
of a barge-mounted crane, ensuring the collection
of segments from the precasting site and their positioning in the final structure. A terrestrial crane
was employed for the Gardon, Bourg-Saint Andeal, and Bonpas Bridges. The same crane, maneuvering either on land or over water (on a
barge), assured the positioning of all the segments
used to construct the upstream and downstream
bridges of the Paris Ring Road.
When site conditions are suitable, the same lifting crane may be used both to serve the precasting
yard and to transport the segments to their final
position in the structure (Hartel Bridge, Holland).
This principle w as enlarged successfully during the
construction of the bridges over the Loire River at
Tours (Motorway Bridge and Mirabeau Bridge),
where the segments were placed with the aid of a
mobile portal frame. The portal frame is placed
astride the bridge de c k and moves along a track
supported by two bailey bridges, one either side of
the structure. The track length is approximately
twice that of the typical span, and the track itself is
moved forward progressively as construction proceeds. The bailey bridges are supported on temporary piers driven into the river bed. The segments
are first brought to the bridge deck and then taken
by the mobile portal frame, which transports them
to their final position in the finished structure, Figure 1.47.
Where a mobile truck or crawler crane is used
for placement, there are often difficulties in the
510
Technology and Construction of Segmental Bridges
positioning of the key segments at midspan, because the finished structure on either side of the
key segment prevents the crane from maneuvering
properly and hinders the positioning of the segment, which may be carried out only from the side
of the structure. For the B-3 Motorway Bridges a
special apparatus was designed to place those segments in the cantilever arm to be constructed in the
direction of the completed structure, Figure 3.95.
Two longitudinal girders are braced together and
rested on the pier head of the cantilever to be constructed at the front, and on the existing structure
at the rear. The apparatus consists of a mobile
winch-trolley, ensuring the hoisting and positioning o f the seg m ents, and an ad v anc ing tro lley
situated at the rear and equipped with a translation
motor. The front and rear supports are conceived
in such a manner as to transmit the vertical loads
through the segment w ebs.
The segments on the other side of the cantilever
are easily placed by the mobile crane. This beam
may easily be used to ensure cantilever stability
d u ring c o nstru c tio n w hen the p iers are no t
sufficiently rigid to support unsymmetrical loading. The cantilever is rigidly fixed to the girders by
clamping bars capable of resisting both tension and
compression. The crane and the girders, used together, will allow a 130 ft (40 m) span to be erected
in four working days.
Placement of segments with a mobile crane has
found another application in the construction of
small-span structures such as three-span motorway
o v erp asses (see the d isc u ssio n o f the A lp ine
Motorway, Section 3.15, and Figure 3.103). The
segments are precast in a central factory, transported to the various sites by road and positioned
b y a m o b ile c rane ac c o rd ing to the erec tio n
scheme, which consists essentially of the following:
Two temporary adjustable props, easily dismountable, placed at the one-fourth and three-fourths
points of the central span.
Temporary supports with jacks allowing cantilever
construction
Temporary prestress to tie the segments together
before stressing the final prestress
Elimination of the classic cast-in-place closure joint
by direct junction of the two cantilever arms face to
face.
Final prestress by continuous tendons instead of
cantilever-type layout.
The total construction time for such an overpass,
including the piers, usually does not exceed two
weeks, of which less than one week is spent on the
bridge superstructure itself. This method has been
used with great success for the Rhone-Alps motorway overpasses, with spans varying between 60 ft
(18 m) and 100 ft (30 m).
11.8.2
THE BEL4iM-ASD-
WI,VCH .klETHOD
The beam-and-winch method of placing precast
segments was conceived for the construction of the
Pierre-Benite Bridges over the Rhone River. This
construction method requires a fairly simple apparatus rolling along the already constructed part
of the cantilever and ensuring the lifting, translation, and positioning of all the segments. The apparatus is shown diagrammatically in Figure 11.63.
It consists of the lifting gear B carried by the trolle)
C rolling along the bridge deck on tracks D. The
segment A is brought, bv land or water, beneath
the p ier in q u estio n, w here it is lifted bv the
equipment. It is then transported to two launching
beams E that cantilever out from the bridge deck,
upon which it continues to advance until reaching
its final position, whereupon it is lifted to its final
level next to the previous segment, Figure 11.64.
This system can, of course, be simplified if the
seg m ent c an b e b ro u g ht b y so m e ind ep end ent
means to a location vertically below its final position in the structure.
As originally conceived, this system was not
c o m p letely ind ep end ent: ano ther c o nstru c tio n
procedure was required to erect the pier segment.
The pier segment was cast in place in the PierreBenite
Bridges. It w as p recast and p laced by a
crane for the Ampel Bridge in Holland and by a
floating barge crane for the Bayonne Bridge over
the river Adour. This weakness was eliminated in
the c o nstru c tio n o f the Saint- A nd re- d e- C u b z ac
Brid g e. Fo r this stru c tu re, the p ier seg m ents,
which form the starting point for each cantilever,
were placed by the same equipment that placed the
typical span segments, Figure 3.72. The equipment
was hung, with the help of cables, to an auxiliary
mast fixed to a lateral pier face. The pier segment
was brought in from the opposite side, lifted and
placed by the mobile equipment’s winches. In the
same position the following segment was located
and the auxiliary mast removed, Figure 3.73. At
this point it was a simple matter to reposition the
mobile lifting equipment in order to place the typical span segments, Figure 3.70.
recast Segments
511
GMFNTS NFAR
PPARATUS
RIGHT RANK
TRANSFER
TROLLEY
PIER BLOCK
4ENT
PONTOON
,tream Bridge, placing apparatus.
evolved and how the original concept has been
modified.
Launching Girders Slightly Longer Than
the Span Length
We first consider the construction method of the
Oleron Viaduct Bridge superstructure, Figure
3.32. The segments were brought along the top
slab until they reached the launching girder, then
lifted by the latter, transported to their final position, lowered so as to come into contact with the
previous segment erected, and prestressed to the
cantilever. The launching girder itself, slightly
longer than the span length, was made up of a steel
trellis beam with an entirely welded rectangular
section w eighing 124 tons (113 mt) and measuring
312 ft (95 m). The maximum span length of the
bridge was 260 ft (79 m).
The launching-girder system consists of two
fixed supports, called tunnel legs, allow ing the
segments to pass between them, one at the rear of
the girder and the other at the center. At the front
end is a mobile prop enabling the girder to find
support on the next pier. The bottom chords of the
girder are used for the rolling track that supports
the segment trolley, which can move the segment
horizontally and vertically and rotate it a quar-
Technology and Construction of Segmental Bridges
512
~23~~118’~141’--ll6’-7”
4
Placlng center segment
118’i141”
259’A
Movmg gantry to next pier
Placing segments in doudle cantilever
FIGURE 11.65. Oleron Viaduct, launching-girder
operations. (A) Rear support, (B) center support, (C)
temporary front prop, (D) prop support, (E) pier segment, (F) temporary support.
ter-turn. Three phases are clearly distinguishable
in the construction of a cantilever, Figure 11.65:
Phase 1:
Placing the pier segment
A support adjustment was carried out with the
help of hydraulic jacks when the girder was resting
on the rear and central supports and the temporary front prop, before installing the pier segment.
The purpose of this adjustment was to obtain the
optimal distribution of the launching girder selfweight among the three supports. While the fi-ont
prop is being installed, the central support rests on
the end of the previous cantilever in the same position in which the rear support will be during the
erection of the typical segments. In this phase the
launching girder rests on two supports and is
therefore statically determinate; nothing can be
done to change the rear-support reaction. While
the pier segment is being placed, however, the
girder is resting on three supports and is statically
indeterminate. It is therefore necessary to ensure
that the reaction at the central support is less than
or equal to that which will be produced by the rear
support during the next construction stage, including the weight of the trolley and the tractor
placed in the near vicinity. Several other structures
have been built with launching girders of the same
generation as the one used for the Oleron Viaduct.
The Chillon Viaduct, Figures 3.43, 11.66, and
11.67, along the bank of Lake Leman used a 400 ft
(122 m) launching girder w eighing 253 tons (230
mt). The maximum span length was 34 1 ft ( 104 m).
The launching girder, of constant rectangular section, was of the suspension type, being suspended
at the one-quarter points by cable stays anchored at
the central mast, which extended above the level of
the launching girder. The supports were hydraulically adjustable, allowing the girder to cope
with different angles of superelevation, Figure
The launching girder rests on three supports-the
rear support, the center support near the end of
the newly constructed cantilever, and the front
prop, which is attached to the front of the next pier
with the help of a temporary prop support.
Phase 2:
Moving the launching girder forward
The girder rolls along on the rear support and the
segment trolley, which is rigidly attached to a metal
framework known as the temporary translation
support, which is fixed to the pier segment. The
rear and central supports are equipped with bogies
and roll along a track fixed to the bridge deck while
the girder is being moved forward.
Phase 3:
Placing
typical
segments
The launching girder rests on two supports, the
central support anchored to the pier segment and
the rear support tied with prestressing bars to the
end of the previously constructed cantilever.
FIGURE 11.66. Chillon Viaduct, launching-girder in
operation.
Placing Precast Segments
513
4th &age
lauchmg
Girder
FIGURE 11.67. (:hillon
Viaduct, launching-girder mo vements.
11.68. The launching girder included three means
of adjustment:
Lateral movement of the trolley in
order to place eccentric segments
Adjustment 02: Lateral translation of the central
support in order to cope with
ho riz o ntal c u rv atu re o f the
structure
A@.stment 03: Vertical adjustment of bogies to
take up the superelevation and so
keep the central support vertical.
Adjustment
D3
FIGURE 11.68. Chillon
adjustments.
Viaduct, launching-girder
Dl:
In order to follow the horizontal curves the
launching girder rotated about the rear support
while moving sideways across the central support,
Figure 11.69. The mobile temporary front prop
was conceived in the same way as the other supports so as to allow the passage of the first segments
to either side of the pier segment.
The Blois Bridge on the Loire River in France
had a 367 ft (112 m) long launching girder w eighing 135 to ns ( 123 m t) , Fig u re 11.70. The
maximum span length was 300 ft (91 m). The
launching girder, of constant triangular section,
could be dismantled and transported by road. All
of the girder components were assembled with
high-strength bolts, ensuring the transmission of
514
Technology and Construction of Segmental Bridges
CONSTRUCTION OF
H OlzI Z 0 rifA~
CONSTRUCTION OF
UORIZONTAL
CURVC
CURVE
( STAGE 1. )
( STAGE 2 )
FIGURE 11.69. Chillon Viaduct, curved span construction.
ELEVATION
FIGURE 11.70. Blois Bridge, launching girder.
SFCTION A
forces by friction between adjoining plates, Figure
11.71.
The use o f a v ery lig ht structural steel
framework carried with it the risk of large deflections. These were reduced and controlled by two
sets of cable stays, passive and prestressed, which
came successively into play during maneuvering of
a segment (upper passive stays) and during the
launching-girder advancement (lower prestressed
stays). This launching girder was later used for the
erection of two other structures: the Aramon
Bridge on the Rhone River, Figure 11.72, and the
2950 ft (900 m) long Seudre Viaduct.
The Saint Cloud Bridge on the Seine, Figure
3.78, is a recent example of the use of a large
launching girder. The girder could place segments
weighing up to 143 tons (130 mt) in spans of up to
335 ft (102 m) with a minimal radius of curvature
in plan of 1080 ft (330 m), Figure 3.79. The weight
of the launching girder was 260 tons (235 mt) and
its total length was equal to 400 ft (122 m).
The adjustments adopted were similar to those
used for the Oleron, Blois, and Chillon bridges.
The launching girder, which used upper passive
stays and lower prestressed stays, was constructed
FIGURE 11.71. Blois Bridge, launching-girder assembly detail.
FIGURE 11.72. Aramon Bridge over the Rhone
River.
with a constant triangular section made up of individual elements assembled by prestressing. This
launching girder is notable, apart from its assembly
by prestress, for its ability to follow extremely tight
curves. The movements used for the Chillon Via-
Placing Precast Segments
duct were, of course, used for this purpose. However, in the Saint Cloud Bridge it was necessary
also for the launching girder to take up several intermediate positions during the erection of a given
cantilever so as to bring each segment to its final
position in the structure. The total lateral translation reached 19.7 ft (6 m) at its maximum. Construction speed of the bridge deck was 130 ft (40
m) per week, including all launching-girder maneuvers. Two other structures erected with the
help of the Saint Cloud launching girder were the
Angers Bridge and the Sallingsund Viaduct.
The launching girder used for the Alpine
Motorway network was conceived for spans and
segment weights of more modest dimensions; it is
typical of lightweight universal equipment that can
be easily dismantled for reuse in another structure,
Figure 11.73. This girder allowed the handling of
segments weighing up to 55 tons (50 mt) over
spans up to 200 ft (60 m).
Reflecting on the launching girders mentioned
above, we note that their evolution centers on two
major characteristics: the structural conception of
the girder and the assembly method (connection
types, number of elements, and so on).
Launching girders tend more and more to be of
the lightweight type, relying on exterior forces to
cope with different loadings. These exterior forces
are provided by the external active cable stays,
which allow the structure to be placed in a condition ensuring a favorable behavior under a given
loading. This approach to launching-girder design
provides more optimal use of materials than did
the first-generation girders of variable cross section.
Another advantage of a constant cross section is
that it facilitates the construction of standard sec-
FIGURE 11.73. Alpine Motorway launching girder.
515
tions that can be interchanged and assembled on
site. In this way the girder length can be varied according to the span length and the weight of the
segments. Connections are made with tensioned
bolts, Figure 11.74, which reduce considerably the
number required and consequently the time
needed to assemble or dismantle the structure.
These connections have recently replaced those
made with high-strength bolts and fishplates, notable on such structures as the Deventer Bridge and
the B-3 Viaducts.
Means of erection adjustments also have improved, tending to reduce the forces applied to the
deck itself by ensuring that the girder supports are
located over the piers or at least in the very near
vicinity.
This natural evolution leads us toward a new
type of launching girder, one whose total length is
slightly greater than twice the typical span length,
allowing the simultaneous placing of the typical
segments of cantilever N and the pier segment of
cantilever N + 1.
Launching Girders Slightly Longer Than Twice the
Typical Span
The first launching girders of this type were used
on the following bridges: Rio Niteroi in Brazil; Deventer in Holland, Figure 3.50; and B-3 South
Viaducts in the eastern suburbs of Paris, Figure
3.93.
The Rio Niteroi Bridge (Section 3.8), linking the
city of Rio de Janeiro with Niteroi, consists of 10
miles (16 km) of bridge deck constructed by four
identical launching girders, Figures 3.55 and 3.56.
Each 545 ft (166 m) long girder could be completely dismantled. The constant triangular sec-
zyx
FIGURE 11.74. Prestressed connections.
516
Technology and Construction of Segmental Bridges
tion, weighing 440 tons (400 mt), could cope with
spans of up to 260 ft (80 m). The connections were
identical in principle to those used for the Blois
girder. Each installation was equipped with three
supports of nontunnel type, one fixed and the
other two retractable.
The erection sequence was as follows, Figure
1.51:
Phase 1: Segment placing
The girder rests on three supports, each one over a
pier. Two segments are erected simultaneously,
one on either side of the double cantilever under
construction. The pier segment of the next cantilever is also placed with the launching girder in
this position.
Phase 2:
M oving the launching gder forw ard
The girder rolls on two temporary translation supports, one placed above the pier of the finished
cantilever and the other above the pier of the cantilever to be constructed. These temporary supports are attached to the trolleys; the launching
girder is lifted, thus freeing the permanent supports; and the trolleys are engaged, enabling the
translation of the launching girder to a position to
erect the next cantilever. The temporary translation supports are equipped with a mechanism allowing transverse movements, as the structure includes a certain amount of horizontal curvature.
The Rio Niteroi girder was equipped with three
sets of active stays: lateral stays, central stavs, and
launching stays. The lateral stays, positioned on
the underside of the two spans and constantly
under tension, ensure the resistance of the girder
while the load (segment) passes near midspan. The
central stays strengthen the girder in the vicinity of
the central support. The launching stays, under
tension while maneuvering the girder, transfer the
front and rear reactions to the central support.
Owing to the length of the bridge and the presence of a large stretch of water beneath the structure, the segments were brought to the launching
girder on barges. The cantilever stability of the
bridge was assured by the launching girder itself,
and ties and props were positioned as construction
proceeded.
The launching g ird er u s e d f o r the D ev enter
Bridge in Holland, Figures 3.49 and 3.50, were
also capable of being entirelv dismantled and of
triangular section. Its total length was 5 12 ft (156
m) f o r a w e i g h t o f 1 9 8 t o n s ( 1 8 0 m t ) . T h e
maximum span length was 243 ft (74 m).
Assembly of the launching-girder elements was
c o nsu m m ated b y p restress b ars no rm al to the
joints. It was supported by the fixed supports, of.
which the rear and the central allowed the passage
of a segment, and two sets of cable stavs: central
stays and launching stays. The translation operations were identical to those of the Rio Niteroi
Bridge, even though only one segment could be
lowered into place at a time.
What was peculiar about this launching girder
was its abilitv to raise itself to its working level bv its
own means, and this from the ground level where
it was assembled. This was made possible bv the
central suspension mast, which acted as a lifting
ja c k.
In the case of the B-3 South Viaducts, Figure
3.92, the constantly varying structure supported b\
200 piers, crossing five railway tracks, the Ourcq
Canal, and several urban roadwavs, was mastered
b y a hig hly m ec haniz ed launc hing g ird er. The
simultaneous placing of two segments of the same
cantilever, each weighing between 33 and 55 tons
(30 and 50 mt) either side of the pier, is controlled
by a radio-controlled servo mechanism that synchronizes the loading at each end of the girder.
A g ain the leng th o f the lau nc hing g ird er w as
slightly greater than twice the typical span length,
TYPICAL
CROSSw8ECTION
FIGURE 11.75. B-3 South Viaduct launching girder. general la\o11t.
517
References
that is, between four and six segments per day.
The average construction speed, including
launching-girder maneuvers, was therefore 200 ft
(60 m) per week.
The B-3 launching girder was recently reused
for the Marne-la-Vallee Viaduct, which carries
high-speed suburban rail for the Paris transport
authority.
References
FIGURE 11.76. B-3 Sourh Viaduct, segment transport tractor.
which varied between 100 and 164 ft (30 and 50
m), Figure 11.75. The girder support reactions
were thus applied in the region of the piers, and
the cantilever stability was ensured by the launching girder itself. This stabilizing device can be seen
to the left of the central support in Figure 11.75.
The segments were supplied by a special eightwheeled tractor moving along the top slab, Figure
11.76. A special device used to unload and store
the segments brought by the tractor freed the latter and removed the supply of segments from the
erection critical path. The cycle of segment placement and girder advancement is represented in
Figure 3.93. The next pier segment was placed
during the same phase as the typical segments.
About two spans were constructed each week-
1. Anon., M anual for Q uality Co ntro l for Plants and Pro duction of Precast Prestressed Concrete Products, MNL116-70, Prestressed Concrete Institute, Chicago,
1970.
2 . Anon., A CI Manual o f Co ncrete Practice, Part I, American Concrete Institute, Detroit, 1973.
3. “Proposed Recommendations for Segmental Construction in Prestressed Concrete,” FIP Commission-prefabrication, 3d Draft, September 1977.
4. “Recommended Practice for Segmental Construction
in Prestressed Concrete,” Report by Committee on
Segmental Construction, Journal of the Prestressed
Concrete Institute, Vol. 20, No. 2, March-April 1975.
5. Anon., PCI Po st- Tensio ning M anual, Prestressed Concrete Institute, Chicago, 1972.
6. Anon., PTI Po st- Tensio ning M anual, Post-Tensioning
Institute, Phoenix, Arizona, 1976.
7. T. J. Bezouska, Field Inspection of Grouted PostTensioning Tendow,
Post-Tensioning Institute,
Phoenix, Arizona, March 1977.
12
Economics and Contractual Aspects
Of Segmental Construction
12.1 BIDDING PROCEDURES
12.2
12.1.1 Single Design
12.1.2 Design and Build
12.1.3 Value Engineering
12.1.4 Alternate Designs
12.1.5 Summary Remarks on Bidding Procedures
EXAMPLES OF SOME INTERESTINti BIDDINGS AND
cDsrs
1 2 . 2 . 1 Pine Valley Creek Bridge, California
12.2.2 Vail Pass Bridges, Colorado
12.2.3 Long Key Bridge, Florida
12.2.4 Seven Mile Bridge, Florida
12.1 Bidding Procedures
A bridge design should on principle be economical
and as a practical matter must fall within budgetary
restrictions of a particular project. The economic
“ moment of truth” for a given bridge design occurs
when bids are received and evaluated.
In a basically stable economy where material and
labor costs are predictable within relatively small
fluctuations, the selection of structure type and
materials is relatively straightforward. This situation prevails when the time required for the design
is relatively short and thus is not affected by economic cycles, or, if the design time is relatively
long, the economic cycles are mild. In an inflationary economy there is no economic stability, and designers are hard put to make rational choices, as
they have no control over economic parameters
that can influence their design decisions. In short,
the p ro b lem is w hether ec o no m ic assu m p tio ns
made during the course of design are valid at the
time of bidding.
518
12.2.5 Zilwaukee Bridge, Michigan
12.2.6 Cline Avenue Bridge, Indiana
12.2.7 Napa River Bridge, California
1 2 . 2 . 8 Red River Bridge, Arkansirs
12.2.9 North Main Street Viaduct, Ohio
12.2.10 Summary of California’s Experience
12.3
INCREASE IN EFFICIENCY IN CONCRETX BRIDGES
12.3.1
12.3.2
Redesign of Chacas Viaducts, Venezuela
Comparison between Tancarville and Brotonne Bridges,
France
REFERENCES
Obviously, the design and the bidding (tendering) of a project are closely related. Contractual
bidding procedures vary from country to country,
and c u rrent ec o no m ic p ressu res are lead ing to
changes in these procedures. The various bidding
methods used in various countries can be broadly
categorized (with some possible variations) as follows: (1) single design, (2) design and build, (3)
value engineering, and (4) alternate designs.
12.1.1
SISGLE
DESIGl\
Heretofore, single design was the major method
used in North America and Great Britain. In this
method, in general, design drawings prepared for
bid are very detailed, to the extent that even the
length and other dimensions of every reinforcing
bar may be given. The bidding period is followed
by a tight construction schedule. The contractor
bids and executes the project in strict accordance
with the bidding documents. No variation from the
documents is allowed unless an error in design is
519
Bidding Procedures
discovered, or a specific detail proves impractical to
consummate, or geological perturbations are discovered that differ from what was assumed in design and delineated in the contract documents.
These changes are authorized by a change order,
and if there is an increase in cost the contractor is
paid an “ extra.”
This system worked well for many vears when
the economy was fairly stable and predictable and
when economic changes were gradual over an extended period. Its disadvantage is its lack of flexibility to accommodate an inflationary economy,
sudden price changes in materials, a rapidly advancing technology, and the current emergence of
specialtv c o ntrac to rs with u niq u e eq u ip m ent o r
skills, proprietarv designs, and patented construction methods. Its biggest advantages are ease in
administering the contract and absolute control
over the final design.
In so m e Eu ro p ean c o u ntries, b y c o ntrast, b id
documents are prepared with the intention that
the contractor will prepare and submit his own
detailed design for the prqject. Thus, bid plans will
be more general and, for a bridge, may show only
sp an leng ths, p ro file, and ty p ical sectio ns. The
contractor may then refine the original design or
submit an alternate design of his own choice, the
responsibilitv for producing the final design and
details being his rather than the engineer’s, This
procedure allows the contractor to use any special
equipment or technique he may have at his disposal. For example, a cast-in-place concrete box
mav be substituted for a steel superstructure where
the contractor has special know-how in concrete
construction, or the change may be less drastic and
involve only a reduction in the number of w ebs in a
box girder.
Verification of the adequacy of the contractor’s
final design is generally carried out by a “ proof engineer” who is retained by the owner or is on the
owner’s engineering staff. In order to minimize
d isag reem ents b etw een the c o ntrac to r and the
proof engineer, European codes have been made
very
specific. As a result, European contractors
usually maintain large in-house engineering staffs,
although they may also use outside consultants.
The outcome apparently is a savings in construction cost, achieved by the investment of more desig n tim e and effo rt than in the sing le-d esig n
method.
The advantage of the design-and-build method
is that in an atmosphere of engineering competition, innovative designs and construction practices
advance very rapidly. The state of the art of designing and constructing bridges advances in response to the need for greater productivity. The
disadvantage is the lack of control over the selection of the type of structure and its design. There is
some concern, too, that quality of construction may
suffer as a consequence of overemphasis on productivity and initial cost. However, the contractor
is usually required to produce a bond and guarantee his work over some period of time, and any defects that surface during this period have to be repaired at his expense. Whether such a system could
be adopted in the United States is debatable.
1 2 . 1 3 VALUE
E.\‘GI.~EERI,~G
Value engineering is defined by the Society of
American Value Engineering as “ the systematic
application of recognized techniques which identify the function of a product or service, establish a
value for that function, and provide the necessary
function reliability at the lowest overall cost. In all
instances the required function should be achieved
at the lowest possible life-cycle cost consistent with
requirements
for
performance,
maintainability,
safety, and esthetics.“ ’
In 1962 the concept of value engineering became mandatory in all U.S. Department of Defense
armed services procurement regulations (ASPR).
Bef o re this tim e v alu e eng ineering had b een
applied to materials, equipment, and systems. The
advent of ASPR provisions introduced value engineering concepts to two of the largest construction agencies in the United States-the U.S. Army
Corps of Engineers and the U.S. Navy Bureau of
Yards and Docks. Soon thereafter the U.S. Bureau
of Reclamation and the General Services Administratio n (GSA ) ad o p ted and inserted v alue engineering clauses in their construction contracts,
and the U.S. Department of Transportation established a value engineering incentive clause to be
used by its agencies.
Several value engineering clauses (or’ incentive
clauses) are in use today by many agencies. In general, they all have the following features’:
1.
A paragraph that defines the requirements of
a proposal: (a) it must require a change to the
contract and (b) it must reduce the cost of the
contract without impairing essential functions.
Economics and Contractual Aspects of Segmental Construction
520
2. A “ d o c u m entatio n” paragraph that itemizes
the information the contractor should furnish
with each proposal. It should be comprehensiv e eno u g h to ensu re q u ic k and ac c u rate
evaluations, d etailed eno u g h to ref lec t the
contractor’s confidence in its practicability, and
refined to the point where implementation will
not cause undue delay in construction operations. Careful development of this paragraph
and meticulous adherence to its requirements
w ill p rec lu d e sc atter-sho t p ro p o sals b y the
c o ntrac to r and b u rd enso m e rev iew b y the
agency.
3.
A paragraph on “ submission.” This paragraph
details the procedure for submittal.
4. A paragraph on “ acceptance,” which outlines
the right of the agency to accept or reject all
proposals, the notification a contractor may
expect to receive, and appropriate reference to
proprietary rights of accepted proposals.
5.
A paragraph on “ sharing,” which contains the
f o rm u la f o r d eterm ining the c o ntrac t p ric e
adjustment if the proposal is accepted and sets
forth the percentage of savings a contractor
may expect to receive.
As generally practiced by highway agencies in
the United States, a value engineering proposal
must indicate a “ substantial” cost savings. This is to
preclude minor changes such that the cost of processing offsets the savings to be gained. Some other
reasons for which a value engineering proposal
may be denied are as follows:
Technical
noncompliance.
Delay in construction such that the cost savings
would be substantially nullified.
Proposed change would require resubmission of
the project for any number of various permits,
such as environmental impact statement, wetlands
permit, and navigation requirements. Resubmission would in all probability delay construction and
nullify any cost savings.
Savings resulting from a value engineering proposal are generally shared equally by the agency
and the contractor, after an allowance for the contractor’s development cost, the agency’s cost in
processing the proposal, or both. As practiced in
the United States, all contractors must bid on the
design contained in the bid documents, and only
the low bidder on the base bid is allowed to submit
a value engineering proposal. This is, of course,
v alu e eng ineering ’ s b ig g est d isad v antag e. An)
nu m b er o f c o ntrac to rs m ay hav e m o re costeffective proposals that they are not allowed to
submit because they were not low bidder on the
base bid. Its advantage is that to some degree it allows contractor innovation to be introduced.
Alternate designs, as it is developing in the United
States, basicallv is an attempt to produce a hvbrid
sy stem consisiing o f the b est elem ents o i the
single-design and the design-and-build methods. It
attempts to accomplish the following:
Retain for the authorizing agency control over the
“ type selection” of the structure and its design
Provide increased competition between materials
(structural steel versus concrete or prestressing
strand versus bars) or construction procedures
(cast-in-place versus precast segmental or balanced
cantilever versus incremental launching, and so
04
Provide contractor flexibility (construction
dures, methods, and/ or expertise)
proce-
This method has developed, with encouragement
from the Federal Highway t\dIninistratiorl, as an
anti-inflationary measure to combat dramatic increases in highway construction costs. A technical
2
Advisory published by the Federal Highway Administration states:
Because qf.fluctuating
economic conditions, it isfelt that
on multiple repetitive spans, long spans or major bridges,
or where there is an extended period qf‘design from conception of the project to a release for bds, there can be no
assurance of price stability fbr n particular material or
construction methodoloCg. With alternate de.siLgns, no
matter how the economy changes, more designs are ctzlailable at the time of biddt’ng that are likely to be suited to the
prevailing
economic
conditions.
General recommendations regarding alternate
designs from the same document’ are as follows:
1.
To receive the most economical construction
between basic structural materials, consistent
with geographic, environmental,
ecological
o r o ther site restric tio ns, there sho u ld b e
maximum opportunity for competition between structural steel and concrete.
521
Bidding Procedures
2. W ithin env iro nm ental, aesthetic , site, and
other constraints, the plans and bid documents should show or otherwise indicate what
alternative types of structures will be allowed
or considered. The contractor should be allowed the option to bid any designated alternative design that is consistent with the contractor’s expertise, available equipment, and
so on.
3.
Bid documents and the contract plans should
clearly indicate the design criteria and what
tvpe of alternative designs and/ or contractor
options will be acceptable. Determination of
practical and economical alternatives and/ or
contractor options should be developed in the
preliminary design.
4. Bid d o c um ents sho uld b e c o nsid ered as
“ open” documents in regard to construction
method, erection systems, and prestressing
svstems.
5.
Consistent design criteria should be used for
alternatives; for example, if load factor design is used, it should be used for all alternatives.
6. Sp an leng ths sho uld b e id entified o n the
contract plans. However, other than where
pier locations are constrained by physical and
geological conditions at the site, consideration
should be given to allowing a tolerance in pier
location to avoid placing a particular alternative at an economic disadvantage. For example, in a typical three-span structure, the side
span should be approximately 80 percent of
the center span for structural steel, 70 percent for conventional cast-in-place concrete
on falsework, and 65 to 70 percent in segmental balanced cantilever construction.
To avoid an economic disadvantage to a particular superstructure alternative, alternative
substructure designs may be required. Limitations on the substructure, such as allowable
axial lo ad and m o m ent, sho uld b e clearly
identified on the contract plans.
8. Where specific design requirements are not
covered by the American Association of State
Highway
and
Transportation
Officials
(A A SH TO ) Brid g e Sp ecificatio ns, the co ntractor should be allowed to use other recognized codes and standards where applicable.
How ever, the alternativ e d esig n sho u ld
document where these provisions are to be
used, why the AASHTO requirements do not
apply, and which articles of the substituted
code or standard are to be used. Such provisions should be subject to approval by the engineer and appropriate agencies.
9.
Prebid conferences are to be encouraged as a
means of communication between the engineer, highway agencies, and contractors.
10.
In order to allow a contractor adequate time
to investigate the various alternatives and
prepare plans, it is recommended that the
advertising time be commensurate with the
size and complexity of the project with a
minimum of 60 days.
11.
In o rd er to allo w ad eq u ate rev iew and
checking of the low bidder’s proposal, award
of contract should be extended commensurate with the size of project.
Specific recommendations* regarding
concrete alternates are as follows:
1.
prestressed
To increase the competition in post-tensioned
concrete construction, it is recommended that
plans and other bid documents allow conventional cast-in-place on falsework, precast prestressed span units, and segmental construction or combinations thereof.
2. Segmental construction should allow the following at the contractor’s option:
a. Prec ast o r c ast-in-p lac e seg m ental c o nstruction.
b.
Any of the post-tensioning systems-that
is, strand, wire, or bars or combinations
thereof.
c.
following
construction
Any
of
the
m etho d s: b alanc ed c antilev er, span-bysp an, p ro g ressiv e p lac ing , inc rem ental
launching, or combinations thereof.
d.
Exterior dimensions of the cross section
should be fixed. At the contractor’s option,
the thickness of webs and flanges may be
v aried to ac c o m m o d ate p ro p o sed c o nstru c tio n and erec tio n m etho d s and
post-tensioning systems, providing that
any changes in the dead weight, shear, and
so on are accommodated in the design.
7.
3 . T h e c o n t r a c t p lans sho uld ind icate the
maximum and minimun final prestressing force
(P,) and moment (Pr x e) required, after all
losses, for the final condition of the structure
-that is, dead, live, impact, and all superimposed loads. Any increase in prestressing force
Economics and Contractual Aspects of Segmental Construction
522
requirements as a result of the method of construction, erection, or type of tendon system
should be evaluated at the shop drawing stage.
4.
Changes in
accompanied
stress force
compressive
eccentricity of prestress should be
with appropriate changes in preto produce the same minimum
stress due to prestress.
5.
The minimum prestress force should be such
that under any loading condition, both during
and after construction, stresses will be within
allowable limits. Consideration should be given
to secondary moments due to prestress, redistributed moments due to creep, and stresses
resulting from thermal gradient (between the
top and bottom of the girder and between the
inside and outside of webs).
6. C o ntrac to r rev isio ns to c o ntrac t p lans, w ith
supporting calculations, should be submitted
to the engineer for approval.
REtMARKS
12.1.5 SU,MMARY
P R O C ED U R ES
OlV BIDDIiVG
All of the bidding procedures described above
have one thing in common: they all attempt to
produce the lowest initial cost by competition in
construction and/ or design. All of the last three
approaches (design-and-build, value engineering,
and alternate designs) require decisions based on
comparisons of basic structural materials, structure
types, construction methods, and so on. This implies that the basic premise in the selection process
is equivalency-comparable service, performance,
and life-cycle cost of the facility.
Life-cycle costs refer not only to initial cost, but
also to maintenance and any rehabilitation costs
during the life of the structure. True cost of the
project must be considered. What may be initially
least expensive may in the long run, when future
costs are accounted for, be actually most expensive.
Some newer structure types and designs are at the
fringe of the state of the art and have only been
used in the United States within the last decade or
less. Thus, an adequate background of experience
is unavailable to evaluate life-cycle costs. The estimation of life-cycle costs may be difficult in many
cases, such as for new and progressive bridge designs. Functionally, alternative structures are designed to the same criteria. Only years of operational experience can provide the data base for
reasonably estimating life-cycle costs and thereby
true equivalency in design insofar as cost is in-
volved. However, the problem of adequacy of data
does not diminish the importance of the question
and the need to attempt to answer it.
Another anti-inflationary measure used in recent
years is that of stage construction. This concept
may take one of two forms. Major structures, bec au se o f their siz e, lend them selv es to stag e
c o nstru c tio n-that is, sep arate su b stru c tu re and
superstructure contracts. Usually several years will
elapse between bidding and awarding of the substructure contract and the superstructure contract.
The economic superstructure span range for different alternative types and materials is a variable.
In this form of stage construction the substructure
is let first; thus the spans for the superstructure design become fixed. This may or may not impose an
economic disadvantage to specific superstructure
alternates. The substructure must be designed for
the largest self-weight superstructure alternative,
which may or may not be the successful superstructure
alternative. It ap p ears that this f o rm
of stage construction may be to some extent selfcanceling or counterproductive to cost savings.
With a total alternative design package, the substructure (foundation, piers, span arrangement)
can also have alternatives commensurate with the
superstructure
alternatives.
The other form of stage construction concerns a
large project, containing many bridges, that is subdivided for bidding purposes into a number of
smaller projects. Its primary purpose is to encourag e sm all c o ntrac to rs bv p ro v id ing prqjects o f
m anag eab le siz e, thu s Inc reasing c o m p etitio n.
However, certain construction techniques, by virtue of the investment in sophisticated casting or
erection equipment, require a certain volume of
work to amortize the equipment and be competitive. Depending upon the size of the subdivided
contract, this form of stage construction in some
instances may also become counterproductive.
The value engineering concept can be divided
into two major areas of application: during design
and during construction. Value engineering procedures in the design stage may result in very
specific recommendations based on a certain set of
assumptions at a particular point in time for the
design. If conditions change during the interval
between the design decision and the actual construction, which can be several years, conditions on
w hic h the assu m p tio ns w ere b ased m ay hav e
changed. Such changes could make the original
value engineering decision incorrect. The alternative design concept, on the other hand, does not
make all such specific design decisions at an early
523
Examples of Some Interesting Biddings and Costs
stage but retains some options in order t