Dear Readers,
Scientific Journal of Civil Engineering was established in December 2012. This effort, led by Faculty of Civil Engineering (FCE) – Skopje creates opportunities for staff and postgraduate students at FCE to publish the results of their research activities for the general scientific and professional community.
SJCE is intended to publish the contributions in the fields of; structural engineering, transportation engineering, geotechnics, hydro technical engine-ering, construction management and geodesy. SJCE is published bi-annually and is available online at FCE web site
( www.gf.ukim.edu.mk
).
It is my pleasure to introduce the Second Issue of VOLUME 3 of the Scientific Journal of Civil
Engineering (SJCE).
At the end of the third year of publication of our journal we are now able to think with justification about the realization of the ideas which led us to lunch the journal, the positive results achieved together with authors and readers, but also about new challenges for the future.
The second issue of volume 3 presents a special issue with selected articles from the last Symposium of Macedonian Association of Geotechniks (MAG), organized in city of Struga this year.
This issue includes 5 articles and some of them are invited papers. Articles are in the field of: probabilistic concept in analysing of soil bearing capacity, layer influence on dynamic behaviour of soil, laboratory experiments on soil dynamic characteristics, laboratory experiments for soil improvement and impact of ground vibrations induced by pile driving.
It’s our honour to publish several articles which are specially selected from the number of quality articles published in the proceedings from the latest symposium of Macedonian Association of
Geotechnics. Special thanks to my colleague assist.
Prof. Jovan Papic, who help me in the selection of the articles, and also special gratitude to all authors whose papers are published in this issue.
We continue to invite all researchers, practitioners and members of the academic community to contribute through their articles to the development and maintenance of the quality of the SJCE journal.
We are particularly pleased to publish the results of research, best practice, case studies, ideas for solutions of complex problems, proposals of innovations and the results of experience on important projects.
Sincerely Yours,
Prof. Ph.D. Darko Moslavac
December, 2014

Faculty of Civil Engineering -Skopje
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Email: Darko Moslavac
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Faculty of Civil Engineering -Skopje
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1000 Skopje
Rep. of MACEDONIA
Prof. Ph.D. Darko Moslavac
University Ss. Cyril and Methodius
Partizanski odredi 24, 1000 Skopje
Rep. of MACEDONIA tel. +389 71
368 372; Email: moslavac@gf.ukim.edu.mk
ISSN: 1857-839X
Prof. Ph.D. Darko Moslavac
University Ss. Cyril and Methodius,
Rep. of Macedonia
Prof. dr. sc . İbrahim Gurer
Gazi University, Turkey
Prof. dr
University of Belgrade, Rep. of
Serbia
Vienna University of Technology,
Austria
Miodrag Jovanovic
Rep. of Serbia
Dr.h.c.mult. Dr.techn.
Dr.h.c.mult. Dr.techn.
Prof.dr.ir.
Heinz Brandl
Vienna University of Technology,
Prof. dr. sc.
J.C. Walraven
Heinz Brandl
Zalika Črepinšek
University of Ljubljana, Slovenia
Delft University of Technology,
Netherland
Delft University of Technology,
Netherland
Viktor Markelj
PhD, Assoc. Prof. Jakob Likar
University of Maribor,
University of Ljubljana, Slovenia
Slovenia
ITA Croatia
Davorin KOLIC
Prof. dr. sc. Stje pan Lakušić
University of Zagreb, Croatia
University of Ljubljana, Slovenia
Marc Morell
ITA Croatia Institut des Sciences de l’Ingénieur de Montpellier, France
Prof. Ph.D. Miloš Knežević
University of Montenegro
Prof. Ph.D. Milorad Jovanovski
University Ss. Cyril and Methodius,
Prof. Ph.D. Cvetanka Popovska
University of Montenegro University Ss. Cyril and Methodius,
Rep. of Macedonia
Prof. Ph.D. Ljupco Lazarov
University Ss. Cyril and Methodius,
Rep. of Macedonia
Prof. Ph.D. Goran Markovski
Rep. of Macedonia
University Ss. Cyril and Methodius,
Rep. of Macedonia
University Ss. Cyril and Methodius,
Prof. Ph.D. Ljupco Lazarov
University Ss. Cyril and Methodius,
Rep. of Macedonia
Prof. Ph.D. Goran Markovski
University Ss. Cyril and Methodius,
Prof. Ph.D. Elena Dumova
Jovanovska
University Ss. Cyril and Methodius,
Prof. Ph.D. Elena Dumova
Jovanovska
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PRINT
Evropa92
Mile R. Mihajlov
STRESS ANALYSIS OF BOOMS – PANEL COMBINATION STRUCTURE
György L. BALÁZS
POTENTIALS IN USE OF EBR AND NSM STRENGTHENING METHODS FOR CONCRETE STRUCTURES
RIB-STIFFENED AND COVER-PLATE FULL STRENGTH CONNECTIONS IN
STEEL FRAMES FOR SEISMIC APPLICATIONS
Paulo Lourenco, Graca Vasconcelos
MODERN MASONRY STRUCTURES AND EARTHQUAKES: REINFORCED OR UNREINFORCED?
Cvetanka Popovska, Mattew Jones, Pavlina Zdraveva, Violeta Gesovska
WATER RESOURCES AND THE CHALLENGE OF CLIMATE CHANGE
31
49
69
7
19
5 Page

7
N. Davidovic, Z. Bonic, V. Prolovic
15
Josif Josifovski
25
V. Shesof, V. Gadza, I. Zafirova, J. Bojadzieva, K. Edip
35
B. Susinov, J. Josifovski, S. Abazi
43
Zvonko Tomanovic
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 5
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624.073 624.042: 519.97
A UTHORS
Mile R. MIHAJLOV
Dr. Ing. Former Assistant
Ss. Cyril and Methodius University, Macedonia
Faculty of Civil Engineering Skopje milemihajlov@yahoo.com
Diffusion problems take place quite often in engineering structures. They are common to concrete, wooden and metal construction as well. The reason for diffusion to arise are cutouts and any discontinuity in the structure and external loading. As a consequence to diffusion of stresses, direct or shear, concentration of stresses takes place. To avoid miss – calculation of concentration of stresses , care must be taken of theoretical modeling the stress analysis. In this paper is presented mathematical modeling of two booms-panel structure subjected to concentrated loads. Finite element stress field assumption was used to analyze the diffusion problem. Principal of virtual stresses is employed. The results are tabulated. From the numerical results, it is possible to conclude that concentration of shear stress along the boom takes place.
shear stress, direct stress, boom, panel, diffusion, eigen value, eigen vector
Diffusion problems in engineering structures are very common today. They take place in structures with cut-outs where discontinuity exists. Usually they are dominant in thin sheets which can not take concentrated loads.
In such cases it is common to introduce strong booms rigidly attached to the sheet to diffuse the load into the sheet.
In this paper will be analysed thin panel reinforced by two longitudinal booms that are subjected to concentrated loads at one end, and are fixed to support at the other end, together with the panel. Up to now, most of the theoretical work treats this problem as one dimensional, such that the booms take the axial load, and the panel takes only shear load. In practice it is possible to prove that the panel diffuses the axial load by shear stress and thus inducing tensile stresses as well.
Scientific Journal of Civil Engimeering, Volume 3, Issue 1, June 2014 7 Page

Probabilistic Concept In The Computational Analysis Of The Soil Bearing Capacity
Authors
Nebojsa DAVIDOVIC
Ph.D. Full Professor
Faculty of Civil Engineering and Architecture, Nis dnebojsa@gaf.ni.ac.rs
Zoran BONIC
Ph.D. Full Professor
Faculty of Civil Engineering and Architecture, Nis zokibon@gaf.ni.ac.rs
V. PROLOVIC
Ph.D. Full Professor
Faculty of Civil Engineering and Architecture, Nis vprolovic@gaf.ni.ac.rs
In computational analysis of the soil bearing capacity the definition of an appropriate model due to the complexity of the material (soil) and uncertainties inherent in the influencing factors is difficult. In traditional deterministic methods representative values (usually an average or the lowest value obtained from field and/or laboratory test results) of soil properties are selected by the engineer. The limitations of deterministic approach are that it does not consider the uncertainty of input parameters. In this paper, an alternative probabilistic concept, which uses the Monte Carlo Simulation to account for input parameter uncertainty, is presented.
This approach enables bearing capacity to be quantified in the form of a cumulative distribution function (CDF) from which the values of soil bearing capacity associated with target reliability levels of 90% and 95% (usually needed for design) can be estimated.
Probabilistic concept, deterministic approach, bearing capacity, soil parameters, uncertainty, variability.
For the design of various structures such as buildings, bridges, dams, roads, tunnels, etc., accurate and detailed subsurface information at the site are required. In general, the design of foundations must satisfy two primary requirements: (1) An adequate safety against shear failure; and (2) the total and differential settlements of the foundation must be within limits that can be tolerated by the superstructure
[1]. These two aspects of the design are normally treated separately, in the sense that they are treated as strength and deformation problems respectively, however, bearing capacity usually governs the design process. If bearing capacity is over-estimated, soil will fail, leading to serious consequences and fatalities.
If, on the other hand, bearing capacity is underestimated, undue costs are usually incurred.
Consequently, an accurate prediction of bearing capacity is important for a safe and reliable design of shallow foundations.
The geotechnical engineer’s task is to explore the subsurface conditions at a project site and determine the capacity of the soil to carry
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 7
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N. Davidovic, Z. Bonic, V. Prolovic the load without collapsing or experiencing intolerable deformations. Based on experience and supported by theory, the geotechnical engineer interprets the information in order to predict foundation performance.
The ultimate bearing capacity (qu) of the underlying soil is a key factor in the design of shallow foundations. A number of methodologies for determining the ultimate bearing capacity currently exists in the literature. Based on the ultimate bearing capacity, the allowable bearing capacity (qa) to be used in designing the foundation is calculated through the use of an appropriate factor of safety (deterministic approach) or by adopting a probabilistic approach (as will be illustrated in this paper).
Most geotechnical analyses in general practice are treated as deterministic. These involve analyses using representative values of design parameters, usually an average or the lowest value obtained from field and/or laboratory test results, and application of a suitable factor of safety to arrive at an allowable loading condition
[2]. There are a number of various solutions for the calculation of soil bearing capacity. The fist method was developed by Terzaghi in 1943.
Field tests in Canada by Meyerhof (1963) lead to modification factors. Finally, Brinch Hansen in
Denmark (1970) and Vesic in the USA modified these factor to a greater refinement.
The solution according to the „Regulations for the foundation of building structures (1990)“ is based on the Brinch-Hansen’s extended solution for eccentric and inclined load, from which directly calculated allowable bearing capacity ( q a
): q a
=
V
A a
′
= 0 , 5 ⋅ g ⋅ B ⋅′ N g
⋅ s g
⋅ d g
⋅ i g
+
( c m
+ q
0
⋅ tan ϕ m
)
⋅ N c
⋅ s c
⋅ d c
⋅ i c
+ q
0
Where: s c
, s q
, s
γ
- shape factors (depend on the ratio B/L ); d c
, d q
, d
γ
- depth factors (depend on the ratio D/B ); i c
, i q
, i
γ
- inclination factors (depend on the angle φ and ratio H/V ).
(1)
F c
F
φ
c m
= c F c tan ϕ m
= tan ϕ F ϕ
( F c
= 2 ÷ 3 )
(2)
(
F ϕ
= 1 .
2 ÷ 1 .
8
)
(3)
Bearing capacity factors N
γ
and N c
are obtained from the table for the reduced value of φ m
.
Depending on the eccentricity ( effective footing width (
( L’ = L –2× e
L
B’ = B –2× e
B e
B
, e
L
) the
) and length
) are determined. Effective area of footing ( A’ = B’ × L’ ) is part of the total area of footing ( A = B × L ), which is centrally loaded by vertical component ( V ) of the total load. Allowable vertical component of the resultant ( V case must meet the requirement: V a
≤ V a
.
) in this
In the deterministic approach, a single set of soil properties such as cohesion, friction angle, and unit weight are selected by the engineer based on some rational method. The analyst, although he may implicitly recognize the uncertainty in the various input parameters, selects single values for each parameter. The Figure 1 shows a schematic representation of a deterministic analysis. It can be seen that in the model, which was created for the purpose of deterministic analysis, each input parameter (x, y, z) is represented by the unique value (x m
, y also unique.
m
, z m
). The result of this analysis is
8
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Probabilistic Concept In The Computational Analysis Of The Soil Bearing Capacity
Figure 1.
Schematic representation of the deterministic analysis with 3 input parameters (x, y, z)
Based on the foregoing, it can be concluded that the deterministic analysis does not take into consideration the possible (and likely) variability of the assigned soil parameter values.
Uncertainties associated with geotechnical engineering can be divided into two categories: the inherent or natural uncertainty (aleatory), the uncertainties due to the lack of perfect knowledge
(epistemic) (Figure 2) [3].
Human error would be considered a third source of uncertainty, however, as it is difficult to isolate, its effects are usually included in compilations of statistics on aleatory uncertainty [4].
Figure 2.
Types of uncertainty associated with geotechnical engineering
Aleatory (from Latin aleator meaning “gambler” or alea meaning “die”) uncertainty represents the natural randomness of a soil property, such as the spatial variation of the soil layer properties cohesion and internal angle of friction. It is also called inherent variability. This type of uncertainty is unpredictable and therefore cannot be reduced or eliminated.
Epistemic (from Greek επιστημη meaning
“knowledge”) represents the uncertainty due to lack of knowledge of a soil characteristic.
Sampling uncertainty is present because the parameters are estimated from a limited set of data, while testing uncertainty is due to imperfections of an instrument or of a method to
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 9
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N. Davidovic, Z. Bonic, V. Prolovic register a quantity. Epistemic uncertainty can be reduced by collecting more experimental data, by improving the measurement methods and by using more refined models.
It is obvious that uncertainty is unavoidable, due to the lack of perfect knowledge or to the incomplete information about available data. For this reason the determination of uncertainty necessarily requires the application of probability theory, which quantifies and integrates uncertainty into the design process in a consistent manner [5].
In a probabilistic analysis the geotechnical parameters are treated as random variables.
A random variable (X) is a mathematical function defined on a sample space that assigns a probability to each possible event within the sample space.
Cumulative Distribution Function (CDF) describes the probability measures that a random variable
X takes on a value less than or equal to a number x, for every value x.
f
F
X
( x ) = P [ X ≤ x ] (4)
Probability Density Function ( PDF ) gives a description of a set of possible values x of a random variable X , and the probability ( P ) of each value.
X
( x ) = P [ X = x ]
Both functions are shown in Figure 3.
(5) variable X
Figure 3.
Probability functions of the random variable X: PDF (above), and CDF (bottom)
The position and shape of each Probability Density
Function is defined by its central moments:
The mean (expected) value (
is the sum of the probability of each first central moment.
μ x
) of a random possible outcome of an experiment multiplied by its value. The mean value is also referred to as the
Another important characteristic of a random variable is its measure of dispersion or variance
( σ x
2 ), also referred to as the second central moment.
A more understandable measure of dispersion is the standard deviation ( root of the variance.
σ x
) given by the square
As it is hard to specify whether the dispersion of a variable is large or small only on the basis of the standard deviation, it is more convenient to use the coefficient of variation ( COV ), which is defined as the ratio of the standard deviation over the mean value of the random variable:
COV = σ x
/ μ x
(6)
Among soil properties, unit weight, cohesion, and internal angle of friction are most directly used to evaluate bearing capacity of shallow foundations in many available methods. The variation of a parameter is described with the coefficient of variation ( COV ) of its distribution.
Typical values of COV of soil properties taken from some publications are listed in Table 1 .
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Probabilistic Concept In The Computational Analysis Of The Soil Bearing Capacity
Table 1.
Coefficient of Variation of Soil Properties
Parameter
Unit weight ( γ )
Type of soil
Internal angle of friction ( φ’ )
Cohesion ( c u
)
Sands
Clay
Clay and sand
Sandy soil
Clays
In probabilistic analyses soil parameters as random variables are usually described using normal and lognormal probability distribution functions.
COV (%)
1 - 10
5 - 10
3 - 7
2
5 - 15
12 - 56
5 - 15
25 - 30
20 - 50
References
Lee et al. (1983)
Lumb (1974)
Duncan (2000)
Christian et al. (1994)
Lee et al. (1983)
Lee et al. (1983)
Phoon & Kulhawy (1999)
Lee et al. (1983)
Lumb (1974); Lee et al. (1983)
The normal distribution is specified by a mean
( μ ) and standard deviation ( σ ). The PDF and
CDF for a normal distribution are illustrated in
Figure 4 .
Figure 4.
Normal distribution: (a) PDF, (b) CDF
The lognormal distribution is used when logarithm of the random variable is described by a normal distribution ( Figure 5 ). It is often used to describe non-negative variables (eg soil strength) and parameters that vary significantly (eg coefficient of permeability) .
Figure 5.
Lognormal distribution: (a) PDF of lnX, (b) PDF of X
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 11
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N. Davidovic, Z. Bonic, V. Prolovic
While the deterministic approach assumes that the variability in the underlying soil is insignificant, the probabilistic design method accounts primarily for the variability of the soil properties such as internal angle of friction ( φ ), cohesion ( c ), unit weight ( γ ), etc..
If the input parameters are uncertain, the result of the analysis is necessarily uncertain.
That is, the results of any analysis based on input parameters represented by probability distributions is itself a probability distribution. The
Figure 6 shows a schematic representation of a probabilistic analysis with 3 input parameters.
In this case the analyst explicitly represents the input parameters as probability distributions, and propagates the uncertainty and variability through to the result, such that the result itself is also a probability distribution.
Figure 6.
Schematic representation of the probabilistic analysis with 3 input parameters (x, y, z)
For the probabilistic analysis of the soil bearing capacity method based on Monte
Carlo Simulation ( MCS ) is the most suitable, because the procedure is direct and does not require detailed knowledge of probability theory.
This method is close to the real answer and is therefore used as reference for comparison with other probabilistic methods results.
probability distribution function ( PDF ) and any correlation exists between that input variable and the other available input variables;
The generated input values from Step (1) are entered into the Equation (1) and calculates the deterministic value of soil bearing capacity;
Common procedure of probabilistic analysis of bearing capacity of the soil, which is based on Monte Carlo Simulation ( MCS ), can be represented through the following steps [6]:
Steps (1) and (2) are repeated a number of times
(hundreds or thousands) as part of the MCS, until certain acceptable convergence of the results is met. Convergence is deemed to have occured if the change in the statistics describing the distribution of predicted bearing capacity is 1.5% or less;
For each of the bearing capacity input variables
( c’ , φ’ , γ , B , D ) a random value is generated in relation to parameter uncertainty of the input mean value, coefficient of variation ( COV ),
All simulation results (the calculated values of soil bearing capacity) are used to determine the cumulative distribution function ( CDF ) from which
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Probabilistic Concept In The Computational Analysis Of The Soil Bearing Capacity the values of soil bearing capacity associated with target reliability levels of 90% and 95%
(usually needed for design) can be estimated.
Schematic representation of the abovementioned Monte Carlo Simulation is shown in
Figure 7.
Figure 7.
Schematic representation of Monte Carlo Simulation
In order to illustrate the procedure set out above, the probabilistic analysis for bearing capacity of strip footing is investigated. The input data are as follows:
B = 2.0 m ; D f
= 1.5 m ; γ = 18 kN/m 3 ; μ c
= 5 kPa ;
μ
φ
= 30° ; COV c
= 27% ; COV
φ
= 10% [7].
deterministic value of bearing capacity, are shown in Figure 8 .
Figure 8.
Results of Monte Carlo Simulation – CDF of calculated values of soil bearing capacity
The deterministic ultimate bearing capacity is determined by applying Terzaghi Equation: q u
= c ⋅ N c
+ q
0
⋅ N q
+ 0,5 ⋅ ã ⋅ B ⋅ N
ã
Calculated value of q u
is equal to 1067 kPa.
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014
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N. Davidovic, Z. Bonic, V. Prolovic
For required reliability levels of 90% and
95%, the corresponding bearing capacities are estimated from CDF to be equal to 730 kPa and 658 kPa, respectively. On the basis of these values, equivalent factors of safety of 1067/730 = 1.5 and 1067/658 = 1.6 are calculated.
These results indicate that the factor of safety of 3 (usually required in the deterministic analysis of the soil bearing capacity) is conservative. The results also indicate that the uncertainty in the values of c’ and φ’ can significantly affect the bearing capacity of strip footing and therefore can not be ignored, but should be quantified as described above and introduced in the computational analysis of the soil bearing capacity.
Although traditional deterministic approach in the computational analysis of the soil bearing capacity is simple and straightforward, it does not consider the uncertainty in the various input parameters in a rational manner. In order to incorporate these variations, alternative probabilistic analysis is performed. In this approach, input soil parameters are treated as random variables and the influence of these input random variables on the output random variable (allowable bearing capacity) is studied.
In this paper, probabilistic approach that utilizes
Monte Carlo Simulation ( MCS ) was used to obtain bearing capacity of strip footing. It was shown that the applied probabilistic method enables bearing capacity to be quantified in the form of a cumulative probability distribution function ( CDF ) that provides bearing capacity predictions for reliability levels of 90% and
95% (usually required in practice). The results also indicate that the suggested factor of safety of 3 (typical for deterministic analysis) is conservative.
All the above mentioned facts indicate the importance of adopting probabilistic analyses to calculate bearing capacity of strip footings.
Based on the example presented in this paper, it can be concluded that the probabilistic analysis is not significantly more demanding (in terms of the required level of knowledge, the time required to execute the analysis, the scope of research, inputs, costs, etc.) than traditional deterministic analysis, and that in turn provides great opportunities for the modeling of uncertainty and variability of parameters, which are included in the calculations.
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[1] Meyerhof, G.G. (1951): „The ultimate bearing capacity of foundations“,
Geotechnique, Vol.2, No.4, pp. 301-331.
[2] Griffiths, D. V., Fenton, G. A., Manoharan,
N. (2002): „Bearing capacity of rough rigid strip footing on cohesive soil: probabilistic study“, Journal of Geotechnical and
Geoenvironmental Engineering; 128: pp.
743–755.
[3] Baecher, G. B., Christian, J. T. (2003):
Reliability and Statistics in Geotechnical
Engineering , John Wiley & Sons.
Chichester, West Sussex, England;
Hoboken, N.J.
[4] Davidović, N., Prolović, V., Stojić, D.
(2010): „Modeling of Soil Parameters
Spatial Uncertainty by Geostatistics“, Facta
Universitatis, Series: Architecture and Civil
Engineering, Vol. 8, N° 1, pp. 111 – 118.
[5] Russelli, C. (2008): „Probabilistic Methods applied to the Bearing Capacity Problem“,
PhD Thesis, Institut für Geotechnik der
Universität Stuttgart.
[6] Davidović, N. (2013): „Probabilistic concept and its application in geotechnical stability analyses“, PhD
Thesis, University of Niš, Faculty of Civil
Engineering and Architecture.
[7] Shahin, M. A., Cheung, E. M. (2011):
„Probabilistic Analysis of Bearing Capacity of Strip Footings“, ISGSR 2011 – Vogt,
Schuppener, Straub & Bräu (eds),
Bundesanstalt für Wasserbau, pp. 225-230.
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Evaluation Of The Layer Influence On Dynamic Behavior Of Structure Over Layered Soil
Authors
Josif JOSIFOVSKI
Ph.D. Assistant Professor
University of Ss Cyril and Methodius
Faculty of Civil Engineering jjosifovski@gf.ukim.edu.mk
A simple coupled finite with boundary element model of structure over soft soil layer is analyzed and dynamic behavior of the structure is determined. The three dimensional dynamic soil-structure interaction analyses is performed in frequency domain with the help of the SSI-3D program package. Dynamic interaction between structure and layered soil is examined for internal (seismic) and external (wind) excitation.
Using the Substructure method the problem of soil-structure interaction is modeled using different mathematical formulations, namely the
Soil which is discretized by Boundary element method and the Superstructure by Finite element method. Such a coupled system at the common interface is used with computational procedure to determine the dynamic response.
Since solution is in frequency domain, a signal of inverse Fourier Transformation is used to convert the results back into time domain. The objective has been to evaluate the influence of a layer on the dynamic response of singe frame excited by wind and earthquake. The obtained results show that the behavior of the influence dynamic soil stiffness and damping in certain cases, such as this one with layer over stiffer half-space, is significant, especially in the case of seismic excitation.
Soil - structure interaction, coupled
FEM - BEM model, linear elastodynamic analysis, earthquake, wind excitation.
Soil-structure interaction
It is often a practice with respect to structural response to earthquakes, for very stiff foundation systems to apply that the excitation to the structural support points as free-field earthquake motions at that location, in other words, to neglect the effects of Soil-Structure
Interaction (SSI). In reality, however, the structure always interacts with the soil to greater extent during earthquakes, imposing soil deformations that cause the motions of the structure soil interface to differ from those would have been observed in the free-field.
Therefore, final response of the structure will be dependent of the dynamical characteristics
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 15
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Josif Josifovski of the soil. The nature and amount of this interaction depends not only on the soil stiffness but also on the stiffness and mass properties of the structure.
To derive the basic equation of motion, it’s sufficient to examine a simple structure embedded in soil for earthquake excitation.
The base of structure usually a basement with adjacent walls is first assumed at to be flexible. The structure is represented schematically by Fig. 1. Subscripts are used to denote the nodes of the discretized system.
The nodes located on the soil–structure interface are denoted by “b” (for base), the remaining of the structure by “s”.
The dynamic stiffness of the system will be comprised of stiffness portions from the both substructures, the actual structure or better superstructure and the soil with excavation, where on the other hand the soil substructure with excavation, is denoted by “g” (for ground).
The dynamic equations of motion will be formulated in the frequency domain. The amplitudes of the total displacements are denoted by { u t }, which are functions from discrete values of the circular frequency
. The total (superscript t) expresses that the motion is referred to an origin that does not move, reference point. The order of this vector equals the number of dynamic degrees of freedom of the total discretized system, but this vector can be decomposed into subvectors { u s t } and {u b t }.
motion takes the form given by Equ.1.1, in matrix notation.


[ ]

K s bb
 
K g bb

 
 

{ }
{ }




 K g bb
{ }

{ } 

(1.1)
Both substructures contribute to the dynamic equilibrium equation in the nodes “ b” lying on the soil–structure interface. The contribution of the soil is through interaction forces of the soil, thus depend on the relative motion to { u b g } .
For other types of dynamic excitation, for example external excitations like the wind load, which is a second loading case in this study. The equation of motion is transformed in the following form, (Equ.1.2),


[ ]
[ ]
[ ]

K s bb
 
K g bb

 
 
{ }
{ }

  u b 
=


{ }
{ }

 (1.2)
When modeling certain problems from the practice a usually made assumption is to discretaize the base of the superstructure
(sometimes the whole basement with adjacent walls) as a rigid system. This additional kinematic condition makes the whole analysis to be simpler while reducing some degrees of freedom. Therefore, the total motion at the base function of the total rigid-body motions of a point “ O” u b t b t
{ } which can be expressed as a as,
{ } transformation matrix.
= [ ] { }
, hence
[ ]
-
The interaction effect associated with the structure is termed kinematic interaction (Equ. 1.3) and the corresponding mass-related effect is called inertial interaction (Equ. 1.4). The equation of motion for the problem of soil-structure interaction can be divided theoretically into kinematic part, where the system mass is set to zero,



 (
(
+ ξ
+ ξ
)
)
[ ]
[ ] ( +
( i ξ
+ ξ
) 

K s bb
) [ ]
  K g bb

 

 
 

{ }
{ }

 


 K
{ }
 g bb


{ } 

(1.3) and inertial part of the interaction given by
Equ.1.4,
Figure 1.
Soil - structure interaction (a) System and (b) Model
For seismic (internal) excitation, the first loading case in this study, hence that the nodes which are not directly in contact with the soil (the subscript “s” stands for structure) are not loaded. Therefore, the equation of
16
|
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



[ ]
[ ]
[ ]

 K s bb
 
K g bb






=

{ }
{ }


 u i b


= ω 2






M s bb

 


 
{ }
{ }


 u b k


(1.4)
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Evaluation Of The Layer Influence On Dynamic Behavior Of Structure Over Layered Soil
The term ω 2
{ } from Equ. (1.4) represents the negative - acceleration amplitude of the kinematic motion. For the inertial-interaction part of the dynamic response, the load vector, thus consists of the negative inertia loads (mass of structure time’s acceleration) determined from { u k } .
The physical interpretation of the kinematic and inertial part of interaction and the computational procedure through the use of this concept is illustrated in Fig. 2 . Summarizing, in the first step by omitting the mass of the superstructure and subjecting the dynamic system to the load vector (which depends on the free-field motion), the kinematic motion is calculated first.
This means that the loading is applied in the actual dynamic analysis calculating the inertial interaction part. Thus, the two subsystems can be assumed to interact with each other with the feedback from the superstructure to the dynamic model from the first step.
Figure 2.
Computational procedure of SSI
There are in general two methods available for dynamic analysis in solving a Soil-Structure
Interaction problem:
• Direct (both the superstructure and the soil are analyzed as one coupled system);
• Substructure (soil-foundation mechanism and the superstructure are represented as two independent mathematical models or substructures);
The Direct method is probably the easier method to analyze soil-structure interaction for seismic excitation to modeling a significant part of the soil around the embedded structure and applying the free-field motions at the fictitious boundaries.
However the number of dynamic degrees of freedom in the soil region will be very height resulting in significant running time. As the law of superposition has to be implicitly assumed for the soil-structure interaction analysis, it is computationally more efficient to use the substructure method. Consequently, the idea of the substructure method arises as better method for solution of the SSI problem, but this approach allows one to use completely different methods in discretizing the structure (e.g. FEM) and the soil substructure for example with the Boundary
Element Method (BEM), Fig. 3. This combination will make the whole analysis less computer demanding.
Figure 3.
Coupling of the subsystems
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 17
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Josif Josifovski
In the present analysis the following assumptions are made;
• Dynamic analysis through Substructure method (combination of FEM+BEM);
• Small displacements;
• 3D isotropic linear elastoplastic analysis
(assumes the law of superposition to be valid);
• Computation performed in frequency domain using the computer program SSI-3D.
The problem of soil structure interaction is analyzed on a simple model of frame structure over soft soil layer, given with Tab. 1 and Fig. 4.
a)
18
|
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Figure 4.
Analyzed problem of frame structure on layered soil (a) geometry and (b) Excitations
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Evaluation Of The Layer Influence On Dynamic Behavior Of Structure Over Layered Soil
Table 1.
Numerical discretization
FEM
Total
Number of columns
3d beam element
Number of girders
3d beam element
BEM
Total
Number of Foundations
Foundation 1,2,3
Number of Surfaces (d=0)
Surface 1
Number of Interfaces
Interface 1
From material and geometrical point of view more details are presented in the fillowing text.
The soil is assumed to be isotropic linearelastic semi-infinite continuum. The 3D model of soilfoundation mechanism is discretized using BEM with soil material characteristics given in Tab. 2.
Hence, (1) - designates the properties of the soil layer and (2) - of the half-space ;
34 Elements, 21 nodes
6
3x6 Elements, 6 x 6 nodes
4
4x4 Elements, 8x4 nodes
833 Elements,1003 nodes
3
85x3 Elements, 96x3 nodes
1
288 Elements, 360 nodes
1
264 Elements, 330 nodes
As an interesting case example a layer over stiffer half-space is considered.
In Tab.3 the geometrical characteristic and material properties of the foundation system represented by three spread foundations are presented.
Table 3.
Geometrical and material properties of the
Foundation
Layer Ec u
GN/m 2 r
/ kg/m 3
2a 2b d m m m
Ix m 3
Iy m 3
Iz m 3
Found 31.5 0.4 2500 1 1 0.33 0.076 0.076 0.137
Figure 5.
Model of the soil substructure
Table 2.
Soil material properties
Layer G u r c s c p
MN/m 2 / kg/m 3 m/s m/s / x
1
2
45 0.4 2000 150 367
125 0.4 2000 250 612
0
0
/ b c s
/c p
/
0 0.41
0 0.41
For the sake of the simplicity, this analysis considers
3D Structure represented by single 3D frame model.
The superstructure is discretizied using FEM with
3D beam elements see in Tab. 4 and Fig. 6.
Table 4.
Geometrical and material properties of the
Structure
Layer Ec u r x b d Ix Iy Iz
GN/m 2 kg/m 3 % m m m m 3 m 3 m 3
Colum 31.5 0.25 2500 5 0.4 0.4 0.0036 0.0021 0.0021
Girder 31.5 0.25 2500 5 0.4 0.5 0.0055 0.0042 0.0027
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 19
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Josif Josifovski
(86) and in combination with the real measured event, Tab. 5 and Fig. 7.
The recorded data of the wind-storm event are courtesy of “Arbeitsgruppe Aerodynamik im Bauwesen”-RUB. The height distribution of the wind load which can be seen in Fig.4
Figure 7.
Time history of the wind (storm) excitation
Figure 6.
Model of the complete Structure
(coupled BEM+FEM model)
When subjected to dynamic loads, the structure oscillate in a way that depends on the nature and de-formability of the supporting ground, the geometry and inertia of the foundation and superstructure. Such an excitation may be in the form of support motion due to waves arriving through the ground during earthquake, an adjacent explosion or the passage of a train, or it may result from dynamic forces imposed directly or indirectly on the foundation from operating machines, vehicles moving on the top of the structure, or wind. In this analyses two types of dynamic loads will be examined, wind load (case of external excitation) and seismic
(or internal excitation).
The wind load is modeled under the German construction standards and regulations DIN 1055 T4
Table 5.
Characteristic values of simulated wind load
Nt
/
1024
Wind
Lt sec.
100 dt sec.
0.1
t sec.
51.2
Time domain ů(t)max m/s
46.8
* fast Fourier Transform (Cooley & Tukey 1965)
20
|
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The ground motion that affects the building can be summarized in a conceptual way as follows. The waves that create motion emanate from the line of fault rupture, and so approach the building from a given direction.
The nature of the seismic waves and their interactions is such that actual movement at the ground will be random: predominantly horizontal, often with some directional emphasis and sometimes with a considerable vertical component.
In this investigation, the foundations of structure are assumed to be rigid. It is traditionally assumed that most of the energy of the earthquakes is transmitted through waves which produce horizontal movements of the soil and structure, therefore the horizontal free field excitation of some recorded earthquake is a usual assumption. For this analysis we choose the El Centro earthquake described in
Tab. 6 and Fig. 8.
q(t)max kN/m2
1.45
Nf
/
512
Frequency domain df fmin
Hz
0.008
Hz
0 fmax
Hz
5
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Evaluation Of The Layer Influence On Dynamic Behavior Of Structure Over Layered Soil
Table 6.
Characteristic values of simulated earthquake
Nt
/
Lt
Earthquake sec.
dt sec.
t sec.
ü(t)max m/s2
1024 100 0.1
51.2
* fast Fourier Transform (Cooley & Tukey 1965)
2.63
time domain u(t)max Nf cm
10.86
/
512 df
Hz
0.008
frequency domain fmin fmax
Hz
0
Hz
5
Figure 8.
Acceleration time history “El Centro” earthquake
The program SSI-3D with the described computational approach and method should realistically sim-ulate and predict the dynamic behavior or response of the structure supported on rigid foundations embedded in soil layer over half-space.
After fFT* the load is transformed into frequency domain and one obtains the Fourier amplitude spectrum of the simulated wind load. In the next step the whole (coupled) system is solved in the frequency domain. The time history response (in the highest point node -21, see
Fig.4) of the structure is deter-mined (Fig.9).
For comparison the same example is analyzed with the program ANSYS (different pro-gram is used for the purpose of calibration) where fixed base conditions assumed, in other words, the effects of soil-structure Interaction (have been ne-glected (Fig.10).
The results confirm the fact that in reality the structure always interacts with the soil and the right way to investigate the dynamic behavior of certain structure will be to include the influence from the soil substructure.
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 21
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Josif Josifovski
Figure 9.
Time history of the structure response due to wind load with SSI
Figure 10.
Time history of the structure response due to wind load without SSI
If both responses (w/ SSI and w/o SSI ) are compared it is evident that the system with
SSI has significantly larger damping due to the inclusion of the soil damping. The system with SSI has shows a greater displacement amplitude due to the influence of the layer stiffness which is considerably lower in respect to the one of the structure.
The problem of soil structure interaction for earthquake excitation produces a complex relation between the two subsystems. In this case the effects of SSI are even more pronounced because the soil ground does not only serve to reduce the building oscillations but at the same time also to introduce the load to the structure.
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Evaluation Of The Layer Influence On Dynamic Behavior Of Structure Over Layered Soil
Figure 11.
Time history of the structure response due to seismic load with SSI
Figure 12.
Time history of the structure response due to seismic load without SSI
Comparing the response curves visually the difference is evident with higher displacement amplitude sin the case of a system w/ SSI.
Moreover, the system w/o SSI presents very stiff response due to the fact that the flexibility of the soil is not considered.
The inclusion of the soil introduces two main modifications in the dynamic behavior of the
Structure:
In general it can be concluded that the difference in the response of structures due to diverse approaches
(Fig.9 and Fig.11 systems w/ SSI and Fig.10 and Fig.12 systems with fixed boundary conditions = w/o SSI).
• First the flexibility of the soil reduces the natural frequencies of the Structure (or makes the structure more flexible) in comparison to the one with fixed base condition, on the other side, additional damping is included due to wave radiation from the oscillating Structure.
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 23
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Josif Josifovski
• The effect of the additional dissipation of energy through radiation damping is not sufficient to make up for the added flexibility to the support nodes. Therefore the response of the soil-structure system is almost double the value of the structure with fixed base.
The ratio between the natural frequencies of the system w/ SSI over w/o SSI characterizes the inertial interaction. If the ratio is lower then one then this indicates high internal interaction, while a ratio equal to one means no inertial interaction. In soft soil conditions, structures on shallow foundations usually display a high inertial interaction, where on the other hand structures on deep foundations display almost no inertial interaction.
The nature and amount of the interaction strongly depends on the dynamic soil stiffness, especially when there is a presence of softer soil layers we can experience the amplification of the incoming seismic waves. Therefore, it is from great importance when reliable types of structures had to be properly analyzed including the effects of the soil structure interaction. In this context this study proves that the influence if SSI can be significant which can be explained by the presence of softer soil layer over second stiffer soil stratum, from which the emitted waves of the oscillating foundation are partially reflected and returned back to their source. In this process the waves are amplified thus structure will experience changes in its overall response.
The degree of interaction is very high in the case of shallow foundation systems and depends of the degree of activation in the response which is dependent from the excitation frequency, amplitude and the soil stiffness.
In the cases of some simple structure (classified in the lowest category) the soil influence can be neglected if logical explanation is offered, but in all other cases this effects have to be considered especially when loose (soft) soil layer is present.
Similar provisions are included in Eurocode 8 which deals with the seismic structural design.
1. Clough, R. W. & Penzien, J.: Dynamic of
Structure, McGraw-Hill, Inc., New York 1993,
ISBN 0-07-011394-7
2. Cooley, J. W. & Tukey, J. W.: An Algorithm for the Machine Calculation of Complex Fourier
24
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Page
Series, Mathematics of Computation, Vol. 19,
No. 19 (1965) 297 – 301.
3. Dominguez J.: Boundary Elements in
Dynamics, Computational Mechanics
Publications, Southampton 1993, ISBN
1-85312-258-0
4. Gazetas, G. & Mylonakis, G.: Seismic Soil-
Structure Interac-tion: New Evidence and
Emerging Issues. Geotechnical Special
Publication 75, Geotechnical Earthquake
Engineer-ing and Soil Dynamics III, Vol. 2
(1998) 1119 – 1174.
5. Gazetas, G.:Analysis of Machine Foundation
Vibrations: State of Art, Soil Dynamics and
Earthquake engineering, Vol.2, No.1 , pp 2-42, 1983
6. Hirschauer, R.: Kopplung von Finiten
Elementen mit Rand-Elementen zur
Berechnung der dynamichen Baugrund-
Bauwerk Interaction, HEFT 31 , des
Grundbauinstitutes der Technischen Universität
Berlin , Berlin 2001, ISBN 3-7983-1883-2
7. Huh, Y. & Willms, G.: Bauwerk-
Baugrund-Wechselwirkung, Ermitlung des Schwingungsverhaltnes starrer
Gründungskörper mit Randelementmethode,
SFB 151-Berichte Nr.2, Ruhr-Universität
Bochum, November 1984
8. Huh, Y.: Anwendung der
Randelelementmethode zur Untersuchung der Dynamischen Wechselwirkung zwischen
Bauwerk und geschichtetem Baugrund,
Dissertation am Lehrstuhl Theorie der
Tragwerke, Ruhr-Universität Bochum,
Mittelung Nr.86-13, Dezember 1986
9. Kythe, Prem K.: An Introduction to Boundary
Element Meth-ods, CRC Press, Inc., Boca
Raton Florida 1995, ISBN 0-8493-7377-8
10. Meskouris, K.: Structural Dynamics (Models,
Methods, Examples), Erst&Sohn, Berlin 2000,
ISBN 3-433-01327-6
11. Peterson, C.: Dynamik der Baukonstruktionen,
Friedr. Vieweg & Sohn Verlagsgesellschaft mbH, Brainschweig / Wiesba-den 1996, ISBN
3-528-08123-6
12. von Estorff, O.: Zur Berechnung der dynamischen Wechselwirkung zwichen
Bauwerken und Ihrer Umbgebung mittels zeitabhänigiger Randintegralgleichung,
Mitteilung Nr.86-10 des SFB 151, Ruhr-
Universität Bochum, Dezember 1986
13. Wolf, John P.: Dynamic Soil Structure
Interaction, Prentice-Hall, Inc., Englewood
Cliffs, New Jersey 1985, ISBN 0-13-221565-9
14. Wolf, John P.: Foundation Vibration Analysis using simple Physical Models, Prentice-Hall,
Inc., Englewood Cliffs, New Jersey 1994, ISBN
0-13-010711-5
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Laboratory Experiments On Soil Dynamic Characteristics Of Npp Site
Authors
Vlatko SESOV
Ph.D. Full Professor
University of Ss Cyril and Methodius
IZIIS Skopje vlatko@pluto.iziis.ukim.edu.mk
Vera GADZA
B.Sc. Civil Engineer
University of Ss Cyril and Methodius
IZIIS Skopje vera@pluto.iziis.ukim.edu.mk
Irena ZAFIROVA
B.Sc. Civil Engineer
University of Ss Cyril and Methodius
IZIIS Skopje irena@pluto.iziis.ukim.edu.mk
Juliana BOJADZIEVA
M.Sc. Assistant
University of Ss Cyril and Methodius
IZIIS Skopje jule@pluto.iziis.ukim.edu.mk
Kemal EDIP
M.Sc. Assistant
University of Ss Cyril and Methodius
IZIIS Skopje kemal@pluto.iziis.ukim.edu.mk
The paper summarized the results from experimental investigation on dynamic properties of the soil samples collected from
NPP site. Recommendations and procedures given in IAEA Safety Standards Series and US
NRC Regulatory Guides have been followed.
A series of strain controlled cyclic simple shear tests have been carried out on dry soil samples ranging from small-medium to large shear strain levels. Simple shear condition is the one of the most representative strain condition of in-situ ground during seismic event and very suitable for evaluating ground response. Stable hysteresis curves from the tests enable clear definition of the nonlinear reduction of the shear modulus and damping ratio f the investigated soil samples
Cyclic simple shear test, shear modulus, NPP
The earthquake case histories have revealed that the local geological conditions have specific effect on the characteristics of ground motion upon the surface during the earthquake.
Depending on the characteristics of the soil layers and the characteristics of the excitation at the level of the seismic subsoil, the mentioned effect can be higher or lower. It is expressed through the amplitude-frequency modification of the surface seismic excitation in respect to the corresponding excitation at the level of seismic bedrock. The influence of soil layers on the amplitude and frequency content of ground motion can be substantially. Therefore investigation of the subsurface conditions at a nuclear power plant site is very important at all stages of the site evaluation process.
The investigation programme depend of the stage site evaluation should provide necessary information for appropriate characterization of the subsurface ground conditions . To perform reliable and realistic site specific seismic response analysis, which actually define the seismic input for the further seismic assessment of the structures and equipment performance of nuclear power plant, static and dynamic material properties of soil and rock profiles should be obtained. Shear moduli and material damping for each soil layer at the site are one
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 25
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V. Shesof, V. Gadza, I. Zafirova, J. Bojadzieva, K. Edip the most important soil parameters for seismic site response analysis. In cases of ground response involving no residual displacement the response is determined mainly by the shear modulus and damping characteristics of the soil under reasonably symmetrical cyclic loading condition. A comprehensive research studies of the factors affecting the shear moduli (G) and damping (D) of soils have been made by many researchers (Seed and Idriss 1970; Hardin and Drnevich 1972; Lee and Finn 1978; Seed et al.1986; Vucetic and Dobry 1991; Rollins et al.1998; Vucetic et al.1998; Darendeli 2001;
Roblee and Chiou 2004; Zhang et al. 2005). In these studies it was suggested that the primary factors affecting moduli and damping are: strain amplitude ( g) , mean effective principal stress
( s m
’), soil type and plasticity index (PI). Other factors that appear to be lees important are: frequency of loading, number of loading cycles, overconsolidation ratio and degree of saturation.
In Fig. 1 a typical shear stress-strain relationship is shown for fine-grained soil materials at undrained cyclic loading. The maximum shear modulus (G max
) refers to modulus at very small strains in the range of g =10 -4 and less. Taking into account the range of ( g ) to which (G in which the soil behaves as a linear material,
(G max max
) refers
) could be defined as a constant which in that range correlates stresses and strains.
The secant modulus (G) has been defined through the secant of the hysteresis loop which passes through the extreme points, as shown in Fig. 1.
The secant modulus (G) could be interpreted as an average modulus in the (± g ) domain. In this way it is used for definition of an equivalent linear model for which (G) is a constant correlating ( t ) and ( g ) in amplitude range from ( -g ) to ( +g ).
The material damping ratio (D) represents the energy dissipated by the soil, Figure 2.
Mechanisms that contribute to material damping are friction between soil particles, strain rate effect, and nonlinear soil behavior. The hysteretic damping ratio can be calculated as it shown on
Fig.2 where:
A
H
= the area enclosed within the hysteresis loop
( t-g ) representing the damping energy ( D W)
A oab
= area of triangle (Oab) representing the strain energy (W).
Theoretically, there should be no dissipation of energy in the linear elastic range for the hysteretic damping model. However, even at very
26
|
Page low strain levels, there is always some energy dissipation measured in laboratory specimens.
At higher strains, nonlinearity in the stress strain relationship leads to an increase in material damping ratio with increasing strain amplitude.
Figure 1.
Definition of shear moduli G
Figure 2.
Definition of damping D
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Laboratory Experiments On Soil Dynamic Characteristics Of Npp Site
done by Hardin and Drnevich, 1972, regarding the level of mean principal effective stresses and
1972, regarding the level of mean principal effective stresses and number of cycles on G.
They obtained a negligible difference of (G) for testing even a number of cycles exceeding
exceeding 100. Effects of mean principal effective stresses can be clearly seen, increase of effective stresses leads to higher values of G.
Figure 3.
Effect of  ,  , , ' and number of cycles N upon G (Hardin and Drnevich, 1972)
Based on a review of laboratory test data published in the literature, Vucetic and Dobry
1991 [4] published stress-strain curves, which are shown in Fig. 4. The effect of soil plasticity
index of the soil increases: G
faster with over-consolidation ratio OCR;
G increases faster with geological age; the normalized curve G/ G





versus the curve of damping (D) versus g g first two effects relate to small strains
rises;
falls and G degrades less after N cycles of a given g g
. The
=10 -4 while last three effects relate to larger strains above
 =10
-2 .

increases
falls and G degrades less after N cycles of a given  . The first two effects relate to small strains  =10 -4 g
-2
=10
increases max max
-4
Figure 4.
Effect of Cyclic stiffness degradation on G/G max
for soils of different plasticity indices
(Vucetic & Dobry, 1991)
The current state of practice for determining G and D for ground response analysis involves: measuring shear wave velocity V max
(Vucetic & Dobry, 1991)
S
in the field
for soils of different plasticity indices
(Vucetic & Dobry, 1991)
27
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P a g e

V. Shesof, V. Gadza, I. Zafirova, J. Bojadzieva, K. Edip
The current state of practice for determining G and D for ground response analysis involves: measuring shear wave velocity V
S
in the field and estimating or measuring the variation of
G and D with g primarily in the laboratory. Insitu geophysical survey is one of the widely used methods for measurement of Vs which is directly related to small strain shear modulus
G max
. (G max s
2 = r V ). Even the measuring accuracy of conventional laboratory tests at small strain amplitudes has been improved most of the studies strongly suggested to used the G max
determined by in-situ measurements.
The selection of testing techniques for measurement soil properties requires careful consideration and understanding of the specific problem at hand [2]. Efforts should always be made to use laboratory tests that replicate the initial in-situ stress conditions and the anticipated cyclic loading conditions as closely as possible.
Procedures and techniques for several laboratory methods like triaxial testing method, resonant column method for evaluation of soil properties is given by many national and international standards (ASTM, JGS, BS etc).
Cyclic simple shear test method for definition of shear moduli and damping of soil is not fully covered by any standard. Even it is generally difficult to perform simple shear tests in ideal conditions because of several disadvantages
(non-uniformity of strain distribution and normal stresses, limitation due to K
0
condition) still the cyclic simple shear test is capable of reproducing earthquake stress condition much more accurately than is the cyclic triaxial test.
Therefore it is not justified for the cyclic simple shear tests to set aside from standardized method for evaluation of dynamic properties of soils. Some of the disadvantages of cyclic triaxial and cyclic simple shear tests can be avoided by cyclic torsional shear tests which allow isotropic and anisotropic initial stress conditions and can impose cyclic shear stresses on horizontal plane with continuous rotation of principal stress axes 5 .
The present study adresses applicability and usefulness of cyclic simple shear tests for evaluation of soil properties under dynamic loading at wide range of shear strain.
Summarized in this paper is laboratory procedure for estimating the variation of G/
28
|
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Gmax and D with g for representative soil sample collected from one NPP site. It should be emphasized that normalized shear moduli and damping were subject of this study while other soil investigations regarding site evaluation, ground conditions etc. are not discussed here.
The NPP site is characterized by several soil layers. The soil deposit is stratified up to 35-
30 m depth. Below 35 m engineering bedrock consisted of marly clay, marls and sandstone is located. The nonlinear behavior of these layers will have influence on ground motion parameters during the expected earthquakes and the dynamic interaction between soil and the facilities of the nuclear power plant. For analysis and definition of these effects, it is necessary to define dynamic characteristics of the site soil layers. For this purpose, laboratory testing of samples taken from characteristic soil layers was performed at
University “Ss Cyril and Methodius”, Institute of Earthquake Engineering and Engineering
Seismology, IZIIS Soil dynamic laboratory in Skopje, Macedonia in order to define the strain compatible shear modulus values and damping values for representative soil layers.
The selection of the tested materials was made by careful examination of the quality of the received samples. Depending on the type, nature and conditions upon arrival at the laboratory the soil samples were tested as relatively undisturbed and disturbed samples. Handling and storage of samples was done following the procedures prescribed in US NRC Regulatory Guide 1.138 [9]. The identification markings of all samples were verified immediately upon their arrival at the laboratory. Tested as relatively undisturbed samples were those from the coherent materials and materials containing a high percentage of clay fractions. Samples from sand material were tested as reconstituted samples. The velocity of pouring rate, height of pouring and tapping energy were parameters which were applied during the preparation process to re-produce as close as possible insitu stress state [3].
Physical properties for soil materials subject to dynamic testing (grain size distribution, unit weight, specific gravity, water content, plasticity index etc..) were used from previous performed in-situ and laboratory investigations, Table 1.
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Table 1.
Physical properties of soil samples
Physical properties of soil samples
Lithological description
Bore-hole
Depth (m)
Exp.
No.
Vertical loading
 [kPa]
Dry unit weight Relative
 max
[kN/m 3 ]
 min
[kN/m 3 ] density
Dr [%]
Humidity w [%]
Mc 7
10.0-10.7
1 161 17.2 12.2
Silty clayey sand
2 165
3 161
17.2
17.2
12.2
12.2
60
65
62
Table 1.
Fine grain clayey sand
Mc 9
14.6-15.4
1 232
2 230
3 232
Lithological description
1
2
Exp.
No.
290
270
Medium grained sand
Medium grained sand
Clay
Silty clayey sand
Mc 2
19.0-19.5
Mc 7
1
2
1
2
340
337
Fine grain clayey sand
Mc 4
3
3
1
337
Mc 9
1 2 439
2
3
433
25.0-26.0
3
1 290
2
429
Medium grained sand
Mc 2
4
19.0-19.5
1 431
Mc 5 1
2
113 sand
5.3-5.6 2
3 337
1 120
Medium grained
Mc 5
Mc 4 3
1
2
3
4
127
19
Sandy Clay
Clay
9.1-9.4 2 1 194
Mc 5
3
3
192
1 19
Sandy Clay 9.1-9.4 2 194
3 192
15.5
15.5
15.5
10.3
10.3
10.3
78
79
79
Dry unit weight Relative
 max
[kN/m 3 ]
 min
[kN/m 3 ] density
Dr [%]
Humidity w [%]
12.2
64
15.5
12.2
12.2
10.3
10.3
78
67
67
18.4
10.3
13.3
65
66
19.7
69
18.4
18.4
18.4
18.4
18.4
18.4
13.0
13.0
13.0
13.3
13.3
13.3
13.3
67
67
65
66
69
68
68
19.7
19.5
29
28.3
27.2
21
29
28.3
27.2
21
19.5
19.5
19
and the control system which optimize dynamic range both in measurements and control of force and displacement. Two cylindrical samples both
Figure 5.
Schematic view of the cyclic simple shear apparatus CSSA
Figure 5.
Schematic view of the cyclic simple shear apparatus CSSA
5 | P a g e
Scientific Journal of Civil Engineering • Volume 3 • Issue 2 • December 2014
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 29
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Figure 5.
Schematic view of the cyclic simple shear apparatus CSSA
P a g e

V. Shesof, V. Gadza, I. Zafirova, J. Bojadzieva, K. Edip
The CSSA used in this study enables cyclic shear tests with constant volume and strain control; cyclic shear tests under constant vertical load with control of shear strains and cyclic shear tests with load control. shear moduli G. By using the obtained values for the shear moduli G for each strain level, established were the G diagrams of the relationships between the shear moduli and the shear strains.
The CSSA used in this study enables cyclic shear tests with constant volume and strain control; cyclic shear tests under constant vertical load with control of shear strains and cyclic shear tests with load control.
The cyclic simple shear tests on soil samples were performed in 7 series. Performed within each series were at least two to four experiments under equal loading and dynamic excitation conditions. Samples from 5 different sand soil layers were tested, Table 1. Dynamic excitation was applied in the form of short series of cyclic simple shear loads with frequency of 0.1 Hz by controlling the shear strains (strain control). The excitation was
V. Shesof, V. Gadza, I. Zafirova, J. Bojadzieva, K. Edip maximum amplitude of shear strains.
Permanent recording of horizontal displacements and shear stresses was performed throughout the experiment. In this way, the hysteretic relationships between the shear moduli and the shear strains.
 relationships were obtained for each strain level.
Ƴ
Ƴ
The cyclic simple shear tests on soil samples were performed in 7 series. Performed within each series were at least two to four experiments under equal loading and dynamic excitation conditions. Samples from 5 different sand soil layers were tested, Table 1. Dynamic excitation was applied in the form of short series of cyclic simple shear loads with
The CSSA used in this study enables cyclic shear tests with constant volume and strain control; cyclic shear tests under constant vertical load with control of shear strains and cyclic shear tests with load control.
displacements and shear stresses was
The cyclic simple shear tests on soil samples were performed in 7 series. Performed within each series were at least two to four experiments under equal loading and dynamic excitation conditions.
Samples from 5 different sand soil layers were tested, Table 1. Dynamic excitation was applied in the form of short series of cyclic simple shear loads with frequency of 0.1 Hz by controlling the shear strains (strain control). The excitation was applied step-by-step, with variation of the maximum horizontal displacements and shear stresses was performed throughout the experiment. In this way, the hysteretic t-g relationships were obtained for each strain level.
Ƴ
Dividing each value of the modulus G by G max corresponding to the least shear strain that
The dynamic shear moduli are defined as
V. Shesof, V. Gadza, I. Zafirova, J. Bojadzieva, K. Edip are determined for each strain level
G/G max
 curves were determined. The normalized curves (G/G
 . Figure 1 presents the basis for determination of secant max
 ) are suitable for presentation of the phenomenon of moduli decrease with increase of shear strain (  ). A plot of G/G max
is known as normalized modulus reduction curve and this refers to very important nonlinearity in the soil behavior under dynamic loads.
Damping of soil was defined by damping ratio
D that represents a perc entage of the critical damping. It was determined by using the relation given in Fig.2.
Dividing each value of the modulus G by G max corresponding to the least shear strain that was obtained during the corresponding test, obtained were the normalized shear moduli
G/G max
for each strain level. Based on the normalized shear moduli values, the functional
G/G max
 curves were determined. The normalized curves (G/G max
 ) are suitable for presentation of the phenomenon of moduli decrease with increase of shear strain (  ). A plot of G/G max
is known as normalized modulus reduction curve and this refers to very important nonlinearity in the soil behavior under dynamic loads.
Damping of soil was defined by damping ratio
D that represents a perc entage of the critical damping. It was determined by using the relation given in Fig.2.
Shear stress versus shear strain relationships for each selected soil material was derived from performed tests. Results for  relationships at seven different levels of shear strain are presented in Figures 6, 7 and 8.
Shear stress versus shear strain relationships for each selected soil material was derived from performed tests. Results for  relationships at seven different levels of shear strain are presented in Figures 6, 7 and 8.
Figure 6.
 relationships different strain levels - Mc9 Figure 7.
 relationships different strain levels – Mc5
The dynamic shear moduli are defined as secant shear moduli G that correspond to the extreme points of the hysteretic curves. They are determined for each strain level g . Figure 1 presents the basis
Scientific Journal of Civil Engineering • Volume 3 • Issue 2 • December 2014 6 | P a g e for determination of secant shear moduli G. By using the obtained values for the shear moduli
G for each strain level, established were the Gg diagrams of the relationships between the shear moduli and the shear strains.
Laboratory Experiments On Soil Dynamic Characteristics Of Npp Site
Dividing each value of the modulus G by G max corresponding to the least shear strain that was obtained during the corresponding test, obtained were the normalized shear moduli G/G max
for each strain level. Based on the normalized shear moduli values, the functional G/G max
g curves were determined. The normalized curves (G/G max
g ) are suitable for presentation of the phenomenon of moduli decrease with increase of shear strain ( g ).
A plot of G/G Scientific Journal of Civil Engineering • Volume 3 • Issue 2 • December 2014 6 | P a g e reduction curve and this refers to very important nonlinearity in the soil behavior under dynamic loads.
max
is known as normalized modulus
Damping of soil was defined by damping ratio D that represents a percentage of the critical damping. It was determined by using the relation given in Fig.2.
Shear stress versus shear strain relationships hysteresis curves at higher levels of strain are clearly observed. These relationships were for each selected soil material was derived from performed tests. Results for t-g relationships at seven different levels of shear strain are presented in Figures 6, 7 and 8.
1.0
Figure 7.
 relationships different strain levels – Mc4
Mc4 basis for definition of normalized G/G damping D curves for different level of shear strain, Figs 7, 8, 10, 11, 13, 14 clearly observed. These relationships were basis for definition of normalized G/G
Bh.No:
Mc 9
8, 10, 11, 13, 14
Exp.1
Exp.2
max
and damping
D curves for different level of shear strain, Figs 7,
1.0
max
and
Bh.No:
Mc 5
Depth: 15.0-15.6 m
Exp.1
Exp.2
Exp.3
30
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0.8
0.8
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014
0.6
0.6
0.4
0.4
0.2
0.2
0.0
1.0E-003 1.0E-002 1.0E-001
Shear strain  (%)
1.0E+000 1.0E+001
0.0
1.0E-003 1.0E-002 1.0E-001
Shear strain  (%)
1.0E+000
Figure 9.
G/Gmax curves versus  - Mc9 Figure 10.
G/Gmax curves versus  - Mc5
1.2
1.0
0.8
Bh.No: Mc 4
Depth: 25.0-26.0 m
Exp.1
Exp.2
Exp.3
Exp4.
0.6
0.4
0.2
1.0E+001
0.0
1.0E-003 1.0E-002 1.0E-001
Shear strain  (%)
1.0E+000 1.0E+001
Figure 11.
G/Gmax curves versus  - Mc4
It can be seen from Figures 9 and 10 that reduction G/Gmax curves slightly differ in the medium range of shear strain values. Soil material from the same depth, for borehole
Mc5, Fig. 10 behave more rigidly compare to results for material from borehole Mc9, Fig.9.
Scientific Journal of Civil Engineering • Volume 3 • Issue 2 • December 2014
7 | P a g e

Figure 7.
 relationships different strain levels – Mc4
1.2
1.0
Bh.No:
Mc 9
Depth: 14.6-15.4 m
Exp.1
Exp.2
Exp.3
0.8
1.2
1.0
0.8
max
Bh.No:
Mc 5
Depth: 15.0-15.6 m
Exp.1
Exp.2
0.6
0.6
 relationships different strain levels – Mc4
0.4
0.4
max
1.0E-002 1.0E-001
Shear strain  (%)
1.0E+000
1.0E-003 1.0E-002 1.0E-001
Shear strain  (%)
1.0E+000 1.0E+001
1.2
1.0
0.8
0.6
0.4
1.0
0.8
0.6
0.4
Bh.No:
Mc 9
Depth: 14.6-15.4 m
Exp.1
Exp.2
Exp.3
Figure 9.
1.2
1.0
G/Gmax curves versus
Depth: 14.6-15.4 m
Exp.1
Exp.2
Exp.3

1.0
- Mc9
1.2
1.0
Bh.No:
Mc 5
Depth: 15.0-15.6 m
Exp.1
Figure 10.
G/Gmax curves versus
Bh.No:
Mc 4
Depth: 25.0-26.0 m
Exp.1
Exp.2
Depth: 15.0-15.6 m
Exp.1
 - Mc5
0.8
Exp.2
Exp.3
Exp4.
Exp.2
0.8
0.8
0.6
0.6
0.6
0.4
0.4
0.4
0.2
0.2
0.2
0.2
0.2
0.0
1.0E-003 1.0E-002 1.0E-001
Shear strain  (%)
1.0E+000 1.0E+001
0.0
1.0E-003
Figure 9.
1.0E-002
0.0
Shear strain  (%)
G/Gmax curves versus

Shear strain 
- Mc9
G/Gmax curves versus
1.2

0.0
0.0
1.0E-002
1.0E+001
 (%)
G/Gmax curves versus
- Mc9
Figure 11.
 - Mc5
G/Gmax curves versus  - Mc4
G/Gmax curves versus  - Mc5
1.0E+001
1.0
0.8
0.6
1.2
1.0
Bh.No:
Mc 4
Depth: 25.0-26.0 m
Exp.1
Exp.2
Exp.3
Exp4.
Bh.No:
Mc 4
Depth: 25.0-26.0 m
Exp.1
Exp.2
Exp.3
Exp4.
P a g e
0.4
0.6
0.2
0.4
0.0
1.0E-003 1.0E-002 1.0E-001
Shear strain 
0.2
1.0E+000 1.0E+001
Figure 11.
G/Gmax curves versus  - Mc4
1.0E-003
1.0E-002 1.0E-001
Shear strain  (%)
1.0E+000
G/Gmax curves versus g - Mc4
G/Gmax curves versus 
1.0E+001
- Mc4
P a g e

V. Shesof, V. Gadza, I. Zafirova, J. Bojadzieva, K. Edip
It can be seen from Figures 9 and 10 that reduction G/Gmax curves slightly differ in the medium range of shear strain values. Soil material from the same depth, for borehole Mc5,
Fig. 10 behave more rigidly compare to results for material from borehole Mc9, Fig.9.
35
3 5
30
30
25
25
20
20
15
15
10
5
0
V. Shesof, V. Gadza, I. Zafirova, J. Bojadzieva, K. Edip
10
V. Shesof, V. Gadza, I. Zafirova, J. Bojadzieva, K. Edip
Exp.1
5
Exp.2
Exp.3
35
0
1.0E-002 1.0E-001 1.0E+000 1.0E+001
30
Shear strain  (%)
30
1.0E-002 1.0E-001
Shear strain  (%)
Bh.No:
Mc 5
Depth: 15.0-15.6 m
Exp.1
Exp.2
1.0E+000
Figure 12.
Damping ratios versus  - Mc9 Figure 13.
Damping ratios versus  - Mc5
1.0E+001
35
30
25
20
15
10
5
0
1.0E-003
Figure 12.
1.0E-002 1.0E-001
Shear strain  (%)
Damping ratios versus 
Bh.No:
Mc 9
Depth: 14.6-15.4 m
Exp.1
1.0E+000
Exp.2
Exp.3
1.0E+001
At the same time if we look at the results for damping D, Figures 12 and 13 the material from borehole Mc 9 shows higher damping than material from borehole Mc 11 from medium to large levels of shear strain. This characteristic behavior of soil material under
- Mc9
25
20
10
5
Figure 13.
Damping ratios versus 
Bh.No:
Mc 9
Depth: 14.6-15.4 m
Exp.1
- Mc5
Exp.2
medium to large levels of shearing is one of
20
15
10
5 the crucial step that has to be take into account when site response analysis is performed.
Results from the cyclic simple shear tests were used also to define the influence of the effective stresses upon shear moduli G, Figure
14.
Bh.No:
Mc 5
Depth: 15.0-15.6 m
Exp.1
Exp.2
0
Figure 13.
Damping ratios versus g - Mc5 medium to large levels of shearing is one of
0
15
1.0E-003 1.0E-002 1.0E-001
Shear strain  (%)
1.0E+000 1.0E+001 1.0E-003 1.0E-002 1.0E-001
Shear strain  (%)
1.0E+000 1.0E+001
10
5
0
Figure 12.
At the same time if we look at the results for damping D, Figures 12 and 13 the material
1.0E-003
Damping ratios versus
1.0E-002 1.0E-001

Depth: 15.0-15.6 m than material from borehole Mc 11 from
- Mc9 Figure 13.
Damping ratios versus the crucial step that has to be take into account when site response analysis is performed.
 - Mc5
Results from the cyclic simple shear tests were used also to define the influence of the effective stresses upon shear moduli G, Figure
14. At the same time if we look at the results for damping D, Figures 12 and 13 the material the crucial step that has to be take into account when site response analysis is damping D, Figures 12 and 13 the material from borehole Mc 9 shows higher damping than Results from the cyclic simple shear tests material from borehole Mc 11 from medium to effective stresses upon shear moduli G, Figure
14. levels of shearing is one of the crucial step that has to be take into account when site response analysis is performed.
Results from the cyclic simple shear tests were used also to define the influence of the effective stresses upon shear moduli G, Figure 14.
Presented results in Figure 14 show increase
(increasing of overburden pressure). The results
32
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Figure 14.
Variation of Initial shear moduli G with depth
Presented results in Figure 14 show increase
(increasing of overburden pressure). The stiffness along the depth of the foundation soil.
Results from the tests show stable   hysteretic behavior for medium to large strain level while for small strains less than 10 it was observed relatively large scattering of data. Therefore some of the results for Gmax and damping D for low level of strains were depth not take into account and not presented in figures. Even the measurement system of the apparatus was improved with two LVDT (direct and indirect measurements) it was not enough to get stable data for strain level less than 10
Regarding the estimation of G max designer to upgrade the given G/G
-3 .
(low shear strain moduli) strong suggestion was given to
curves with the results from in-situ geophysical survey. to get stable data for strain level less than 10
Regarding the estimation of G max
-3 .
(low shear strain moduli) strong suggestion was given to designer to upgrade the given G/G survey. max
curves with the results from in-situ geophysical
Figure 14.
Variation of Initial shear moduli G with depth
Presented results in Figure 14 show increase in values of initial G moduli with the depth
(increasing of overburden pressure). The results can be used for modeling of soil stiffness along the depth of the foundation soil.
Results from the tests show stable   hysteretic behavior for medium to large strain level while for small strains less than 10 -3 it was observed relatively large scattering of data. Therefore some of the results for Gmax and damping D for low level of strains were
Results from the tests show stable  
Scientific Journal of Civil Engineering • Volume 3 • Issue 2 • December 2014 8 | data. Therefore some of the results for Gmax not take into account and not presented in figures. Even the measurement system of the
P a g e apparatus was improved with two LVDT (direct and indirect measurements) it was not enough
P a g e to get stable data for strain level less than 10
Regarding the estimation of G max
-3 .
(low shear strain moduli) strong suggestion was given to designer to upgrade the given G/G survey. max
curves with the results from in-situ geophysical
Scientific Journal of Civil Engineering • Volume 3 • Issue 2 • December 2014 8 | P a g e

Laboratory Experiments On Soil Dynamic Characteristics Of Npp Site
Results from the tests show stable t g hysteretic behavior for medium to large strain level while for small strains less than 10 -3 it was observed relatively large scattering of data. Therefore some of the results for Gmax and damping D for low level of strains were not take into account and not presented in figures. Even the measurement system of the apparatus was improved with two
LVDT (direct and indirect measurements) it was not enough to get stable data for strain level less than 10 -3 . Regarding the estimation of G max
(low shear strain moduli) strong suggestion was given to designer to upgrade the given G/G max
curves with the results from in-situ geophysical survey.
For layered soil, the strain compatible shear modulus values and damping values for each layer are the bases for the derivation of the mathematical model of the layered soil and necessary input for realistic site response analysis of the NPP site. Strain dependent G/Gmax and D curves for representative soil layes at one NPP site are developed by series of strain controlled cyclic simple shear tests. Tests were performed taking into account the most important factors which influenced the shear moduli and damping of soils: shear strain amplitude, mean principal effective stress and PI. Cyclic simple shear apparatus used in this study enables to produce reliable and very stable modulus reduction and damping curve for wide range of strain
(medium to large level of shear strain). Results from performed experimental investigations used in site response analysis provide better understanding of seismic performance of foundation soil and reliable estimation of ground motion. Experimentally defined relationships for shear modulus and damping provide solid base for modeling of soil structure interaction.
Improvement concerning the estimation of small strain shear moduli can be done by combining the in-situ G curves. max
with laboratory derived G/G max
[1] Manual for operation of the Cyclic Simple
Shear Apparatus – Dames and Moore ,
London UK, April 1981,
[2] Tatsuoka,F., Muramatsu,M. and Sasaki,T.
(1982), “Cyclic undrained stress-strain behavior of dense sands by torsional simple shear test”, Soils and Foundations,
Vol.22, No.2, pp.55-70.
[3] Tatsuoka,F., Ochi,K., Fujii,S. and
Okamoto,M. (1986), “Cyclic undrained triaxial and torsional shear strength of sands for different sample preparation methods”, Soils and Foundations, Vol.26,
No.3, pp.23-41.
[4] Vucetic M., Dobry R., Effect of soil plasticity on cyclic response – Journal of
Geotechnical Engineering, Vol.117 No.1
January 1991
[5] Ishihara, K. Soil Behavior in Earthquake
Geotechnics, Oxford Science Foundation,
1996
[6] Tatsuoka, F. Selected papers on deformation characteristics of geomaterials, 1988-2000, Geotechnical
Laboratory, Department of Civil engineering, University of Tokyo, Tokyo ,
Japan
[7] Massarsch, K. R. 2004. Deformation properties of fine-grained soils from seismic tests. Keynote lecture,
International Conference on Site
Characterization, ISC’2, 19 – 22 Sept.
2004, Porto, 14 p.
[8] Jianfeng Zhang; Ronald D. Andrus; and C. Hsein Juang, Normalized Shear
Modulus and Material Damping Ratio
Relationships, Journal of Geotechnical and Geoenvironmental Engineering, Vol.
131, No. 4, April 1, 2005. ASCE, ISSN
1090-0241/2005/4-453–464/
[9] US NRC Regulatory Guide 1.138 ;
Laboratory Investigations of Soils and
Rocks for Engineering Analysis and
Design of Nuclear Power Plants
[10] IAEA Safety Guide, No. NS-G-3.6;
Geotechnical Aspects of Site Evaluation and Foundations for Nuclear Power
Plants
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 33
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16 | Page
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624.012.4:692.92 678:666.97
A UTHORS
György L. BALÁZS
Ph.D. Professor
Budapest University of Technology and Economics
Faculty of Civil Engineering balazs@vbt.bme.hu
Concrete structures may be subjected to severe loadings, like earthquake, impact or increase of loads otherwise suffer deterioration of material properties. External strengthening methods provide convenient possibilities to increase capacity of concrete structures. Two convenient ways of strengthening are the use externally bonded reinforcements (EBR) or the use of near surface mounted (NSM) reinforcements. Materials of these reinforcements are today mainly FRP (Fibre
Reinforced Polymer) with polymer based adhesive. Present paper intends to review the main characteristics of EBR and NSM strengthening methods including design, detailing and application aspects.
retrofitting, strengthening, FRP,
EBR, NSM
Concrete structures may need strengthening due to the deterioration of material properties
(including excessive cracking or deflection), increase of loads or overloading or even modification of strengthening system.
Engineers are always looking for clever ways of strengthening with the following requirements:
- considerable increase in service life with limited maintenance after strengthening
- easy application even by keeping the live load (at least partly)
- short period of time for strengthening work
- aesthetic solution not disturbing the appearance of the structure
- economic solution at least in relative terms.
Externally bonded reinforcement (EBR) and near surface mounted (NSM) reinforcement are the most convenient strengthening methods today by using fibre reinforce
Scientific Journal of Civil Engimeering, Volume 3, Issue 1, June 2014 19 Page

Laboratory Investigations To Determine The Effect Of Lime Improvement Of Soft Silty Soil
Authors
Bojan SUSINOV
M.Sc. Assistant
University of Ss Cyril and Methodius
Faculty of Civil Engineering Skopje bojan_gt@hotmail.com
Josif JOSIFOVSKI
Ph.D. Assistant Professor
University of Ss Cyril and Methodius
Faculty of Civil Engineering Skopje jjosifovski@gf.ukim.edu.mk
Sead ABAZI
M.Sc. Assistant
University of Ss Cyril and Methodius
Faculty of Civil Engineering Skopje sead_aba@hotmail.com
Lime stabilization as a method for soil improvement is benefitial for number of important engineering properties, such as: strength, resistance to fracture, resilient properties and reduced swelling.
This paper briefly describes the application of quicklime to stabilize soft soil. Many laboratory tests have been carried out on silty soil to determine the improvements for lime in varying percentages.
The laboratory investigations for different lime contents preparing with optimum moisture content and sample curing under controlled conditions were carried out. The investigation of both treated and untreated soil are focused mainly on the strength parameters, changes in the plastic properties, compaction requirements, California
Bearing Ratio and compressibility characteristics of the lime-soil mixtures. Significant improvement and stabilization of the silty soft soil has been observed for 4 percent of lime admixtures.
Stabilization, soil improvement, quicklime, silty soft soil.
Soils vary widely in engineering properties and often local soils are not adequate to meet the support requirements of a construction project
(Slag Cement Association, 2005). Constructions over soft soil are one of the most frequent problems in many parts of the world (Emilliani and Ismail, 2010). Thus, soil stabilization has become the major issue in geotechnical and structural engineering. The literature review has shown that many researchers analyze the effectiveness of using different materials as soil stabilizers. Several materials can be used as soil stabilizing agents e.g. lime, cement, fly ash and their mixtures.
Soils can be improved by adding lime to the soil, mixing thoroughly with a measured amount of water, and densely compacting the mixture.
Lime stabilization is particularly important in road construction for modification of sub-grade soils, sub-base materials and base materials. It may be used for shorter-term soil modification e.g. to provide a working platform at a construction site.
Through stabilization, it has been found that not only mechanical properties were improved, compressive strengths and bearing capacity were increased, but also durable pavement was created. Lime improves
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014 35
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Page

B. Susinov, J. Josifovski, S. Abazi the strength of soils by three mechanisms: hydration, flocculation and cementation. The first and second mechanisms occur almost immediately upon introducing the lime, while the third is a prolonged effect. The main objective of this paper is to evaluate the effects of the lime on silty soft soil.
Disturbed soil samples were obtained from excavation pit at 2.0m depth. The soil samples were subjected to classification laboratory testing.
A summary of physical properties of the tested soil is presented in table 1. The grain size distribution curve indicated that the soil is composed of
62% silt, 30% fine sand and 8% clay. Based on the Unified Soil Classification System (USCS) the soil is low plasticity clay (CL). The soil also classified as A-6 (9) soil in accordance with the
AASHTO classification system.
Table 1.
Physical properties of untreated soil
Characteristics
Colour
Natural water content [%]
Field dry unit weight [kN/m 3 ]
Specific gravity
Passing No. 200 sieve [%]
Clay content (d<2µm) [%]
Plasticity Index [%]
OMC [%]
MDD [kN/m 3 ]
CBR [%]
Activity
USCS
AASHTO Class. System (GI)
Values, description dark brown
29.2
15.1
2.79
73
8
12.2
15
18.3
2.67
1.52
CL
A-6 (9)
The form of the lime could be either quicklime
(CaO), or hydrated lime (Ca(OH)2). Quicklime hydrates with the soil moisture to become hydrated lime and therefore acts as a better drying agent before providing the calcium to react with the silica and alumina in the soil. Table 2 presents the chemical compositions of quicklime used in this research. The optained LOI is 27.26 %.
Table 2.
Chemical compositions of quicklime compound
SiO
2
Al
2
O
3
Fe
2
O
3 quicklime
[%]
0.10
0.61
1.45
compound
SO
3
MgO
CaO quicklime
[%]
0.11
0.97
69.15
The laboratory testing procedures include determining optimum lime requirements and moisture content, preparing samples, and curing. Curing is important for
36
|
Page chemically stabilized soils because lime-soil reactions are time and temperature dependent.
All laboratory tests were carried out according to Macedonian standards but also some recommendations of ASTM standards had been used.
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Laboratory Investigations To Determine The Effect Of Lime Improvement Of Soft Silty Soil
The tests were carried out on specimens of soillime mixtures with different percentages of lime (2,
4, 6, 8, 10 %) added with respect to the dry weight of soil. To determine the optimum lime content, the
Eades and Grim pH test was performed. Change of soil plasticity under the effect of lime in various percentages was measured after 1 hour of primary mixing, while changes in the Proctor rates was calculated after a mellowing period of 24 hours. The
California bearing ratio (CBR) was performed after a sample curing for 7 days at 40°C and then soaked for
4 days. The specimens for oedometer test cured 7 days at 23°C. The cylindrical samples, of dimensions
50x110 mm, were prepared and compacted with optimum moisture content to obtain the Unconfined
Compression Strength (UCS) after curing for 3 days at 50°C, 7 days at 45°C and 28 days at 23°C.
The Eades and Grim test is used to approximately determine optimum lime content required to satisfy immediate lime-soil reactions and still provide significant residual calcium and high system pH
(about 12.4 at 25°C) (Dallas and Yusuf, 2001).
This is necessary to provide proper conditions for long-term pozzolanic reaction that is responsible for strength and stiffness development.
The values in table 3 indicate that there is significant increase in pH when 2% lime is added, but the increase diminishes as lime is further added to the soil.
Table 3.
Physical properties of lime-soil mixtures
Lime [%] pH [/]
LL [%]
PL [%]
PI [%]
0
29.7
17.5
12.2
2
29.3
23.3
6.0
4
30.4
22.9
7.5
6
29.7
23.0
6.7
8
8.21
12.26
12.35
12.41
12.42
30.0
23.3
6.7
All pH values of the different mixtures are in correspondence with the recommended values.
Liquid limit (LL), plastic limit (PL) and plasticity index
(PI) data obtained on the five mixtures are presented in table 3. The LL seems unaffected by the lime content, the PL increases and the PI decreases when 2% lime is added to soil. The further addition of lime does not change the plasticity.
Standard Proctor compaction test was conducted on the five mixtures. The addition of lime to the soil caused reduction in the maximum dry density (MDD) and increase in the optimum moisture content. The typical compaction curves of different soil-lime mixtures are presented in
Figure 1.
Figure 1.
Moisture – dry density relationship
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B. Susinov, J. Josifovski, S. Abazi
Although not investigated, is expected that the time of curing can contribute by increase in the optimum moisture content.
The California bearing ratio (CBR) of a compacted soil is determined by comparing the penetration load of the tested soil to that of a standard high quality crushed stone rock. The results are used to evaluate the relative quality and strength of a soil. The results presented in figure 2 and 3, indicate that as lime content is increased there is an increase in the CBR value. If 4 % lime is added to soil and cure 7 days, the CBR increase is significant. Following the recommendation for optimal lime content (AASHTO Classification), the strength and deformability parameters were investigated for 4, 6 and 8 % of lime.
a) b)
Figure 2.
а) Load versus penetration data obtained from a CBR test and b) CBR values for penetration of 2.54 mm for different lime content
The apparatus used was standard one-dimensional oedometer. Soil-lime mixtures were blended and moistened, and then allowed to sit for 3 hours before compaction to simulate the delay that typically occurs in the field. All remolded specimens were left 7 days to cure before testing. This process allowed the water to be distributed uniformly within the sample without any loss of moisture. Figure 3 shows values of the compressibility modulus Mv
(Eoed) obtained on all four mixtures.
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Figure 3.
Modulus of compressibility versus lime content
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Laboratory Investigations To Determine The Effect Of Lime Improvement Of Soft Silty Soil
There is an increase in Eoed as lime content increases to 4%. Unexpected decrease in Eoed from 4 to 6 % is registered, especially in light of the values obtained for 8 % of lime content.
To evaluate the effect of lime content, UCS samples are prepared for un stabilized and stabilized soil at three lime contents (4, 6 and
8%).
Lime is thoroughly mixed with the dry soil at
OMC and placed in plastic zip-lock bags for 1-24 hours. After the mellowing period specimens were compacted in accordance with standard proctor test.
Specimens were divided into three sets, each consist of two with same lime content. Every set of samples was cured as explained before to evaluate the effect of curing time.
Additionally, an identical set of replicate samples is tested with capillary soak to evaluate the effect of moisture conditioning.
The effect of lime content is presented in Figure 4.
a) b)
Figure 4.
UCS versus lime content а) cured at 45°C and b) cured at 23°C
By comparing UCS of soil-lime specimens cured at 45°C, it is seen that unconfined compression strength increases as the lime content increases to 4%. As the lime content increases to 6 or 8%, the UCS decreases. The specimens cured for 7 days with 4% lime content showed the highest
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B. Susinov, J. Josifovski, S. Abazi values of UCS (1259 kPa). It is clear that only
4% lime is sufficient amount to double the UCS of the soil.
When the specimens cured for 28 an 56 days at 23°C the unconfined compression strength increases as the lime content increases to 6%.
As the lime content increases to 8%, the UCS decreases.
The effect of curing time is presented in Figure 5.
a) b)
Figure 5.
UCS versus curing time а) 3 and 7 days and b) 28 and 56 days
It is clear that the UCS increases by curing time for same curing conditions. Specimens cured at 45°C for 7 days show 65% higher UCS than the specimens cured for 3 days at the same temperature. Also, the specimens cured at
23°C for 56 days show 71% higher UCS than the specimens cured for 28 days at the same temperature.
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After curing period, some specimens were exposed to soaking for 24 hours to evaluate the effect of moisture conditioning on UCS. Figure
6 presents the results from UCS performed on the specimens after 7 days at 45°C curing time and 24 hours capillary soaking.
Scientific Jounal of Civil Engineering, Volume 3, Issue 2, December 2014

Laboratory Investigations To Determine The Effect Of Lime Improvement Of Soft Silty Soil
Figure 6.
Effect of moisture condition on UCS for different lime percentages
It is seen that the unconfined compression strength decreases with increasing moisture after soaking.
The untreated soaked specimens don’t show any compressive strength.
The mixture of lime and silty soil material, has significantly improved the mechanical properties.
Even at 2 % of lime, a reduction of moisture content and the plasticity index is around 40% and 45%, respectively. The addition of quicklime increases the optimum moisture while decreasing the maximum dry density due to lower specific weight. Thus, the quicklime can be applied in soils with high moisture content resulting with more efficient compaction.
It would be interesting to investigate the long-term performance of lime stabilized soil.
[1] Beeghly J.H.: „Recent experiences with lime – fly ash stabilization of pavement subgrade soils, base, and recycled asphalt“, International ash utilization symposium, 2003, pp.4
[2] Dallas N.L., Yusuf F.A.M.S.: „Example problem for application of the national lime association
MDTP to ascertain engineering properties of lime-treated subgrades“, Technical document.
2001, pp.1-24
[3] Emilliani A.G., Ismail D. S. A.: „Geotechnical
Properties of Fly Ash and its Application on Soft
Soil Stabilization“, 2010, pp.1.
Addition of lime does have effect on the CBR values. The CBR has improved up to 16 times when
8% of lime is added to soil and cured 7 days and even better results are expected for longer period of time. Lime also can improve the compressibility characteristics of the soil. 4% of lime can increase the compressibility modulus up to six times.
[4] National lime association (www.lime.org):
„Mixture design and testing procedures for lime stabilized soil“, Technical brief, 2006
[5] Сусинов Б.: „Стабилизација и модификација на слабоносиви почви со додаток на вар“,
Skopje, 2013.
Unconfined compression strength increases depending on the lime content and the duration of the specimen curing, but decreases with increase in the moisture after soaking. The largest increase is observed in specimen with 4% lime where for 7 days of curing at 45°C, the stabilized soil shows 2 times greater strength compared to the unstabilized soil. In this context, the same soil-lime specimens soaked for 24 hours show greater UCS.
[6] Susinov B.: „ Lime stabilization of silty soft soil “,
5thiYGEC, Paris, 2013.
[7] Slag Cement Association : „Use of Slag Cement in Soil Cement“,Bulletin No. 25, 2005, pp.1.
A general conclusion would be that the most suitable amount of lime to improve the strength characteristics of this soil would be 4%. On the basis of pH, LL, PL and PI, it can be concluded that 2% lime is optimal.
[8] Taha R. et all: „Use of cement by-pass dust in soil stabilization“, Engineerig journal of the University of Qatar, vol 14, 2001, pp.61-76.
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Methods For Predicting Impact of Ground Vibrations Induced By Pile Driving on The Old Masonry Wall Buildings and Their Monitoring
Authors
Zvonko TOMANOVIC
Ph.D. Professor
University of Monte Negro
Faculty of Civil Engineering Podgorica zvonko@ac.me
Pile driving in urban areas raises the question of adverse effects of induced soil vibrations on the adjacent buildings. Design and dimensioning of buildings and structures takes into account the soil oscillations generated by earthquakes as natural inducers. However, ground vibrations resulting from pile driving can also cause partial or total collapse of the adjacent buildings.
Two basic mechanisms of instability are loss of load-bearing capacity under a foundation or damage of structural system. Prediction of damage of structural system caused by artificial vibrations through application of theoretical models is a demanding and unsafe procedure.
Measurement of ground vibrations (velocity and acceleration) arising from pile driving cannot be directly correlated with the load-bearing capacity of structural elements and overall structural integrity. Namely, the exact method of control of the structural load-bearing capacity under the impact of artificial vibrations is not defined by any standard or technical regulation.
Certain standards (such as EC3, DIN 4150,
ÖNORM S 9020) define maximum allowable soil velocity of vibration for particular type of adjacent buildings. Although they can be helpful in the risk assessment of possible damages, such standards cannot guarantee that some damage will not occur even if the soil vibration velocities and its frequencies are within the allowable limits. Therefore, during the pile driving in urban areas it is necessary to perform monitoring of the buildings with high risk damage. This paper contains an overview of methods for assessment of threat to buildings exposed to the vibrations induced by pile driving and also monitoring methodology and results of vibration measurements and monitoring of masonry wall building originally built in 1930 during the pile driving in surrounding location.
ground vibrations, piles, wave propagation
Driven piles are used in cases where ground conditions do not allow a structure to be
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supported directly on the underlying soil (shallow foundations). Such piles represent a good technical solution from the aspects of load-bearing capacity and cost-effectiveness. However, in terms of vibrations, noise and environmental pollution, application of driven piles has some serious shortcomings, especially when used in urban areas. In principle, the hammer blow noise and environmental pollution that is mainly manifested through diesel engines exhaust emissions are far less an issue compared to the ground vibrations induced by pile driving.
For the engineering purposes, the impact of vibrations on the surrounding buildings is quantified by a maximum (peak) soil particle velocity induced by pile driving. This velocity is used to assess the potential structure damage risk. Another important indicator of potential damage, in addition to the particle velocity, is the seismic wave frequency and number of cycles to which the structure will be exposed.
After assessment of the level of threat to particular buildings which are exposed to the effects of ground vibrations caused by pile driving, it can be resorted to the ultimate of prohibiting the use of this technology if assessed that it can cause serious damages or collapse of the adjacent buildings. If the intensity of vibrations is such that can impact the safety of the structure, including cosmetic damages as well, it is necessary to perform monitoring of the structure that is threatened by vibrations during the pole driving process. Corrective measures can be applied where as appropriate either through construction technology or by strengthening the structure which is at threat.
Zvonko Tomanovic
When driven piles are planned to be applied for foundation of a building in urban area, it is necessary to identify any nearby building which can be endangered by adverse effects resulting as a consequence of these activities. Particularly strong vibrations are generated when a free-fall diesel hammer of heavy weight, using relatively small number of blows is used as a pile driver
(Fellenius, 2006). A ground vibration of certain amplitude and frequency that occurs at pile driving is actually caused by the hammer impact on the pile cap (Figure 1). This is followed by propagation of mechanical waves through the pile body, along the pile shaft and finally at the pile toe where this vibration is further transmitted to the soil particles which start to vibrate and generate seismic waves.
These waves are then transmitted through the soil and the ground water, causing dynamic response of the structures of the buildings which are in the vicinity of the pile driving area, through a complex interaction of soil and the structure of the building, producing specific vibration amplification. The soil in which the pile is driven has a significant impact on the intensity of vibrations and propagation of elastic waves (Massarsch et al., 2008).
During the propagation of these seismic wave
- vibrations, some minor ground deformations can occur, which can be measured by specific techniques (White et al., 2003). However these topics will not be further considered in this paper as they do not have a direct impact on the prediction of damage to the nearby structures.
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Figure 1.
Schematic illustration of transfer of vibrations during pile driving
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Methods For Predicting Impact of Ground Vibrations Induced By Pile Driving on The Old Masonry Wall Buildings and Their Monitoring
The pile driving procedure itself induces ground vibrations and it also causes an adverse environmental impact through the increased noise levels and emission of exhaust gases from diesel engine operations (the environmental aspect is not the focus of this paper). Impact of vibrations on the nearby buildings can initiate several adverse processes that may cause partial or total collapse of the buildings. The impact of vibrations can reflect on a) the load bearing capacity of the foundation soil and b) the structure of the building. The impact on the foundation soil can be manifested through the compaction of soil particles that may cause settlement of structures, occurrence of liquefaction and total loss of the load-bearing capacity of the foundations. The impact of vibrations on the structure of building depends primarily on the structural system, building material and type of foundation. Modern buildings with reinforcedconcrete and steel frames are much more resistant to vibrations than the old structures with brick walls or stone walls.
It is worth mentioning the impact of vibrations on the non-structural components of the building, such as wall mortar, non-structural walls, parapets, doors and windows etc. Damages to the non-structural components of the building (the so called cosmetic damages) will not pose risk to the stability of the building; however they can fully endanger the functionality of the building. The issue of functionality is particularly important in case of the industrial buildings in which sensitive equipment is used for manufacturing process or in case of the underground systems such as waterline, gas line etc.
An exact theory which describes a general case of the impact of vibrations generated by pile driving on the adjacent buildings has not been developed up to date. A certain progress can be achieved by application of numerical models based on the finite element method or boundary element method; however these consume a lot of designing time with the calculation results being uncertain. On the other side, based on the experience acquired from the pile driving a significant number of correlations has been established, and such correlations are incorporated in large number of standards applicable to this area. These standards can be helpful when assessing whether the vibrations of certain soil particle velocity and vibration frequency caused by the pile driving pose threat to the adjacent buildings. However, there is no guarantee that some minor (cosmetic) damages of the building structure, adverse foundation settlement or disruption of the building functionality will not occur during the pile driving.
In addition to the factors above, eventual damage of the building can be affected by the number of cycles, meaning the number of piles that will be driven during the construction works. Settlement of structures increases with the higher number of cycles, particularly in the pure sand environment. One of the first laboratory studies of the seismically induced settlement of sands was performed by Seed and Silver (1972). However, the number of cycles or the duration of the construction works during pile driving is incorporated in standards mainly for the sole purpose of assessment of human response to vibrations and increased noise during the construction works (BS 5228) and not for the assessment of the vulnerability of structures of adjacent buildings to vibrations.
Mechanical oscillations of the building foundations can be registered and analysed based on the amplitude and frequency properties of the seismic waves in the conditions of a real mechanical excitation of soil that is produced by the causative agent of oscillations. A standard approach to measurement in such cases involves the use of proper seismic measuring instruments and sensor arrays (geophones) with high mechanical sensitivity and proper frequency response range. However, assessment of the effects of vibrations on the adjacent buildings is far more complex issue, due to the complex process of the wave propagation through the soil, foundation-soil interaction and the building structure itself. The impact of vibrations on the damage to the buildings can be assessed based on statistical data. Basically, this approach has been used in developing of the majority of standards which govern this area.
At pile-soil interaction, after the hammer impact on the pile cap, in the ground occurs a transmission of different seismic waves and wave fronts. A spherical front of P and S waves is induced at the pile toe. A conical front of the S wave is formed on the pile shell. When the wave front reaches the terrain surface, the vibration energy is transformed into the surface R wave,
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Zvonko Tomanovic
Figure 2 (Athanasopolus et al., 2000; Deckner et al., 2012).The wave propagation through the soil is followed by their attenuation as a result of geometric and material damping. Geometric damping is caused by a change in the soil volume (expansion) and material damping is a consequence of the internal friction and hysteresis.
Figure 2.
Schematic representation of different seismic wave types generated at pile driving
Predicting the level of vibrations during construction of thebuilding founded on piles, in the circumstances when there are nearby buildings that may be threatened during the construction works, has a significant impact on technical solutions and cost price. Conservative approach almost always overrates the impact of vibrations and increases the price of construction works, yet shortens the time that would be spent on analysis of the maximum level of vibrations. On the other side, if the level of vibrations is underrated, damages to adjacent buildings may occur and the human response can lead to suspension of works, which can result in the increase of costs and delay of works. For this reason it is a common practice nowadays to base the level of vibrations on the in-situ measurements during actual construction of piles at the particular site. By this it is possible to include all the effects which cannot be incorporated even in the most complex numerical models without minor or major approximations with significantly longer period of time necessary to obtain the solution.
In the recent decades, based on the experience acquired from measurement of propagation of vibrations through soil, several mathematical correlations between the hammer stroke energy, distance from the hammer impact point and measurement point and particle velocity has been developed. One of the first empirical methods proposed by Attewell & Farmer (1973),is the following: where are:
- predicted maximum peak particle velocity at distance operation,
(1) from the piling
- energy of hummer blow in one cycle,
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Methods For Predicting Impact of Ground Vibrations Induced By Pile Driving on The Old Masonry Wall Buildings and Their Monitoring expressed in Joules (Nm) and
– parameters relating to the soil characteristics, piling technique and ground profile.
Parameter typically varies in the range from
0.5 (soft soil) to 1.5 (stiff soil), while value of parameter varies from 0.5 to 1 in different analysis and standards. EUROCODE 3 (EC 3 part 5, EN1993, hereinafter referred to as EC3) recommends a value of 1 for parameter
Diagram of predicted ground vibrations during dynamic piling according to EC3 is illustrated in
Figure 3, (White et al., 2002).
Figure 3.
Diagram of predicted ground vibrations caused by dynamic pile driving according to EC 3
The intensity of an artificially generated quake is sometimes assessed by seismic scales, for instance by MCS (Merkalli-Cancani-Sieberg).
Although there are similarities between the artificially caused seismic waves and those occurring due to natural earthquakes, there are significant differences, particularly in quake duration of dominant period of oscillations, seismic waves spectral content etc. Because of these differences, the seismic intensity scale cannot be applied directly for assessment of the intensity of ground shaking in case of artificially caused vibrations such as blasting, pile driving, dynamic soil compaction etc.
Russian scale for assessment of seismic intensity of artificially caused quakes – vibrations has been developed at the Institute of Physics of the Earth, Russian Academy of Sciences.
The allowable level of seismic intensity in case of artificially caused quakes (blasting, technical vibrations etc.) for structures of different earthquake-resistance is related to the velocity of particle oscillations. The scale has twelve seismic degrees and the buildings and structures are classified in four classes:
Class I - buildings of particular importance, buildings of specific interest for the state and architectural monuments. Artificial induction of quakes near these structures is possible only in extraordinary cases.
Class II - industrial buildings of particular importance - pipelines, large-capacity production halls, mine head-frames, water towers and similar structures with design life of over 20 to 30 years; residential buildings with large number of residents, houses of culture, cinemas and similar buildings.
Class III - industrial buildings and office buildings of relatively small dimensions, with no more than three storeys;
Class IV - structures and industrial which accommodate expensive equipment, yet damage
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Zvonko Tomanovic to such buildings would not endanger human life and health;
Swedish standard SS 02542 11 (SS1999) is based on a 30-year experience in driving different types of piles, sheet pile, sand compaction in wide range of Swedish soils. The standard specifies vibration levels which are acceptable for different types of buildings, but it does not consider noise, environmental protection and impact of vibrations on the operations of different industrial machines.
A hazard level is defined by a vertical component of the wave velocity for comparable foundation types; however no consideration is given to liquefaction, soil compaction and similar effects.
A horizontal component of velocity is taken into consideration only in a limited number of situations. Vibrations should be measured at the foundation level in the closest location from the piling site (Massarasch et al., 2008).
Allowable ground oscillation velocity is calculated based on the following Equation:
(2) where:
- Vertical component of vibration velocity
(from 9 to 15mm/s),
- Building factor (from 0.5 to 1.7),
- Material factor (from 0.65 to 1.2),and
- Foundation factor (from 0.6 to 1.0).
German standard DIN 4150/3 defines three categories of structures and specifies respective allowable oscillation velocities which change in correlation with the oscillation frequency of the excited material particles.
This standard has established a correlation between the oscillation velocity and oscillation frequency of the artificially excited particles of soil. Standard DIN 4150/3 has identified three types of structures and for such structures has specified allowable oscillation velocities for the ground material particles, in particular:
1. Buildings used for commercial purposes, industrial buildings and similar reinforcedconcrete structures.
2. Residential buildings and buildings of similar purpose or structure.
3. Buildings that, because to their particular sensitivity to vibration, cannot be classified under lines 1 and 2 and the preservation of which is of intrinsic value - damaged buildings or listed buildings - delicate structures (historical monuments etc.).
Allowable limit values for oscillation velocity according to DIN are also illustrated in Figure 4.
Maximum allowable limit values for soil particles velocity at the ground level are specified for the defined frequency ranges.
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Figure 4.
Allowable limit values for oscillation velocity according to DIN 4150/3
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Methods For Predicting Impact of Ground Vibrations Induced By Pile Driving on The Old Masonry Wall Buildings and Their Monitoring
Eurocode 3 defines a level of maximum velocity in terms of human comfort and adverse effects on buildings, while the level of maximum velocity in terms of building damage prevention is limited to the values shown in Table 1 (Bolton, et al., 2002).
Table 1.
Maximum acceptable transient vibration according to EUROCOD 3
Type of structure
Buried services
Heavy industrial
Light commercial
Residential
Architectural merit
Maximum acceptable transient vibrations [mm/s]
25
15
10
5
2
Austrian standard ÖNORM S 9020 distinguishes between four building classes and specifies respective allowable ground oscillation velocities. The classes are defined with respect to the structural system, building material, and building and terrain conditions, Table 2. Limit value for soil particle velocity ranges from 5 to
30mm/s (Table 3), similar as defined by EC3.
Table 2.
Building classes according to the Austrian Standard ÖNORM S 9020
Buildung class
I
II
III
IV
Structure
Industrial structures
- Frame structures (with or without core) of steel and/or reinforced concrete
- Wall structures (concrete, pre-fabricated segments)
- Engineering timber structures (e.g. halls)
Residential buildings
- Frame structures (like I)
- Wall structures (like I)
- Structures with concrete slabs, walls made of concrete, brick, masonry with cement or lime mortar
- Timber structures, except half-timbered buildings
Frame structures with less strength than structures of class I and II: Structures with basement slabs of concrete or brick arches and pre-fabricated parts, timber beam or pre-fabricated brick slabs in the upper floors. Brick lined half-timbered buildings.
Listed buildings particularly sensitive to vibrations
Table 3.
Reference values for the allowable peak velocity magnitude vR,max [mm/s] (measured at the foundation) according to the Austrian Standard ÖNORM S 9020.
Building class
I
II
III
IV
Allowable peak velocity magnitudemax vR, max [mm/s]
Infrequent Blasting (weekly)
30
20
10
5
Frequent
Blasting (daily)
24
16
8
4
Long-lasting
Temporary vibrations
12
8
4
2
As mentioned before, the vibrations induced by pile or sheet pile driving in urban areas can cause serious adverse effects (Athanasopoulos 2000,
Rockhill 2003). For this reason, the prediction of the pile driving impact in urban areas is of significant importance to making proper engineering decisions.
Monitoring of potentially threatened buildings during the construction phase also has an important role in the risk assessment during execution of works.
The need for monitoring and risk assessment during pile driving works should be defined in the structural design (foundation). This design, based on geological conditions, adopted construction technology, vicinity of adjacent buildings and use of simplified models and formulas can provide
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Zvonko Tomanovic an answer if an adverse effect of pile driving on the nearby buildings and environment can be expected during the execution of works. One of the equations that can be useful for this phase of analysis is provided by Head & Jardine (1992), and it reads:
(3)
Where parameters have the same meaning as referred to in Equation (1). Equation (3) is basically
Equation (1) with the adopted value n=1, (White,
2002). Eurocode uses the same equation with certain recommended values for parametercwhich relates to the soil characteristics, ranging from 0.5 for soft soil to 0.7 for stiff soil.
With a need for monitoring of vibrations during the execution of works defined, the flowchart on Figure 5 specifies main activities to be taken to control the adverse impact of vibrations caused by pile driving.
Generally, the adverse impact of vibrations on nearby buildings depends on the intensity and frequency characteristics of the ground vibrations arising from piling and characteristics of the threatened building and soil. Considering the previously mentioned, the first two columns of the flowchart show main activities which are designed to lead to the data about a) the amplitude and frequency range and interpretation of data about vibrations which are induced by application of particular pile driving technology on the specific site; b) the assessment of vulnerability of building exposed to pile driving, including the settlement of building and ground. The third column shows main steps for monitoring of the nearby building which is threatened by vibrations arising from pile driving.
Figure 5.
Flowchart of main works and actions related to risk assessment and monitoring of residential building during pile driving
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Methods For Predicting Impact of Ground Vibrations Induced By Pile Driving on The Old Masonry Wall Buildings and Their Monitoring
Soil particles during pile driving operations have three velocity components which can be measured simultaneously using a triaxial geophone. Minimum two geophones are used, allowing for simultaneous measurement of velocity at two mutually distanced positions, most often between the pile and the nearby building. Peak particle velocity is defined as the maximum of the vector sum of three mutually perpendicular particle velocities (Hiller et al.,
1998), Equation (4).
(4)
To obtain maximum particle velocities as defined in Equation (4), during pile driving operations it is necessary to perform on-site measurement of three perpendicular components of vibrations which feature in Equation (4). Triaxial geophones, which permit for independent measurement of velocities in three orthogonal directions, are used for measurement of the velocity components. One direction of geophone is set in the direction of the source of vibrations (pile that is driven), while the second one is positioned in vertical direction and the third one - in horizontal, perpendicular to the longitudinal direction. It should be mentioned that
Equation (4) takes into account the maximum measured values of the velocity components, although these may not occur simultaneously.
Therefore, the maximum peak velocity is actually a simulated resultant of the simultaneous action of all component velocities.
During seismic wave propagation through the soil, as the distance from the source of vibrations
(pile) grows, energy dissipation (caused by soil material damping) occurs, whereby frictional losses occur during propagation. In addition to frictional losses, the dissipation can occur to some smaller extent, as a result of seismic energy reflection and refraction in relatively homogenous soil. These effects are more evident in the conditions of heterogeneous ground with stiff soil layers and in case of a small depth of bedrock.
Key indicator of the condition of building and in particular its resistance to vibrations during pile driving is the seismic resistance during natural seismic impact. In this sense, the first step to be taken in the analysis of the building to be exposed to industrial vibrations is to define a maximum level of seismic excitation the building can sustain without major damages during natural seismic impact. Although (as it was mentioned) there are significant differences in intensity and frequency range between natural seismic waves and pile driving induced waves, the resistance of building on earthquake impact represents a sound starting point for the assessment of the building resistance to dynamic impact (vibrations).
Resistance of the building during natural seismic impact (according to the applicable technical rules and standards) can be assessed exactly, regardless of the structural system and building material. Additionally, degree of seismic resistance expressed through the MSC scale allows for assessment of building resistance to industrial vibrations according to some standards. Although being experience-based, these standards provide quite acceptable results for the engineering practice.
Second important aspect of assessment of building resistance to vibrations is the applied building material, its age and natural resistance to cyclic loading. Therefore, masonry wall structures (rock, brick etc.) are far less resistant to cyclic and impact loading compared to concrete, reinforced-concrete or steel structures. Third important element of the building resistance to vibrations is the functional resistance. Sometimes the building resistance to damage is not critical; however the building cannot be used during construction period due to the unbearable physiological effects, or in case of industrial buildings machines cannot operate because of the increased level of external vibrations.
A specific aspect of analysis is the assessment of possible impact of vibrations on the soil compaction, which can cause settlement of foundation or occurrence of liquefaction that can lead to a complete instantaneous collapse of the foundation soil.
Monitoring of building during pile driving operations is performed based on the monitoring program. Monitoring program in relation to the impact of vibrations caused by pile driving has to be developed in several phases that are not common for similar programs of new structures or monitoring programs during operational phase of building under common loads and impacts.
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Zvonko Tomanovic
In case of old buildings and historically important buildings, an important step in developing a monitoring program is familiarisation with the available technical documents, construction technology and acquisition of other relevant information. Comprehensive understanding of the construction technology, applied building materials and challenges encountered during construction period represent a prerequisite for understanding of condition of the existing building or any visible damages.
In cases when there is no available information on the existing condition of the building, such information should be obtained based on the fieldworks the aim of which is to define the condition of foundation, condition of the applied vertical and horizontal support elements of the structure, geological conditions under and around the building, condition of groundwater etc. A particular segment in the analysis of the existing condition is testing of the quality of building materials, which is performed partly on the building itself and partly by means of laboratory testing of samples.
Based on the general assessment of condition of building and structural system, it is possible to define the procedures for monitoring of the building during construction period, i.e. duration of industrial vibrations. Response of the building to vibrations is monitored by means of visual inspections, measurement of width of the existing cracks, monitoring of the soil on site etc. The scope of visual inspections should include structural elements but also the cosmetic damages which compromise the building functionality.
Residential building Vila Lepes was built in 40s of the last century. Based on the original construction drawings, technical description dating back in 1936 and photos taken during construction period (See Figure 6a and Figure 6b), it has been ascertained that the structure of the building is consisted of the brick masonry walls (at the floor level tied in concrete beam rings that are reinforced by massive stirrup iron), wood beam ceilings and timber roof structure. The thickness of walls varies, starting with 60cm at the ground floor and decreasing with the number of stories to reach some 30 cm at the last storey. Footings are
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7.0), the timber mezzanine floors in one half of the building have been removed and replaced with reinforced-concrete slabs.
Figure 6.
a) Building under construction 1936; b)
Present image of the building, 2014;
Resistance of the building during natural seismic impact should indicate global resistance of a building to seismic activities in compliance with the applicable regulations. This evaluation used a numerical model developed in software package “Tower build 7” (the model used for the analysis concerned is illustrated in Figure 7). The resistance of building on earthquake impact offers a general picture about the condition of building, and also helps in categorisation of building when applying certain standards for assessment of vulnerability of building to artificially induced vibrations arising from pile driving.
Seismic analysis of building suggested that all support walls have the capacity to resist
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Methods For Predicting Impact of Ground Vibrations Induced By Pile Driving on The Old Masonry Wall Buildings and Their Monitoring earthquake intensity of 8 degrees in the MSC scale, yet the full capacity of the majority of walls has not been reached. The capacity of walls has only been reached in some walls of the last storey, hence it can be concluded that no major damages to the building are expected in case of an earthquake of seismic intensity between 8th and 9th degree in the MSC scale.
Figure 7.
3D Model for seismic analysis of the structure of Vila Lepes
The analysis indicated that the building structure is consisted of masonry walls which are favourably distributed in the longitudinal and transverse direction. Calculation-based verification of the building response to seismic activity and historical data on relatively small damages caused by the 1979 earthquake (the earthquake intensity in Tivat area was between
8th and 9th degree in the MSC scale), suggest that the building is well constructed and can sustain the design seismic loads, as defined by regulations, without any major damage.
The source of vibration is a pile-driver that presses in the steel protective column into the ground. The pile-driver uses a diesel hammer model DELMAG D30-32/33, which allows for
15 to 20 blows per minute, Figure 8. It was planned to drive total number of 520 piles, the diameter of which is 600mm and the length between 14m and 20m. The wider area of the location concerned is mostly composed of flysch sediments of Upper Eocene age, represented by clay facies, marls and sandstones. Over the flysch sediments layer there are deposits of deluvial-proluvial, alluvial and marine sediments (thickness of 3-25m) of
Quaternary age. On the top of these deposits there is a layer of anthropogenic sediments - embankment. Characteristic geologic profile of terrain along the measured direction 1 (Figure
10) is illustrated in Figure 9.
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Figure 8.
Driving of Piles No 72 and No 226
For the purposes of the damage risk assessment of Vila Lepes, the construction of pile-supported foundations at location UP1-09 Porto Montenegro in Tivat (Figure 10), was preceded by monitoring of mechanical (seismic) effects of driving the steel protective column on the nearby residential buildings. On this occasion, the multichannel seismic apparatus equipped with appropriate seismic sensors was deployed to allow for monitoring of multiple series of excitement induced by a low-frequency mechanical hammer used for driving of protective column during pile installation at characteristic locations.
Physical values of parameters of the particle oscillations (displacement, velocity, acceleration) that occur due to vibrations are measured at different distances from the source of vibration
- pile and foundations of the building. The velocity of soil particle oscillation of the artificially induced medium is prevailingly recorded at the measurement sites. The measurement was performed by means of a mobile seismograph equipped with a 3-component geophone
(Figure 11), whereby each records one of three components of particle oscillation velocity of the excited medium at the site concerned. The measurements involved application of a digital seismograph “INSTANTEL Blast Mate Series
III”, with an 8-channel option. The instrument directly records the velocity of oscillation of the excited particles (velosigrams). Two geophones are placed in the horizontal plane, one in the direction of the detection point (to record longitudinal component of oscillations)and the third geophone is placed perpendicular to the horizontal plane (to record vertical component of oscillations). Based on the frequencies recorded on velosigrams, by means of numerical derivation and integration it is possible to obtain the two
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Methods For Predicting Impact of Ground Vibrations Induced By Pile Driving on The Old Masonry Wall Buildings and Their Monitoring other components of vibrations - displacement
(vibrogram) and acceleration (accelerogram) of the excited particles. The use of fast Fourier transform enables to obtain a complete spectral analysis of the recorded vibrograms, velosigrams or accelerograms.
Instrumental recording of vibrations induced by pile driving at the location of the marina Porto
Montenegro Tivat has been carried out along two directions. Both directions are oriented to the most critical nearby building - Vila Lepes, at which the measurement site MP– 1 is located, and according to the design locations, two piles -
Pile No 72 and Pile No 226, as illustrated in Figure
10.The measurement site MP – 1 is located at the foundation of the building Vila Lepes (Figure
11), while the other measurement sites are located on the ground along two directions - the first being located between MM – 1 and pile No
72: MP – 2, 3 and 4; and the second between MP
– 2 and pile No 226: MP – 5, 6 and 7, Figure 10.
Figure 9.
Characteristic geologic profile of terrain along the measured direction 1
Figure 10.
Disposition of geophysical-seismic measurements of vibrations
Figure 11.
Geophone at the building foundation
(“INSTANTEL Blast Mate Series III”)
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Recording of the series of seismograms was followed by the amplitude and frequency analysis and interpretation of data with the aim to define maximum oscillations that occurred at measurement points (according to the velocity and acceleration parameters of oscillation) as well as the dominant frequencies of excitation. Characteristic recording of the performed measurement is illustrated in Figure 12.
Figure 11.
Three component velosigrams: recording: S - 72 _ MP – 2
Based on experimental data obtained from the in-situ measurements, empirical (power of law) relationship has been defined for the process of attenuation of seismic effects of vibrations as measured through the particle velocity V= f(r), caused by the pile driving operations at the location of the building. The following exponential function has been used:
(5) where:
V - Velocity of particle oscillations caused by vibrations of the excited medium; [mm/s], r- Distance of the source of vibrations from the measurement point - MP; [m],
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K - Constant relating to the pile driving technology and engineering-geological conditions of the working environment n - Constant (exponent) relating to the distance of the measurement point - MP from the source of seismic waves - vibrations and applied power per hammer blow.
The method of least squares has been applied to the acquired data set to define numerical values of the coefficients K and n. The resulting numerical values of these coefficients, for the particle velocities as measured in-situ, are
K=14003 and n=-2.5046, at the regression coefficient R²=0.8924, as shown below:
(6)
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Methods For Predicting Impact of Ground Vibrations Induced By Pile Driving on The Old Masonry Wall Buildings and Their Monitoring
The results of measurement of the particle velocity at discrete points on the terrain surface and functional correlation of the particle velocity with the distance from the pile (6) are graphically illustrated by diagram in Figure 13.
For the purposes of mathematical processing of data, the following recorded data have been used: a) at measurement point MP – 1 (at foundation of the building Vila Lepes), all the recordings for both directions have been used; b) for the remaining measurement points, the records made at driving of deeper elements of piles No 72 and No 226 have been used, given that the vibration intensities originating from these rock mass segments are significantly amplified
(as the pile penetrates into more consolidated bedrock). The non-consolidated rock materials dipping in multiple directions increase the vibration amplitude by multiple times.
For “safety level”(measured particle velocity for all measurement data is below the level defined by expression: V=K_1*r^(-n_1 )), based on aquired data, numerical values of coefficients K_1and n_1have been defined, and further dependence of the particle velocity on the distance:
V = 15000* r -2.4 (7)
Empirical function (3) that is recommended by EC3 for prediction of distribution of particle velocity is also illustrated by diagram in
Figure 13. Energy per blow of the applied pile-driving hammer ranges between 94.765J
- 47.970J, hence the values for both limit values for energy per blow are shown.
The value for parameter c = 0.5 has been adopted based on the ground description and geological profile. Particle velocity defined by mathematical correlations (6) and (7) occurs to be between limit values as per EC3, for the level of the allowable velocity of 5mm/s.
However, prediction of the velocity attenuation according to EC3 offers overrated values of velocity for the distances over approximately
20m, and underrated values of velocity for the distances below approximately 20m.
Figure 13.
Functional correlation of attenuation of intensity of seismic effects of vibrations caused by pile driving
Results of the spectral analysis of the recorded seismograms shows that the range of recorded frequencies of oscillations is below 10Hz, which indicates to the fact that physical and mechanical properties of the foundation soil of the building
Vila Lepes are very unfavourable to these kinds of excitations. According to the standard DIN
4150/3, the intensity of seismic effects of the artificially induced quakes due to pile driving, measured through the particle velocity, is limited to the level of up to 5 mm/s (for the recorded frequency range < 10Hz). The same limit value with a view to the allowable particle velocity for the building concerned is also recommended by EC3. When the relationship power of law (2) according to the Swedish standard SS 02542
11 (SS1999) is applied, the following allowable velocity is obtained: mm/s which is approximate to the limit values as listed in tables of DIN and EC3 standards.
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Pursuant to the limited level of intensity of the seismic effects of vibrations caused by the pile driving operations at the Vila Lepes building location, the following three zones
(in terms of adverse effects of vibrations on the building) have been defined with a view to the distance of the pile installation point from the building: a) distance up to 25.0 m – zone A, in which the intensity of seismic effects of vibrations exceeds the limit value of 5 mm/s, b) distance from 25.0 to 28.0 m - zone B, in which pile driving operations should be accompanied by instrumental recording, due to the “dissipation of data” (safety level) that occurs because of dominant heterogeneity of the engineering-geological and geotechnical conditions of the soil and c) distance over 28.0m – zone C, in which vibrations induced by pile driving will not cause any damage to the structural elements of Vila Lepes.
For the reasons of cost-effectiveness, the
Operator decided to apply the same technology for construction of all piles, including zone A as well, where piles are installed at the distance less than 25m away from the threatened building. In order to eliminate any risk (life and health of the building occupants), the arrangements were made for the occupants to temporarily move out from the building during pile driving operations in zone B. Monitoring of the building is planned to control risk of eventual damages to the residential building Vila Lepes that may occur as a result of the pile driving operations at the location UP1-
09 (at distance from the building between 126m and 18m). The monitoring of building during pile driving operations should include visual inspection, measurement of settlement of the building and measurement of development of width of the existing fissures and cracks in the building.
Defining the method of building monitoring is based on the as-built design of the building
(including also the existing damages), results of the building resistance to natural seismic impact, results of filed measurements of vibrations
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The scope of visual inspection included all visible parts of the building Vila Lepes and the surrounding terrain. Consideration was given to any damage to the building structure such as presence of new cracks in supporting walls of more than 0.3 mm in width and 50cm in length, large pieces of plaster falling off ceilings (pieces larger than 2-3cm) etc. During monitoring of the building, no damages to the building have been observed except for some minor damages of the ceiling due to falling off plaster pieces of 1-2cm in size in the part of the building where ceilings are constructed of wooden beams.
The existing cracks are signs of weakening of the structure of support walls, being potentially the most vulnerable points for further damages in terms of the crack length propagation and width expansion. Considering this, eight points have been selected on the building for measurement of the crack width with the precision of 1/1000mm and installation of control glasses, Figure 14.
After all 520 piles were installed, the maximum measured increase in the width of the existing cracks was 0.05mm.
Measurement of settlement of foundations can indicate to the impact of vibrations on the changes in the structure of soil underlying the foundations. Considering that the upper soil layers at the observed location are consisted of soft sediments that are under a constant level of the ground water, the settlement of the building may occur during pile driving operations as a result of the vibrations induced by pile-driving.
In addition to this, soil compaction due to driving of the protective column for the pile construction causes lateral pressure, which can lead to the soil heaving and uplifting of the building, when driving piles at close distance from the building.
With reference to the above mentioned, the monitoring program accounted for measurement of settlement of the building Vila Lepes. The measurement of settlement was performed geodetically - by precise levelling (0.01mm reading, minimum measurement accuracy
0.1mm). The measurement of settlement was performed at 7 benchmarks installed immediately above the building foundations, at some 20cm
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Methods For Predicting Impact of Ground Vibrations Induced By Pile Driving on The Old Masonry Wall Buildings and Their Monitoring
Figure 14.
Image of the existing crack on the support middle wall of the building on the last storey, with measurement points for measurement of changes in the crack width (a detail is shown on the right) above the existing pavement of the building.
The measured settlements of the building ranged between +0,3mm and -1.1mm (+ uplift,
- settlement).
In terms of load-bearing capacity and costeffectiveness, the driven piles represent a sound technical solution. Nevertheless, from the aspect of vibrations, noise and environmental pollution, application of these piles has some serious downsides, in particular when installed in urban areas. For the purposes of assessment of risk to the nearby buildings, it is necessary to analyse geological, engineering-geological, hydrogeological and seismic and tectonic conditions of the rock mass in the wider area of the location, and further the nature and type of the nearby buildings being located in the segments of urban zones, which belong to the catchment area of the construction site. In compliance with standards, norms and criteria that are applicable in different countries, negative effects of the applied technology of pile driving, according to international standards, are constrained by limitation of the soil particle velocity for particular frequency range.
Applicable standards mainly use the particle velocity for the purposes of risk assessment of potential damages to buildings, although the soil particle velocity itself is not a cause of damage.
Namely, the criteria which feature in the standards are generally established based on the statistical analysis of damages to the nearby structures during pile driving operations and on correlation with the induced ground vibration. Another important indicator of potential damage, in addition to the particle velocity, is the wave frequency and number of cycles to which the structure will be exposed. The wave frequency is included in certain standards
(DIN 4150; BS5228), however the number of cycles to which the structure will be exposed is not quantified in the standards concerned.
The standards concerned categorise the buildings exposed to technical vibrations in a similar way, with almost the same allowable particle velocity at the pile-driving operations.
Empirical function for prediction of wave velocity as recommended by some Eurocode in the analysed case allows for good assessment of the wave velocity in the zone of allowable velocities for residential buildings. However, prediction of the velocity attenuation according to EC3 offers overrated values of velocity for distances over approximately 20m and underrated values of
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Zvonko Tomanovic velocity for distances under approximately 20m.
This is further confirmed by the fact that in this particular case, some thirty piles were installed in the zone which offers approximately two times higher velocities compared to those specified in standards, while on the other side the building did not sustain any structural damages but only some minor cosmetic damages.
Previously mentioned suggests that according to
EC3 the assessed velocity at the same location would be overrated for the purpose of analysis of impact on churches, buildings of historical importance etc., and would be underrated in case of analysis of impact on commercial and industrial buildings of reinforced concrete or steel. Generally, the other standards also define conservative limit values for the particle velocity at impact of the technical vibrations.
The above mentioned indicates that the peak particle velocity induced by technical vibrations, pile-driving in this particular case, due to a large number of influential factors can only be defined reliably by field measurements. The scope of research depends on the complexity of geological medium, presence of underground water, soil type, conditions for occurrence of soil compaction or liquefaction, type of building that is exposed to vibrations (including the structural system and building materials). It is evident that such a large number of influential factors requires field recording of the intensity of seismic effects of vibrations, and later also monitoring of vibrations during pile-driving works, as well as the monitoring of the nearby building that is exposed to vibrations, with an aim to reduce the risk of adverse effects associated with the application of driven piles.
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[4] BS5228 (1992). “Part 4: Noise control on construction and open sites, code of practice for noise and vibration control applicable to piling operations.” British
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[13] Seed, H. B. and M. L. Silver (1972)
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