10NCEE
Tenth U.S. National Conference on Earthquake Engineering
Frontiers of Earthquake Engineering
July 21-25, 2014
Anchorage, Alaska
SIMULATED SEISMIC TESTING OF
PARTIALLY-GROUTED MASONRY SUBASSEMBLAGES
Catherine A. Johnson1 and Arturo E. Schultz2
ABSTRACT
While most masonry structures constructed in the West Coast of the U.S. are reinforced and fully
grouted, almost all reinforced masonry construction in the rest of the country, including regions
of high seismic risk, utilizes partial grouting. The popularity of partial grouting is a result of
lower material costs, faster rates of construction, and reduced potential for installation problems
(e.g. inadequate consolidation of grout). Even though walls are the main lateral load resisting
elements in a masonry structure, the seismic performance of partially-grouted reinforced
masonry wall systems has not been studied sufficiently and is not understood well given the
complexity of their behavior. Recent studies have indicated that current design provisions for
partially-grouted masonry walls are not as well developed as those for fully-grouted, reinforced
masonry.
This paper provides a summary of the test of the first specimen in an ongoing experimental effort
to determine the lateral load resistance characteristics of full-size, three-dimensional subassemblages representing portions of partially grouted (PG) masonry buildings. The subassemblages comprise a shear wall with a window opening, as well as cross walls at either end.
The test specimen includes a concrete foundation and a composite roof (i.e. a hollow-core
concrete slab with a reinforced concrete topping), and it is loaded by subjecting the roof
diaphragm to a quasi-static, cyclic drift history. The test was conducted by researchers from the
University of Minnesota at the RTMD Laboratory of Lehigh University that is one of the 14 sites
in the NEES (Network for Earthquake Engineering Simulation) program sponsored by the U.S.
National Science Foundation (NSF). The specimen is a full-size representation of a portion of a
building structure that will be tested at the University of California, San Diego. This paper
describes the load-deformation characteristics of the specimens including failure modes, load vs.
displacement response, and strength and stiffness deterioration.
1
Grad. Student, Department of Civil Engineering, University of Minnesota, Minneapolis, joh04739@umn.edu
Professor, Department of Civil Engineering, University of Minnesota, Minneapolis, aeschultz@umn.edu
2
Johnson, Schultz. Seismic Testing of Partially-Grouted Masonry Sub-Assemblages. Proceedings of the 10th National
Conference in Earthquake Engineering, Earthquake Engineering Research Institute, Anchorage, AK, 2014.
Simulated Seismic Testing of Partially-Grouted Masonry SubAssemblages
Catherine A. Johnson1, Arturo E. Schultz2
ABSTRACT
While most masonry structures constructed in the West Coast of the U.S. are reinforced and fully grouted, almost all
reinforced masonry construction in the rest of the country, including regions of high seismic risk, utilizes partial
grouting. The popularity of partial grouting is a result of lower material costs, faster rates of construction, and
reduced potential for installation problems (e.g. inadequate consolidation of grout). Even though walls are the main
lateral load resisting elements in a masonry structure, the seismic performance of partially-grouted reinforced
masonry wall systems has not been studied sufficiently and is not understood well given the complexity of their
behavior. Recent studies have indicated that current design provisions for partially-grouted masonry walls are not as
well developed as those for fully-grouted, reinforced masonry.
This paper provides a summary of the test of the first specimen in an ongoing experimental effort to determine the
lateral load resistance characteristics of full-size, three-dimensional sub-assemblages representing portions of
partially grouted (PG) masonry buildings. The sub-assemblages comprise a shear wall with a window opening, as
well as cross walls at either end. The test specimen includes a concrete foundation and a composite roof (i.e. a
hollow-core concrete slab with a reinforced concrete topping), and it is loaded by subjecting the roof diaphragm to a
quasi-static, cyclic drift history. The test was conducted by researchers from the University of Minnesota at the
RTMD Laboratory of Lehigh University that is one of the 14 sites in the NEES (Network for Earthquake
Engineering Simulation) program sponsored by the U.S. National Science Foundation (NSF). The specimen is a
full-size representation of a portion of a building structure that will be tested at the University of California, San
Diego. This paper describes the load-deformation characteristics of the specimens including failure modes, load vs.
displacement response, and strength and stiffness deterioration.
Introduction
In the Midwestern, Eastern, and Northwestern United States of America lateral-load resisting
masonry walls are partially grouted, as opposed to their fully grouted counterparts in the West
Coast. The former have grout only in cells containing steel bar reinforcement, either vertical or
horizontal, to form a grouted network that is surrounded by hollow, unreinforced masonry.
Partially grouted (PG) walls are used due to their ease of construction and lower material costs in
comparison to fully grouted (FG) walls.
Partially grouted walls are currently designed according to the Masonry Standards Joint
Committee standard1. The in-plane shear strength equation for walls in the MSJC standard was
developed using data from experiments on FG walls. For use with PG walls, the net horizontal
area is substituted in place of the gross area, and a coefficient for PG walls is provided1. The
__________________
1
Grad. Student, Department of Civil Engineering, University of Minnesota, Minneapolis, joh04739@umn.edu
Professor, Department of Civil Engineering, University of Minnesota, Minneapolis, aeschultz@umn.edu
2
Johnson, Schultz. Seismic Testing of Partially-Grouted Masonry Sub-Assemblages. Proceedings of the 10th National
Conference in Earthquake Engineering, Earthquake Engineering Research Institute, Anchorage, AK, 2014.
limited amount of experimentation on the PG walls has shown that the MSJC equation without
the PG coefficient is often unconservative because PG walls behave differently from FG walls2,3.
This limited group of experiments do not include full-scale, three dimensional sub-assemblages
that more accurately represent the walls in PG masonry structures. The primary objective of this
study is to further examine the behavior of PG walls through cyclic, quasi-static testing of threedimensional sub-assemblages. The behavior of the first specimen test is described through
failure modes, load vs. displacement response, and strength and stiffness deterioration.
In order to fully understand the behavior of PG walls and evaluate the effect of different aspects
of wall construction, a three-phase study sponsored by the National Science Foundation NEES
program is being conducted by the University of Minnesota, Drexel University, and University
of California, San Diego (UCSD) under the leadership of UCSD. Drexel is conducting a series
of in-plane wall tests to validate previous research, as well as evaluate the effects of
reinforcement configurations, end conditions, axial load, and bed-joint reinforcement. UCSD is
conducting full-scale shake-table tests on partially grouted masonry structures, in order to
understand a composite system. The University of Minnesota tests are meant to further
understand the behavior of PG wall assemblages.
Experimental Program
Program Plan and Specimen Configuration
The wall assemblage test specimen at Minnesota consists of a three-dimensional PG assemblage
containing a window opening. Figure 1a and 1b show that the specimen has a C-shaped plan that
is 6502mm long and 4267mm tall, with 1219mm long flanges at either. Herein the 6502mm
long portion of the wall parallel to lateral loading will be referred to as the web, and the 1219mm
sections perpendicular to loading will be designated the flanges. The wall also contains a
2438mm x 1016mm opening in the center of the web. All concrete masonry units were mortared
only on the face shell, with the exception of units containing vertical reinforcement, where the
webs were also mortared to prevent the grout from flowing out of the cell.
The bond beams were created using closed bond beam units, except where vertical reinforcement
passed through the bond beam. Here open-bottom bond beam units were used. All horizontal
reinforcement bars were hooked around the vertical reinforcement bars using a code-compliant
180° hook, as required by MSJC provisions1. In all bond beams, horizontal reinforcement was
also placed in the flanges. There were four bond beams (one at the top and bottom of the wall, as
well as above and below the opening) and six grouted vertical cells (one at the end of each
flange, the end of the wall, and on either side of the opening as shown in in Figure 1a). A
254mm hollow-core plank was placed on top of the wall and it was extended the entire length of
the wall, resting only on the flanges. A 356mm reinforced concrete beam was poured directly on
the top of the wall, and on three sides of the hollow-core plank. The beam extended above the
top of the hollow-core plank and was monolithic with a 102mm thick reinforced concrete
topping. These measures were taken to emulate a floor or roof system, as well as to ensure
integrity between the wall and the roof diaphragm.
(a)
(b)
(c)
Figure 1. Details of PG Wall Specimen: (a) web layout; (b) side view; (c) plan view.
Materials
The concrete masonry blocks were 400mm (16 in.) long by 200mm (8 in.) wide by 200mm (8
in.) high, and satisfied ASTM C904. SPEC MIX® Portland Cement, Lime & Sand Type S
Mortar was used with a compressive strength of 12.7 MPa (1840 psi) measured according to
ASTM C1095, and it satisfied ASTM C2706 requirements. SPEC MIX® Self-Consolidating
Grout with Fine Mixtures was used with a compressive strength of 29.3 MPa (4240 psi)
measured according to ASTM C10197. The grout had a slump of 200mm (10 in.) and satisfied
ASTM C4768 requirements. All steel in the wall consisted of Gr 60 #4 Deformed bars according,
and satisfied ASTM A6159. Specified compressive strength of the masonry was 13.8 MPa.
Test Procedure and Instrumentation
Strain gages were placed on both ends and the center of all horizontal reinforcement on both the
top and the bottom of the bar. They were also placed on both sides of the vertical reinforcement
at a distance of 50mm below the top of the foundation, and 610mm above the foundation.
LVDTs were used to measure the relative sliding between the foundation and the wall, the wall
and the hollow-core plank, the hollow-core plank and the topping, and the wall and the topping.
Internal LVDTs in the actuator measured the actuator displacements. A string-pot located at the
top of the wall measured total drift. Two LVDTs were used to measure the vertical displacement
from the top of the piers to the foundation, and two LVDTs were placed diagonally on each pier
to measure pier deformation.
Lateral loads were imposed by two actuators attached to loading beams that were anchored into
the concrete topping shown in Figure 2a. An external vertical load was applied to the wall by
means of four Dywidag post tensioned rods that extended from the foundation to the roof
structure. Because the load would increase as strain was induced in the rods from drift, a spring
was placed between the monitoring load cells and the nut to introduce compliance and keep this
load approximately constant. Figure 2b shows the details of this set-up at the top of the wall. In
addition to the externally applied load, the weight of the roof and the loading beams added
additional axial stress. The total axial stress on each pier (the two sections of the wall on either
side of the opening) was approximately 0.3 MPa to simulate gravity loads. Load cells were used
to monitor the load in the Dywidag post tensioned rods, as well as the actuators.
The displacements for each cycle were determined by drift level, and were completed twice in
order to measure strength and stiffness degradation. Figure 3 shows the cycles used. The drift
levels (ratio of displacement at the top of the wall to the height of the wall) were chosen to
include the performance limits defined by the Applied Technology Council10. These limits are
also noted on Figure 3a. The combination of two actuators was used to avoid torsion from
eccentricity of the wall. The test was run until the combined load on the system (i.e., the sum of
the actuator loads) dropped to 50% of the maximum combined load measured for the system.
This was done in order to provide understanding of the wall behavior up through significant
strength deterioration.
(a)
Figure 2.
(b)
Testing configurations (a) Test-setup; (b) Axial load set-up at the top of the wall.
Experimental Results and Discussion
Global Response
Figure 3b shows the load vs. top drift index plot. The drift index for the wall is the ratio of
lateral movement at the top of the wall minus the sliding between the wall and the foundation to
the height of the wall (4267mm). The wall experienced an average maximum shear force, Vexp,
of 273.4 kN (61.58 kips) corresponding to an average wall drift of 0.29%. These are values
averaged over the two loading directions. The drift at the top of the wall continued to increase as
the wall continued to degrade. At 0.45Vexp the top of the wall had reached a drift index of 0.69%.
The end of useful wall capacity, that is functional failure, is defined as the point at which the
peak load for a given cycle drops 15% below the maximum value previously measured (Vexp) or
0.85Vexp. The drift index at 0.85Vexp at the top of the wall was 0.37%.
(a)
Figure 3.
(b)
Loading history details (a) Loading procedure; (b) Lateral Load vs. Drift History.
The wall hysteresis loops in Fig. 3b illustrate some pinching, due to the shear damage that
occurred within the wall. However, significant energy dissipation is observed qualitatively once
functional failure is reached because of the frictional resistance provided by the axial load.
Figure 4a illustrates how the axial load applied through the Dywidag post tensioned rods varied
with drift. As the wall moved away from its neutral position axial load increased. There was a 6
kN difference between the minimum and maximum load applied to the wall. This is a 9%
increase from the minimum load applied, which was found acceptable.
(a)
(b)
Figure 4. Wall response to drift (a) Axial Load vs. Interstory Drift Index; (b) Strain vs. Drift
for horizontal rebar below window that yielded. Center, right, and left refer to the strain gage
location on the bar.
Wall Cracking
The wall behaved elastically through the first two cycles, which brought the top of the wall to a
drift of 0.05%. Once the wall continued into the third cycle (drift of 0.1%) cracking began to
appear. The final crack pattern of the wall can be observed in Figure 5. A majority of the
damage to the structure was located within the piers. Significant diagonal shear cracking
occurred in the piers on either side of the opening. While most of the cracks were along the
mortar lines (joint separation), some cracks extended through the masonry units. Some diagonal
cracks formed in the spandrels above and below the opening and extended from the center of the
wall out to the edges. One of the flanges experienced vertical cracking along the reinforced cell
that was also part of the web of the wall, while the other flange experienced spalling at the
bottom corner adjacent to the web.
(a)
Figure 5.
(b)
Cracking at the end of the test: (a) Cracking in left pier; (b) Cracking in right pier.
Relative Sliding
Very little sliding was observed between the structural components (i.e., foundation, wall, roof
structure). Figure 6 illustrates the relative sliding between each structural component compared
to the interstory drift. The maximum sliding displacement occurred between the foundation and
the wall, and it was 0.17mm. This is only 0.5% of the maximum total drift. There was a
maximum amount of 0.018mm of sliding between the wall and the cast-in-place concrete beam
on top of the wall, 0.11mm of sliding between the hollow-core plank and the topping, and 0.14
mm between the hollow-core plank and the wall. However, the large crack widths that were
created during the tests imply non-trivial amounts of internal sliding, i.e. between masonry
surfaces along cracks.
Rebar Yielding
Strain gages indicated that yielding occurred within the steel reinforcement in three locations; in
two vertical bars, and one horizontal bar. Figure 2b highlights, by means of red circles, the
locations where yielding occurred. Figure 4b and 7 shows the strain in the bar versus the total
drift of the wall. The strains were determined as the average of the measured strain between
strain gages on either side of the rebar in each location. The yielding in the vertical rebar was
found in the foundation, while the yielding in the horizontal rebar was found at the edge of the
wall. The vertical bar yielding was unexpected since the wall was designed to be shear-critical,
and a large factor of safety was adopted to avoid flexural failure. That the bottom of the vertical
bars yielded is indicative of inaccurate understanding of the resistance mechanisms in PG shear
walls. The horizontal bar yielding is consistent with the piers sustaining most of the damage and,
therefore, inelastic action.
Figure 6. Relative sliding between each structural component compared to the interstory drift.
Shear Strain in Piers
Figure 8 shows the shear force vs. gross shear strain relation for each. The gross shear strains
were computed from the deflections measured along diagonal LVDTs placed across each pier.
Because instrumentation was not included in the test to determine the load in each pier, it is
assumed that the piers resisted equal loads during the test. Thus, the shear force in each pier is
one-half of the total lateral load. It is noted that the data for the right pier was stopped 2 cycles
earlier than that for the left pier due to damage to one of the LVDT mounts. Plastic shear strain
began to accumulate after cracking was introduced in the piers. The response of the piers after
reaching peak load was highly pinched as large diagonal cracks formed in the masonry. Upon
unloading and reloading, the piers revert to a frictional type of response as the masonry surfaces
along cracked bed joints slide under the action of the nearly constant vertical stresses. The
response of the left pier is more regular once peak load is reached. The right pier, however,
appears to experience more damage as the wall is deflected further, and the photographic
evidence supports this notion (Fig. 5b). The magnitude of the gross shear strains, approaching
0.2, is further evidence of the extensive cracking damage to the piers.
Code Estimates of Shear Strength
As stated earlier, the maximum shear force measured in the experiment, Vexp, was found to be
273.4 kN (61.58 kips). An objective of the project is to evaluate the accuracy of the current code
equations for shear capacity of partially grouted walls. Because the piers were the elements that
were most highly damaged, shear strength of the wall is defined on the basis of the shear
strengths of the individual piers. The shear strength of the wall was calculated using two
methods: the 2011 edition of the MSJC standard11 (Eq. 1), and the 2013 edition of the MSJC
standard1 which includes a factor (γg) to account for partial grouting (Eq. 2).
⎡
⎛ M ⎞⎤
0.5A h f yh d v
Vn = 0.083⎢4.0 −1.75⎜⎜
⎟⎟⎥ A n f m′ + 0.25σ n An +
sh
⎢⎣
⎝ Vd v ⎠⎥⎦
(1)
⎧⎪
⎡
⎛ M ⎞⎤
0.5A h f yh d v ⎫⎪
Vn = γ g ⎨0.083⎢4.0 −1.75⎜⎜
⎟⎟⎥ A n f m′ + 0.25σ n An +
⎬
sh
⎢⎣
⎝ Vd v ⎠⎥⎦
⎩⎪
⎭⎪
(2)
where M is the maximum bending moment; V is the applied shear force; dv is the depth of
the member; An is the net cross-sectional area; f’m is the masonry compression strength; σn is the
axial stress; Ah is the area of a single horizontal reinforcing bar; fyh is the yield strength of the
horizontal reinforcing bar; sh is the spacing of the horizontal shear reinforcement; and γg is the
grouting reduction factor, which is defined as 0.75 for PG walls.
(a)
(b)
Figure 7. Rebar strain history; (a) Strain vs. drift for vertical rebar at left edge of web that
yielded; (b) Strain vs. drift for vertical rebar at right edge of web that yielded. ‘Foundation’ and
‘Base of Wall’ refer to strain gage location on the bar.
From Eq. 1, the prediction for the shear strength, Vn, is 363.3 kN where Vexp/Vn is 0.75. Whereas,
from Eq. 2, the reduction for shear strength is 272.5 kN, and Vexp/Vn is 1.003. Thus, the grouting
reduction factor, γg, of 0.75 seems to be appropriate for assessing in-plane shear strength for the
wall in this study. It also verifies that without that reduction factor, the earlier MSJC11
expression over-predicts the shear strength of the wall by 33%.
However, a single test cannot validate the MSJC-131 formula for PG shear wall strength. It is not
evident that a constant value for γg will apply to all possible variations of the parameters
describing the configuration of a PG wall. Moreover, another potential flaw in the MSJC-111 and
MSJC-1311 shear strength formulas can be identified. In the analysis the height of the piers was
taken as the height of the opening. Within this section, no horizontal steel is present, and
therefore the Ah in the expression for the shear capacities above was taken as zero. However, as
stated above, yielding was found in both the steel above and below the window opening
implying that they contributed to the shear resistance of the wall. Further investigation into the
effects of the horizontal steel location on the shear capacity should be considered when
evaluating the MSJC in-plane shear strength provisions for PG walls.
When comparing the performance of the three-dimensional sub-assemblage in this paper to the
planar wall tests that have already been completed at Drexel12, a few differences should be noted.
The drift of the Drexel planar walls at maximum shear load was higher than that seen here
(0.34% vs. 0.29%). This is likely due to the flanges adding flexural resistance to deflection at
the top of the wall. It also appears that the flanges provided some confinement to the anchorage
of the bond beam reinforcement, as this test experienced less damage at the edge of the wall
where the bond beam reinforcement is anchored, than the planar wall tests completed by Drexel.
(a)
(b)
Figure 8. Shear strain vs. shear force within piers; (a) Left pier; (b) Right pier.
Conclusions and Recommendations
The test results reported here demonstrate the value of testing three-dimensional partially grouted
assemblages in order to better understand the behavior of PG walls. In comparison to the Drexel
planar walls, it can be inferred that the cross walls (i.e., flanges) in the test reported here affect
the behavior of the wall, reducing the drift, as well as allowing for better anchorage and thus
improved behavior of the horizontal shear reinforcement. In order to confirm this, as well as to
identify other characteristics of the system, additional three-dimensional assemblages tests are
needed. Additionally, further experimentation should be done to confirm the MSJC grouting
reduction factor, γg, as well as other provisions for in-plane shear strength of PG walls.
Acknowledgements
The authors gratefully acknowledge the financial support of the U.S. National Science
Foundation NEES Program (Award CMMI-1208208) and the Graduate School of the University
of Minnesota. The authors also thank the Pennsylvania Concrete Masonry Association (PCMA)
for the donation of the concrete blocks, mortar, and grout, and Oldcastle Concrete Products for
donating the hollow-core plank. The authors are also grateful to Professors Jim Ricles, Director
of the Real-Time Multi-Directional (RTMD) Hybrid Testing Facility at Lehigh University,
Professor Richard Sause, Director of the Advanced Technology for Large Structural Systems
(ATLSS) lab at Lehigh University, and the staff of both facilities for their assistance in the
construction, instrumentation and testing of the masonry wall assemblage.
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