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.
CONSOLIDATION BEHAVIOR OF AN
EMBANKMENT ON BOSTON BLUE CLAY
by
JOSEPH FRANCIS WHITTLE, JR.
BS, Rensselaer Polytechnic Institute
(1967)
Submitted in partial fulfillment
of the requirements for the degree of
Master of Science in Civil Engineering
at the
Massachusetts Institute of Technology
June, 1974
Signature of Author.
Depa •
nt
Certified by............
f Civil Engineer4i,
..
May 10, 1974
.................
. ...............
Thesis Supervisor
Accepted by .........................
Chairman, Departmental Committee on Graduate Studehts of the
Department of Civil Engineering
Archives
AUG 1 1974
ABSTRACT
CONSOLIDATION BEHAVIOR OF AN
EMBANKMENT ON BOSTON BLUE CLAY
by
JOSEPH FRANCIS WHITTLE,JR.
Submitted to the Department of Civil Engineering on May 10,
1974 in partial fulfillment of the requirements for the degree of Master of Science in Civil Engineering.
Since August, 1967, data have been collected from a
heavily instrumented section of an embankment for the proposed Route -95 near Boston, Massachusetts. The embankment is
40 feet high, with crest and base widths of 90 and 260 feet
respectively, and is underlain by 10 feet of fine sand and 135
feet of CL clay. The-field data, and laboratory test results
are presented graphically and in tabular form
The finite element program FEECON was used to determine
undrained deformations and stresses. When used with hyperbolic
stress-strain parameters from CKoUDSS tests on laboratory prepared samples of Boston Blue Clay, FEECON gives good results.
The selection of appropriate hyperbolic parameters is discussed. Based on comparisons with field data, the initial excess pore pressures beneath the embankment are best represen+ aAT t. Hented by the modified Henkel equation, Au = Act
kel's a parameter is related to Skempton's A parameer or the
appropriate stress history and stress conditions (plane strain
or direct-simple
shear).
Pore pressure and settlement data were used to determine the rates of consolidation within the clay. In the top 30
feet of clay where the overconsolidation ratio exceeds 2.5,
the consolidation settlement occurs more rapidly than pore
pressure dissipation, but the reverse is true in the lower 105
feet. Field compression parameters were computed from the
field data. The field RR's are 0.034 to 0.039, and the field
CR's are 0.28 to 0.39. These values are 50% and85% greater
than the laboratory values. Based on the field compression
parameters, the predicted final consolidation settlement at
the top of the clay beneath the embankment centerline is 7.8
feet.
Field coefficients of consolidation were computed by
Gray's transformation for a two-layer system considering lateral drainage with an isotropic permeability. The values
2
which gave the best prediction of settlement versus depth
at several times were based on incremental time analysis of
pore pressure data. These values are cvl = 0.71 ft2 /day (top
30 feet) and cv2 = 0.24 ft 2 /day
(lower 105 feet) and they ex-
ceed laboratory values by a factor of three.
Charles C. Ladd
Thesis Supervisor
Professor of Civil Engineering
Title
3
ACKNOWLEDGEMENT
This thesis would not have been possible without the
support of my wonderful wife, E'Lyse. She has freely given
her ILxve, patience and encouragement, and spent endless hours
reducing and tabulating data.
In addition, the financial assistance I have received through the Haley and Aldrich, Inc. Professional Development Program has greatly facilitated the achievement of
my educational goal.
Finally, my thanks to Professor Charles C. Ladd, who
so skillfully directed all my efforts in the preparation of
this thesis.
4
TABLE OF CONTENTS
Page
1.
2.
3.
Title Page
1
Abstract
2
Acknowledgement
4
Table of Contents
5
List of Tables
8
List of Figures
10
Introduction
13
1.1
Background
13
1.2
Purpose
13
1.3
Scope
14
Subsurface Conditions and Soil
Properties
17
2.1
General Geology
17
2.2
Soil Profile
18
2.3
Soil Properties
19
2.3.1
Index Properties
19
2.3.2
Stress History
19
2.3.3
Compression and Consolidation Data
20
2.3.4
Undrained Shear Strength
21
Construction
History
3.1
General
3.2
Embankment
32
32
Construction
5
32
3.3
4.
5.
6.
Instrumentation
Embankment
Page
Performance
38
4.1
General
38
4.2
Settlement
38
4.3
Pore Pressures
41
4.4
Horizontal
43
4.5
Internal Embankment Stress
Deflection
45
Parameters for Undrained Deformation and 69
Stress Finite Element Analysis
5.1
General
5.2
Granular
Soils
69
5.3
Cohesive
Soils
72
5.3.1
General
72
5.3.2
Hyperbolic Parameters
72
5.3.3
Shear Modulus
75
5.3.4
Poisson's Ratio
76
5.3.5
Bulk Modulus
77
5.3.6
Undrained Shear Strength
77
5.3.7
Initial Stress Level and Ko
77
5.3.8
Pore Pressure Parameters
78
5.3.9
Peat Parameters
78
69
Comparison of Predicted and Measured Un- 95
drained Deformation and Stresses
6.1
Yielding
6.2
Horizontal
6.3
Vertical
95
Deflection
Settlement
6
95
96
Page
7.
Stresses
6.4
Foundation
6.5
Pore Pressures
I
100
Analysis of Consolidation Behavior
123
7.1
Pore Pressure
123
7.2
Consolidation Settlement
125
7.3
125
7.2.1
General
7.2.2
Laboratory Compression Para-127
meters and Predicted Pcf
7.2.3
Field Compression Parametersl2 7
7.2.4
Predicted Final Consolidationl2 9
Settlement
7.2.5
Consolidation
130
Field Coefficients of Consolidationl32
7.3.1
General
132
7.3.2
Full Clay Thickness
132
7.3.3
Reduced Clay Thickness
134
7.3.4
Predicted Consolidation Set-13 5
tlement
8.
Conclusions and Recommendations
163
9.
References
166
Appendix A-1
Date-Day Conversion Chart16 9
Appendix A-2
Instrumentation
170
Appendix A-3
Construction History
173
Appendix A-4
Settlement Data
176
Appendix A-5
Well and Piezometer Data 184
Appendix A-6
Inclinometer Data
204
Appendix A-7
List of Symbols
211
7
LIST OF TABLES
Page
46
4-1
Maximum Measured Excess Head, Sta. 246
5-1
Hyperbolic Axial Stress-Strain Parameters 79
for Granular Materials
5-2
FEECON Parameters, Hyperbolic Axial
Stress-Strain Materials
80
5-3
FEECON Parameters, Hyperbolic Shear
Stress-Strain Materials
81
6-1
FEECON Predicted Horizontal Deflections
104
6-2
FEECON Predicted Initial Settlements
105
6-3
FEECON Settlement Correction Factors
and Corrected Initial Settlements
106
6-4
Change in FEECON Vertical and Horizontal Stresses
107
6-5
Change in FEECON Principal Total Stress
108
6-6
Change in FEECON Octrahedral Total
Stress
109
6-7
FEECON Predicted Initial Excess Head
110
7-1
Pore Pressure Dissipation at
137
7-2
Average Degree of Consolidation on
138
7-3
Incremental Consolidation (Pore Pres-
139
, Sta.246
sure)
7-4
Final Consolidation Settlement (-D)
With Corrected Laboratory RR and CR
140
7-5
Final Consolidation Settlement
(Modified Skempton-Bjerrum) With Corrected Laboratory RR and CR
141
7-6
Field RR Values
142
7-7
Field CR Values
143
8
7-8
Final Consolidation Settlement (1-D)
With Field RR and CR
Page
7-9
Final Consolidation Settlement
(Modified Skempton-Bjerrum) With
Field RR and CR
145
7-10
Composite Final Consolidation
Settlement With Field RR and CR
146
7-11
Incremental
147
7-12
Computed Field Cv's - Full Thickness
148
7-13
Calculation of Field Cv2 From Pore
Pressure, Full Clay Thickness
149
Calculation
150
7-14
Consolidation
(Settlement)
of Field Cv2 from
Settlement, Full Clay Thickness
7-15
7-16
Computed Field Cv's - Reduced
Thickness
151
Calculation
152
of Field Cv2
(Pore
Pressure) Reduced Clay Thickness
7-17
Calculation
fo Field Cv 2
Reduced Clay Thickness
9
(Settlement)
153
LIST OF FIGURES
Page
2-1
Site Location Plan
24
2-2
Nearest Borings
25
2-3
Average
2-4
Atterberg Limits and Stress History
28
2-5
Compression and Consolidation Data
29
2-6
Su/uvc vs. OCR from Triaxial Tests
30
Profile
Sta. 246
27
on Sta. 246 Samples
2-7
Undrained
3-1
Construction History
34
3-2
Embankment
35
3-3
Embankment Profile and Instrumentation,
Shear Strength,
Sta. 246
Density
31
36
Sta. 245
3-4
Embankment Profile and Instrumentation,
37
Sta. 246
47
4-1
Lateral Distribution of Measured
Settlement
4-2
Measured
Settlement,
Sta. 245
48
4-3
Measured
Settlement,
Sta. 246,V
49
4-4
Measured
Settlement,
Sta. 246,90'R
50
4-5
Differential
4-6
Differential
4-7
Excess Head,
Sta. 245
53
4-8
Excess Head,
Sta. 246,
54
4-9
Excess Head,
Sta. 246, 30'R
55
4-10
Excess Head,
Sta. 246, 30'L
10
55a
Settlement,
Settlement,
Sta. 246,
Sta. 246, 90'R
51
52
Page
4-11
Excess Head, Sta. 246, 60'R
56
4-12
Excess Head, Sta. 246, 95'R
57
4-13
Excess Head, Sta.246,
58
4-14
Excess Head, Sta. 246, 225'R
59
4-15
Horizontal Deflection vs. Elevation
60
4-16
I-3 Movement,
Sta. 246, 45'R
61
4-17
I-4 Movement,
Sta. 246, 9'R
63
4-18
I-5 Movement,
Sta. 246, 160'R
65
4-19
I-6 Movement,
Sta. 246, 225'R
67
5-1
Non-linear Stress-Strain Model by
Incremental Method
82
5-2
Finite Element Mesh
83
5-3
Adjusted FEECON Unit Weights
84
5-4
Hyperbolic Transformation of CKoUDSS
Test Data
85
5-5
Resedimented Clay, CKoUDSS Data, OCR = 1
86
5-6
Resedimented
CKoUDSS Data, OCR = 2
87
5-7
Resedimented Clay, CKoUDSS
Data, OCR = 4
88
5-8
Resedimented Clay,
CKoUDSS Data, OCR = 8
89
5-9
Normalized Shear Modulus and Rf Factor,
Resedimented Clay
91
5-10
Shear Strength of Resedimented Boston
Blue Clay
92
5-11
Strength Anisotropy of Boston Blue Clay
93
5-12
Ko for Boston Blue Clay
94
6-1
Finite Element Yielding
111
6-2
Predicted and Measured Horizontal Deflection 1 1 2
Clay,
(End of Construction)
11
160'R
Sta. 246
6-3
Predicted and Measured Initial
Settlement
Page
114
(CD 620) Sta. 246
6-4
Predicted and Measured Initial
Settlement (CD 329) Sta. 245
115
6-5
Corrected Initial Settlement
Predictions
116
Drained Vertical Stresses in Clay
117
6-6
Sta. 246,
6-7
Initial Excess Pore Pressures Under
Embankment Top, C and 30'R
119
6-8
Initial Excess Pore Pressures Under
Embankment Slope 60'R and 95'R
120
6-9
Initial Excess Pore Pressures Outside
Embankment Toe, 160'R and 225'R
121
6-10
Skempton's and Henkle's Parameters
122
7-1
Pore Pressure
Centerline
Dissipation Beneath
154
7-2
Centerline Consolidation
155
Differential Consolidation Settlement,
156
7-3
Sta. 246,
157
7-4
Differential Consolidation Settlement,
Sta. 246, 90'R
7-5
Predicted Final Consolidation Settlement, 158
7-6
Centerline Consolidation (Settlement)
159
7-7
2-D Consolidation, Permeable Top and
Base
160
7-8
2 - D Consolidation, Impermeable Top
and Base
161
7-9
Predicted Consolidation Settlement,
162
Sta. 246, C
12
1.
1.1
INTRODUCTION
BACKGROUND
In August 1967 construction began on an embankment for
Interstate highway
-95 across the Revere-Saugus tidal marsh
northeast of Boston, Massachusetts. This involved a large embankment, 25 to 40 feet high and 2.4 miles long, constructed
over a 40 to 160 feet thick deposit of medium to stiff clay,
known as the Boston BlueClay
(BBC).
The embankment design called for staged construction
with a surcharge to minimize post-pavement settlements. Because of uncertainties in the amount and rate of settlement,
and also end of construction stability, instrumentation was
installed at numerous stations along the embankment. One station in particular, Station 246 + 00, known as the M.I.T. Massachusetts Department of Public Works (MIT-MDPW) Test Section, was heavily instrumented. The instrumentation included
settlement platforms, settlement rods, peizometers, slope indicators and total stress cells. Considerable performance
data have been collected during the seven years since the
start of construction.
1.2
PURPOSE
Various researchers have dealt with different aspects
of the embankment performance. However most of the analyses
13
to date have dealt with perfo -mance during construction (D'Appolonia,
et. al., 1971).
In
tddition, Recker et. al.,
(1973)
and Lambe (1973) have discuss ~d the performance after construction
in a general way, b it no analyses of the data were
made.
One purpose of this re )ort is to present in a readily
usable format all data pertai Ling to the performance
of the
foundation clay at the test s !ction. This will simplify and
encourage further analyses
o
the field data. In addition,
the M.I.T. finite-element pro rran FEECON was used to make an
after-the-fact prediction of Lndrained performance. This permitted an evaluation of this )rograms's applicability to
plane strain loading in Bosto
Blue Clay. It also allowed an
evaluation of the input param !ters used for FEECON. Finally,
and of main interest, an anal 7sis of the consolidation behavior beneath
the embankment
ce Lterline was performed. Compres-
sion parameters and the coeff cients of consolidation for the
in-situ plane strain condition qeredetermined from the field data. Measured and predicted cons )lidation settlements were alsQ
compared.
1.3
SCOPE
Chapter
2 summarizes
the Test Section.
t Le initial in-situ conditions at
The soil pr ifile, boring data, and test re-
sults for samples obtained nea
14
the Test Section are included.
Chapter 3 summarizes the construction history of the Test
Section, including embankment construction and instrument installation. Field data for the foundation clay, including
settlement, inclinometer and piezometer data, are presented
in Chapter
4. In addition, problems encountered with data
collection are discussed. The application of the FEECON program is presented in Chapter 5. This section includes a discussion of the appropriate stress-strain parameters based on
model footing tests, and lab stress-strain-strength data. In
Chapter 6, the FEECON predictions arecompared with other predictions andmeasured data. Chapter 7 presents the consolidation analysis beneath the centerline, based on both pore pressure and settlement data. Various methods of back figuring
field values
for the coefficient of consolidation are dis-
cussed and the results compared to field data. Conclusions
and recommendations for further study are given in Chapter 8.
15
2.
2.1
SUBSURFACE CONDITIONS AND SOIL PROPERTIES
GENERAL GEOLOGY
The following discussion is based on D.E. Reed's
summary of the geology of the Boston area (Reed, 1971). Bedrock at the site is the Cambridge Argillite, which has been
subjected to varying degrees of alteration and weathering.
Diabase dikes and sills, from less than a foot to hundreds of
feet in thickness, are common in the bedrock of the Boston
area. The lowest soil unit is usually till, a heterogeneous
mixture of soil sizes deposited beneath the ice sheet. The
till is frequently covered by outwash sand and gravels, deposited by meltwater from the ice front during retreat of the
ice sheet 14,000 to 15,000 years ago.
Depression of the bedrock due to the weight of the
glacier, and rising sea level due to melting of the retreating ice sheet, formed the Boston Basin - an innundated trough.
Vast quantities of silt and clay size particles settled out
of suspension, forming the marine illitic clay called the
Boston Blue Clay. In places, the thickness of this deposit
exceeds
200 feet.
Sea level lowering associated with a minor readvance
of the ice 12,000 years ago
(the Lexington Substagei
re-
sulted in desication and weathering of the top of the clay.
17
In addition, meltwater streams often covered the clay surface with outwash sands and gravels. The uppermost layer at
the site is peat, which has accumulated since a relative stabilization of the sea level some 2,000 to 3,000 years ago.
2.2
SOIL PROFILE
Figure 2-1 shows boring location and the general site
plan. Available boring data are reproduced in Figure 2-2.
The assumed soil profile is shown in Figure 2-3. This profile
is based not only on boring data, but also on instrument elevations and installation records.
The assumed average profile at Sta. 246 show the natural ground
surface
(top of the Peat) at elevation
of +5 (U.S.G.S.). The groundwater
elevation
(El.)
is +2.5. The sur-
face of the gray shale bedrock (Cambridge Argillite) is at a
depth of 168 ft. or El. -163. This is overlain
by 8 ft. of
silty, clayey sand and gravel which is glacial Till. Standard
Penetration Test values (N values) in the Till range from 24
to 171 blows/ft.
Above
the till, between El. -10 and -145,
is
135 feet of gray CL silty Clay (Boston Blue Clay). Lenses and
thin layers of silt or very fine silty sand occur within the
top 30 ft. of clay, but apparently are discontinuous. The top
10 ft. of clay is a medium-stiff
desiccated
crust with N = 10,
and a bouyant weight of 59 pcf (Guertin, 1967). Below this,
the clay is "soft"
(N = 2-4) with a lower bouyant
18
weight
of
52 pcf (Guertin, 1967). However, fieldvane values for undrained shear strength (su) of 700 - 1200 psf indicate medium
to stiff consistency. Immediately above the clay, between
El. 0 and -10, is 10 ft. of fine to medium poorlygraded
sand.
N values average about 20 blows/ft. The top layer of very
soft Peat and Organic
2.3
Silt has N values of 0 to 2.
2.3.1
SOIL PROPERTIES
Index Properties
Figure 2-4 shows the results of Atterberg Limit Tests
performed on samples from borings close to Sta. 246. Only
data from D22 and D24
(Storch, 1965) and MIT - P11
(Guertin,
1967) are plotted. The average values are: plastic limit,
22%; liquid limit, 44.4%; plasticity
index, 22.4%; natural
water content, 40.6%. There is consistent increase in natural water content with depth to El. -45, where it becomes
roughly constant with depth. There is a similar, but less
well defined relation between liquidity index and depth.
2.3.2
Stress History
The stress history, also based on tests from samples
from the three nearest borings, is depicted in
The upper
Figure 2-4.
60 ft. of clay, between El. -10 and -70, is over-
consolidated with an overconsolidation ratio (OCR) of about
19
11 at the top. The data from the MIT - Pll boring agrees
very closely with the initial vertical effective stress (vo).
The commonly-encountered correlation between liquidity index
and stress history (LI > 100% for OCR = 1) does not occur.
However, the boundary between over- and normally consolidated
clay of El. -70 is supported by field vane and UC and UUC
data.
2.3.3
Compression and Consolidation Data
Figure 2-5 summarizes the compression and consolida-
tion data from Storch (D22, D24) and M.I.T. (MIT - Pll) oedometer testing. Values of the recompression ratio (RR
lCr )
1+eo
show a great deal of scatter, but the average RR is 0.024.
The virgin compression ratio (CR = Ccl+e
) increases roughly
1+eo
linearly from the top of the clay to El. -70, where it becomes constant with increasing depth. Average CR lab values
are 0.15 from El. -10 to -40, 0.19 between
El. -40 and -70,
and 0.21 below El. -70.
Values for
the rate
of secondary compression (Ca)
at the insitu stress are depicted
as plots
of C
Aev
(Ca= Alog t )
versus depth. These values are for MIT - Pll data only. C
is
a stress-dependent parameter, and exhibits expected variation
with depth (Ladd, 1971). In the over-consolidated region, Ca
increases with increasing depth (as the ratio of initial to
maximum past vertical effective stress
20
vo/Ovm approaches 1)
and is roughly constant with depth in the normally consolidated zone
Coefficients of consolidation (Cv) versus depth, at
the insitu stress, are also shown. In addition, estimated Cv
values at the final effective vertical stresses,
vF (from
the program FEECON) are shown. These were taken from Guertin's
(1967) plots of Cv versus
vF/ vm.
the basis of the ratio
vc, on
It is reasonable to assume that the effective Cv ver-
sus depth is represented by the average of both sets of Cv
values. This approach results in the following average values:
Cvl
(above El. -70 for OCR > 1) = 0.27 ft 2 /DAY,
El. -70, for OCR = 1) = 0.09 ft 2 /DAY,
2.3.4
Cv2
(below
Cvl/Cv2 = 3.0.
Undrained Shear Strength
Based on comsolidated-undrained triaxial compression
test, CIUC (undrained shear after isotropic consolidation) and
CKoUC
(undrained shear after K
consolidation), on samples
form MIT - Pll, the normalized Su (Su/avo) vs. OCR relation
is plotted in Figure 2-6. These tests were performed according
to the SHANSEP procedures (Ladd, 1971). They were consolidated
to 2-3 times
vm and rebounded to
priate OCR, followed
av which gave the appro-
by shear.
Figure 2-7 showsvarious Su vs. depth relationships
relationship
of Figure
from field and lab data.
The strength
2-6 was used to compute
Su vs. depth at the Test Section
21
(solid line in Figure 2-7).
In a similar
fashion,
Su for
plane strain active and passive (PSA and PSP) and direct simple shear (DSS) conditions are shown as dashed lines. However,
these values are based on tests of laboratory prepared samples
of resedimented Boston Blue Clay. These techniques and data
are extensively covered in Kinner and Ladd (1970) and Ladd
and Edgers (1972) among others. Unconfined (UC) and Unconsolidated-Undrained (UUC) test results from all three borings
are also presented. Finally, Geonor Field Vane (FV) test results are shown for two tests performed prior to construction
at Sta. 244 + 85.
The strength data shown in Figure 2-7 lead to the following observations:
(1)
There is a significant difference between Su
with the major compressive stress in the vertical direction (SuV from CKoUPSA data) and the horizontal direction (SuH from CKoUPSP data). This indicates the
importance of accounting for strength anisotropy.
(2)
U and UU tests give much more scattered results
and lower strengths than other test methods.
(3)
The FV data are quite consistent below El. -50.
Scattered data above this level could be due to sand
or silt lenses or varying avm due to varying degrees
ofdesiccation. FV data coincides closely with SH
CKoUPSP tests in the NC zone.
22
from
(4)
CIUC and CKoUC test data, the most sophisticated
data an engineering firm may have, may be slightly conservative or very unsafe, depending upon location. Beneath the centerline where olF is vertical, it may be
safe
, but away from the centerline, as O1F tends
toward the horizontal, this type of data is unsafe.
23
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3.
3.1
CONSTRUCTION HISTORY
GENERAL
Throughout this report, reference is made to time in
Construction Days (CD). The date 1 Sept. 1967 was designated
as CD 1. It roughly coincides with the completion of the earliest instrument installations. A Date - Construction Day
conversion chart is given in Appendix A-1.
3.2
EMBANKMENT CONSTRUCTION
The final design
embankment grade for the pavement is
El. +18 ft at Sta. 246 (13 ft. above natural
grade).
To mini-
mize post-contruction settlements, the embankment was preloaded with a surcharge
filling.
(see Figure
to El. +40 ft. in three
stages of
3-1).
Stage lconsisted of excavation of the 5 ft. peat layer
followed by replacement with fill (probably end-dumped to
El. +5) and continued filling to the Stage 1 El. of +9. The
excavation and replacement to original grade ocurred during
Dec. 1-7 1967 (CD 92-98). The stage 1 El.
on 1 Jan 1968
of +9 was reached
(CD 123).
Except for minor construction operations (installation
of instruments and access tunnel), no further filling ocurred
for almost seven months. Stage 2, which included the placement
of about 70% of the final fill height, began on 24 June 1968
32
(CD 298). This stage continued without interruption until the
fill reached El. +36 of 4 Dec. 1968 (CD 461).
Stage
3, consisted
of the placement
The final
stage,
of the final 4 ft. of
fill between 20 April and 12 May 1969 (CD 598-620). Embankment construction is tabulated in Appendix A-2.
The fill material consisted of a well-graded fine to
coarse sand with some fine to medium gravel (SW). Results of
field unit weight tests on various lifts above El. +22 are
given in Figure 3-2 (Wolfskill and Soydemir, 1971). The fill
was compacted with rubber-tired rollers above E
+5. Based
on these data, the average total unit weight is 119 pcf.
Since values as low as 102 pcf occur
in the compacted
fill, a
value of 100 pcf was chosen for the dumped fill between El. 0
and +5.
3.3
INSTRUMENTATION
Figures 3-3 and 3-4 show the embankment profiles and
instrumentation at Sta. 245 and Sta. 246. Although the details
of the-instrumentation are covered in Wolfskill and Soydemir
(1971), a brief
summary is provided
here.
There are two groups of instruments. The major group,
the "construction instruments", was installed at the Test Section after Stage 1 (El. +9) was completed. A minor group, the
"preconstruction instruments", was installed prior to any construction at Sta. 245. These instruments were to provide per33
formance data for the 1st Stage of construction.
Most of the Test Section instruments are accessible
from a tunnel beneath the right (East) side of the embankment.
This was to provide protection from vandals and weather. The
time of initial readings of the instruments are shown in Figure 3-1. Instrumentation is tabulated in Appendix A-1.
At Sta, 245 the settlement instrumentation consisted
of six Borros Points, one of which was installed in the till
and used as a temporary benchmark. The only other instruments
were six M206 single-tube hydraulic piezomenters.
At Sta. 246, all M.I.T.-monitored instrumentation was
located between 30'L and 225'R. There were a few MDPW-monitored instruments located further to the left of the centerline. the M.I.T. settlement devices included four settlement
platforms at the top of the natural sand. Additionally, deep
settlement data were provided by twelve settlement rods in
the clay. These consisted of cased 1" o.d. pipes with round
2 1/2" o.d. plates welded 18" above the tips. A permanent
cased benchmark was installed in the till.
M.I.T. pore pressure data was provided by 38 piezometers
at various offsets and elevations in the clay and till. In addition, 5 porous-point well points were installed in the bottom of the natural sand. Five slope indicators (inclinometers)
were also installed beneath the right side of the embankment.
Finally, three total-stress cell clusters were installed within
the embankment at El. +17.
33a
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4.
4.1
EMBANKMENT PERFORMANCE
GENERAL
Most of the data pertinant to the performance of the
foundation clay are presented and utilized in this report.
The most important data excluded are the total stress cell measurements for the embankment interior(El. +17). In addition,
data from the state monitored instruments were not analyzed.
4.2
SETTLEMENT
Figure 4-1 shows the measured settlement at the top of
the sand and the top of the clay beneath the right side of
the embankment. Settlements are shown for CD 620 (end of construction) and several times afterward, and are dish-shaped,
indicating elastic loading.
Total measured settlements and fill elevation are
plotted versus time (log scale) for Sta. 245 and Sta. 246 in
Figures 4-2 to 4-4. At Sta 245, readings were not taken during
the entire construction period. They appear very erratic because of the small scale. However, the actual variations are
not excessive, being
nly + 0.02 ft. It is not clear from
Sta. 245 data if settlement of the clay due to Stage 1 consolidation was complete prior to Stage 2 filling.
Due to improper installation of some cased settlement
rods at Sta. 246, some of the deep settlement data became
38
unusable. The specifications originally called for a minimum
vertical distance of 0.5 ft. between the bottom of the casing
and the supporting plate on the settlement rod. As actually
installed this distance varied greatly - between 0.17 and
1.27 ft. A combination of insufficient clearance and large
settlements invalidated the data for four settlement rods:
SR4 and SR6 (centerline), and SR10 and SRll (90' R).
Possible corrected settlements for these rods are
shown in Figures 4-2 and 4-3 as dashed lines. Unfortunately,
the casing elevations were only monitored twice: upon installation and 8 November 1973 (asanoutcome of this study). As a
result, the times when casing settlement actually invalidated
the data are unkown. Figures 4-2 and 4-3 indicated that this
occurred before the following times: SR4, CD1200; SR6, CD1200;
SR10, 1300; Srll, CDllO0.
An attempt was made to determine the earliest day when
casing settlement invalidated the four deep settlement points.
It was assumed that the ratio of casing settlement (based on
top elevation and assuming no compression) and the settlement
of the nearest platforms was a constant between the installation date and 8 Nov. 1973. This allowed an estimate of casing
settlement versus time. The earliest invalid day is the day
on which the casing settlement exceeded the rod settlement
by an amount equal to the initial clearance. The resulting
earliest invalid days are: SR4, CD382; SR6, CD860; SR10,
39
CD1470; SRll, CD1037.
Although four rods produced invalid total settlements,
seven sets of differential settlement data were
invalidated.
Casing drag on one rod will result in a reduced apparent differential settlement (or even heave) in the overlying layer,
and increased differential settlment in the underlying layer.
The measure differential settlements at Sta. 246 are shown in
Figures 4-5 and 4-6. The curves plotted represent the measured differential settlements between adjacent settlement
rods. For example, SR1-2 is the difference in total measured
settlement between SR1 and SR2. Due to casing drag the following remarks can be made about the apparent differential settlements
in Figures
4-5 and 4-6:
SR 3-4
too low
SR 4-5
too high
SR 5-6
too low (apparent heave)
SR 6
too high
SR 9-10
too low
SR 10-11
probably too low
SR 11-12
too high
The above errors in measured settlements greatly complicate analysis of the settlement data. All that is certian
is that at the centerline, final consolidation settlement has
been reached above SR3 (El. -43). Similarly, at 90 ft. right,
finalcoDnsolidationsettlement has been reached above SR8
40
(El. -18) and it is probably
close between SR8 and SR9
(El.
-43). In Chapter 7, only differential settlement data of
known validity is used in consolidation analysis, and the data
is plotted
in Figures
7-3 and 7-4.
Although the behavior of the natural sand is not an
object of this study, one aspect is interesting. The differential settlement between SP3 and SR7, 90'R, indicates that the
natural sand dilates after an initial compression. This occurs
after
the fill height
is a maximum
above SP3
(El. +16),
so
that further filling is moving laterally away from SP3.
All settlement data are tabulated in Appendix A-4. The
vertical distribution ofsettlement vs. time is discussed in
Chapter
4.3
7.
PORE PRESSURES
Measured excess pore pressures in feet of water are
plotted vs. time
(log scale) for Sta. 245 and Sta. 246 in Fig-
ures 4-7 to 4-14. Since readings were generally made every day,
only representative data points (except at Sta. 245) are
plotted. The measure excess pressure represents the difference
between the measure total head elevation at a piezometer and
the "hydrostatic" piezometricelevation at the piezometer.
At Sta. 245, the hydrostatic elevation was assumed
equal to the total head elevation indicated by the piezometers
in the natural
sand. At Sta. 246, the hydrostatic
41
elevation
was assumed equal to the water elevation in the nearest well
in natural sand. It must be noted that inreality this is not
correct. The data for the centerline piezometers indicates
that the piezometric water elevations (PWE) in the sand
above the clay and the till below are not equal. As a result
of an apparentartesian condition, the PWE is increased by five
ft. in the till. For simplicity, however, the PWE in the sand
was used for all plots. As a result, the till appears to have
an excess pore pressure of 5 ft. rather than it's true value
of zero
(Figure 4-8, P
11)
Readings at Sta. 245 were not made for the entire construction period. All of these piezometers consisted of Geonor
M206, single-lead, bronze-point hydraulic piezometers. These
appear
to be fairly reliable, although they are slightly
more erratic (+ 2 ft. of head) than the more sophisticated
types at Sta. 246. The limited data available for Sta. 245
indicate that at the center of the clay there was essentially
no pore pressure dissipation between Stages 1 and 2.
The M.I.T. piezometers at Sta. 246 consisted of two
types: Geonor vibrating wire electric piezometers at 30'L,
and two-lead, porous-plastic point hydraulic piezometers elsewhere. Although a reliability analysis has been performed by
Wolfskill and Soydemir (1971), an up-dated review is advisable.
At Sta. 246, the electric piezometers produced consistent
data (+ 0.5 ft. of head) but only short-term reliability (only
33%, or 2 of 6, were operating at CD 2053). The hydraulic
42
type generally showed the same degree
of consistency, but
better long-term reliability (52%, or 16 of 31 installed were
operating reliably at CD 2053).
For comparison with predicted values (using FEECON),
maximum measured"excess heads were taken as the peak values
at the end of Stage 2 filling plus the increase due to Stage
3 filling. It was necessary to estimate excess heads for the
inoperative piezometers. For these, an estimated excess head
vs. time curve was drawn based on operating piezometers
whose locations indicated a similar performance. The results
are shown as dashed lines in the Figures.
The maximum"measured" excess heads are tabulated in
Table 4-1. In addition, representative piezometer and well
data are tabulated in Appendix A-5. The vertical distribution
of excess pressure vs. time is discussed in Chapter 7.
4.4
HORIZONTAL DEFLECTION
No inclinometers were installed with the preconstruc-
tion instrumentation (Sta. 245), but 6 were installed at Sta.
246. Of these, 5 inclinometers between the centerline and
225'R were read by M.I.T.. The instruments were 3 in. in diameter grooved aluminium casings in 5 ft. sections connected
by flexible couplings permitting 6 in. of vertical movement.
Figure 4-15 presents standard plots of horizontal deflection vs. elevation at the end of construction (CD 620) and
43
several times afterward. It was desired to study the horizontal deflections versus time for all depthswithin the clay.
Normal methods of portraying inclinometer data do not permit
this where deflections are relatively small (as at the MITMDPW Test Section). Therefore, the horizontal deflection at
various elevations are plotted vs. time (log scale) in Figures 4-16 to 4-19. Data from the centerline inclinometer
(I-2) are not plotted. It indicated a maximum of only 0.8
inches at the end of loading, indicating slightly asymmetrical loading
The inclinometer data are somewhat erratic, but not
abnormally so (+ 0.5 in.) in view of the great depth of clay
and relatively small deflections. The most erratic data, between CD600 and CD900 and after CD1700, are probably due to
the interchangeable use of several different Wilson torpedoes.
Converely, the most consistent data, between CD1300 and CD
1700, were obtained with the M.I.T. automatic recording instrument
"Beaver"
(Bromwell et. al., 1971).
The data show that the greatest rate of deflection occurs during filling operations. Atthe end of the filling
there is a slower but constant rate of outward creep (related
to log time). Generally, after the end of filling the slopes
of the deflection-time
curvesbetween
El. -10 and El. -70
(OC
zone) are approximately parallel. This indicates that there
is no creep in this zone, and all outward creep is due to the
44
normally-consolidated clay between El. -70 and El. -145. A
study of the differential creep (analogous to the differential settlement) was not undertaken for this report.
The fact that there is no consistently varying break
in the deflection-time curve (with location in the clay)
suports the assumption that horizontal deflection is not a
consolidation-related phenomena. The maximum deflection occurs
beneath the embankment slope ( I-4, 95'R.) rather than at the
toe as one might expect.
The inclinometer data are tabulated in Appendix A-6.
Elevations were based on the average of the casing elevations
at the first and latest readings.
4.5
INTERNAL EMBANKMENT STRESS
When the fill elevation reached +20 ft., three clusters
of stress cells were installed in circular pits excavated to
El. +17. The clusters were placed at the centerline, 30 ft. R,
and 60 ft. R. Each cluster consisted of three Geonor P-100 vibrating wire total stress cells. The datails of this operation and the results are presented in Wolfskill and Soydemir
(1971). Excellent agreement with embankment stresses predicted
by Perloff's
method
(Perloff, et. al., 1970) was achieved
with the fill El. at +40. The data agree with total vertical
stress based on the weight of the overlying fill for the centerline and 60'R and is 85% of that value at 30'.
45
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-X85
5.
PARAMETERS FOR UNDRAINED DEFORMATION AND
STRESS FINITE ELEMENT ANALYSIS
5.1
GENERAL
The finite element program FEECON, described in de-
tail elsewhere, (Simon, 1972, Simon et. al., 1972) was used
to analyze the undrained behavior of the embankment. This
program is exceptionally versatile. It permits the use of
several stress-strain relationships including
bilinear,
hyperbolic, axial stress-strain or shear stress-strain. Anisotropic strengths can be specified for cohesive materials,
and initial shear stress can be accounted for.
For analysis of the Test Section, hyperbolic stressstrain relationships were used for both granular and cohesive soils.
FEECON uses the incremental method in hyperbol-
ic stress-strain models as shown in Figure 5-1. At the beginning of each load increment, the modulus is set equal to the
value of the tangent to the true stress-strain curve corresponding to the existing stress level. It is therefore necessary to use small load increments. Nine increments were used
for the Test Section analysis. The finite element mesh is
shown in Figure 5-2.
5.2
GRANULAR SOILS
A drained hyperbolic axial stress-strain relation was
69
used for the granular soils (fill and natural sand). This
relation is based on work done by Duncan and his associates
(at the University of California, Berkeley) with Kondner's
original suggestions (Simon, 1972). This stress-strain model
is discussed in detail elsewhere (Duncan and Chang, 1970,
Simon 1972).
The initial Young's modulus, Ei, is determined empirically
and is related to the minor principal stress by
Janbu's formula:
Ei =
where
K
is
K
(
n )
n
a dimensionless empirical modulus number, Pa is
the atmospheric pressure in proper units,
3 is the minor
principal stress and n is a dimensionless empirical exponent.
In addition, the tangent Young's modulus is related
to the principal
Et
stresses
=
[ 1
by:
Rf (1-
sinf)(1-73)
2 c cos
where c and
+ 23
2 Ei
sinT
are the Mohr-Coulomb strength parameters; Rf
is the failure ratio, equal to the ratio between the compressive strength (
-
3 )F
and asymptotic stress difference
for the hyperbolic stress-strain curve;
and W3 are the
major and minor principal stresses, respectively.
FEECON also accounts for the stress and strain dependency of the initial and tangent Poisson's rations (viandvt).
The initial Poisson's ratio is represented by the equation:
vi = G - F log 1 0
70
( 3 /Pa)
where G is the value of vi when
equals
3
a, and F is an
empirical constant expressing the dependence of v on confining stress.
The tangent Poisson's ratio is given by:
V.
Vt
Vt = =
[1where d expresses
]2
c 3)
d(l Et
the strain dependency
of v, and -d is the
slope of the line on a transformed hyperbolic plot of Or/
vs
r(
a= axial
strain,
r=
radial
a
strain).
Since the behavior of the embankment material was not an
objective of this report, the stress and strain dependency
of Poisson's ration was ignored. Parameters F and d were
chosen
to be zero, so that
=
i
= G where G is a con-
stant value. The constant Poisson's ratio G was assigned
soils. The empirical
para-
the value
of 0.4 for all granular
meters
n, and Rf which determine E i and Et were chosen
K,
from Mitchell and Gardner's (1971) suggested relationships.
These empirical relations are tabulated in Table 5-1.
For all granular soils the cohesion intercept (c)
was assumed to be zero. Friction angles () and coefficients
of lateral earth pressure at rest (Ko) were estimated, based
on soil type and density
Due to a discrepancy between true ground water elevation (+2.5) and that assumed for FEECON (El. 0), the
actual unit
eights for some soils were adjusted. The pro71
cedure is shown in Figure 5-3. The unit weights of the peat
natural sand, dumped fill and compacted fill to El. +9 were
changed so that the vertical effective stress at the top of
the natural sand was equal for both water level conditions.
All FEECON parameters used for the granular soils are tabuin Table 5-2.
lated
5.3
COHESIVE SOILS
5.3.1
General
A hyperbolic shear stress-strain relation with aniso-
tropic strengths was used for the cohesive soils (peat and
Boston Blue Clay). FEECON incorporates the fact that the iniis usually not zero (Simon, 1972). The
tial shear stress q
hyperbolic relation is written as:
Aq
=
Iq
-
q
=
/[
/Gi
+(Rf/Aqf)y]
in which 1/Gi and Rf/Aqf are analogous to the a and b parameters of Duncan and Chang (1970). Gi is the initial shear
modulus; Aqf is the change in shear stress to cause undrained failure for the appropriate stress condition, whether
DSS, PSA or PSP conditions; and Rf is Aqf/Aq for
=
. All
cohesive input parameters are tabulated in Table 5-3.
5.3.2
Hyperbolic parameters
D'Appolonia, et. al. (1971), Foott and Ladd (1973) and
Simon, et. al. (1972) have shown that the undrained modulus
72
obtained from CKoUDSS tests is a reasonable estimate of the
in-situ undrained modulus for many clays. Therefore, this
approach was used in analysis for the Test Section. Since
CKoUDSS data was not available for the actual undisturbed
samples from the Test Section, data from laboratory prepared sample of Boston Blue Clay were used (Ladd and Edgers,
1972).
Figure 5-4 indicates the method used to determine
the parameters (Gi, Rf) defining the hyperbolic stress-strain
curve. Figures 5-5 to 5-8 present the CKoUDSS data (from
Ladd and Edgers, 1972) for resedimented BBC and the derived
hyperbolic
curves. For the DSS stress
that Thmax = Aqf = su
system, it is assumed
.
When the hyperbolic plot of the test data is made, as
y/ (Th/su)% versus
intercept
(at
%, Rf is the slope of the line and the
= 0%) is su/ G i.
The normalized
initial
shear modulus (Gi/avc) can then be determined fromthe normalized undrained shear strength, and G1 computed for each
layer as follows:
Gi/su = 1/intercept
Gi/Fvc = Gi/su
su/O-avc
Gi = Gi/Fvc X avc
where su = Thmax for DSS case, and Oavc= vertical effective
consolidation stress.
73
At all OCR, hyperbolic plots of CKoUDSS data for resedimented BBC curve downward at low values of shear strain.
The curvature is very slight or non-existent at OCR = 1,
but increases with OCR and creates problems in choosing the
best intercept (su/Gi). Additionally, the plots curve upward at shear strains greater than the failure shear strain.
This is due to the strain-softening nature of the clay,
which is apparent at allvalues of OCR.
The effect of low-strain curvature on hyperbolic
parameters is depicted in Figure 5-4, which shows (schematically) both a normalized shear stress-strain curve and the
equivalent
hyperbolic plot. Various straight-line approxi-
mations of the hyperbolic plot are shown, along with their
equivalent normalized stress-strain plots, computed from
the DSS relation
Th
Fvc
(Simon et. al. 1974)
=
Y(%)
(l10)VC~
Gi
+
Rf y(%)
Thmax/Fvc
It is apparent that the straight line which best fits
the hyperbolic plot is the equivalent of the curve which most
closely approximates the normalized stress-strain data over
the full range of strain up to failure. Such "best fit" approximations to the hyperbolic plots appear to be the best
method of determining hyperbolic parameters from lab data.
74
5.3.3
Shear Modulus
Figure 5-9 indicates the normalized initial shear mo-
dulus (Gi/Ovc) and Rf values used in the FEECON analysis.
Values used for OCR = 1 to 4 were those recommended by Simon et.al. (1974), and were based on comparisons of FEECON
analyses with model footing tests (Simon, 1972, Kinner and
Ladd, 1970). Values used for OCR = 8 are "best fit" values
from hyperbolic plots of CKoUDSS data on the resedimented
clay. The values used in this analysis are shown as dashed
lines.
Initial re-evaluation of model test footing results
indicates that chosen values of Gi and Rf at OCR = 2 and 4
are somewhat too low and too high, respectively. This is in
agreement with "best fit" values from CKoUDSS data at those
OCR's. Additionally, re-evaluation also shows that for NC
clay, there is a reduction
in Gi with increased
than a unique Gi at OCR = 1 as implied by Simon,
vc, rather
et. al.
(1974). This is also in agreement with "best fit" CKoUDSS
data. This da-tafor OCR = 1 (Figure 5-9) shows an inverse
linear relation between Gi/avc and
vc on a log scale.
This is generally the same relation porposed by Janbu (1963):
G ii =
where r is
pa(a3
Pa
h )n e
the dimensionless empirical shear modulus number.
75
Conversely, CKoUDSS data indicate no
avc dependency of
Gi/avc at OCR = 2, 4 and 8.
As a result of these observations, a set of recommended values of Gi/avc and Rf have been determined. These
are portrayed as solid lines in Figure 5-9, and account for
the effect of
vc on Gi/avc. There are several curves inter-
polated between OCR = 1 and 2 for varying
vc
at the top
of the NC clay. Within the NC clay, Gi/avc should vary linearly and inversely as the log of
vc' The recommended
values for OCR = 1 and 2 are based entirely on CKoUDSS data,
and should be considered tentative. It is obviously necessary
to compare further FEECON predictions using these values
with the model footing results.
With FEECON, it is necessary to use a small positive
value for the shear modulus after yielding, Gy. This was
chosen to be one percent of the initial modulus in all cases.
Gy = 0.01Gi
5.3.4
Poisson's Ratio
Poisson's ratio
must always be less than the theore-
tical 0.50 for the undrained case. This is due to the fact
that the term 1/(1-2v) becomes infinity in the finite element
calculations for
= 0.50. An initial Poisson's ratio vi was
chosen as 0.485. With the values chosen for Bulk modulus (K)
and Gi, the yielded Poisson's ratio becomes 0.49985.
76
5.3.5.
Bulk Modulus
The bulk modulus K is kept constant at all stress le-
vels. With an initial Poisson's ratio Vi = 0.485,
Ki
Gi
2 (1+v)
3 (1-2v)
then, Ki = K
=
0.485
= 33G i
This relation was used to determine K for all cohesive soils.
5.3.6
Undrained Shear Strength
Figure 5-10 shows the undrained shear strength para-
meters used in the FEECON analysis. It is necessary to input
the undrained strength in the vertical directions suV, and
the strength ratio Ks (Ks = suH/suV). These parameters are
based on CKoUPSA (uV) and CKoUPSP (suH) tests on resedimented BBC. Figure 5-11 shows the elliptical anisotropic
strength criteria
sed, based on Davis and Christian (1971).
The a/b ratios used are those recommended by Simon, et. al.
(1974). They describe the shape of the Davis and Christian
strength ellipse (a is the major half axis, b is the minor
half axis)
5.3.7
Initial Stress Level and Ko
The initial stress level q
q
= 0.5(1 - K)
vo
77
is expressed as:
This stress is negative for highly oversonsalidated clays,
where K
> 1. The values for Ko were chosen from R.S. Ladd
(1965) data for K
5.3.8
vs.OCR for Boston Blue Clay (Figure 5-12).
Pore Pressure Parameters
FEECON uses Henkel's equation to predict undrained
pore pressures:
Au = Aoct
+ aAToct
where a and k are Henkel's parameters, and Aaoct and Aoct
are change in octahedral normal and shear stress, respectively. Henkel's parameters, a and k, were both set equal
to zero, sothat Au = Aoct.
Other pore pressure and stress
relations were then calculated by hand.
5.3.9
Peat Parameters
Since behavior of the peat was not an object of this
study, little effort was spent indetermination of its parameters. The hyperbolic parameters were taken from CKoUDSS
tests performed on the Taylor River (Maine) peats. Other
parameters were estimated from generally observed performance
of peats in the Boston area.
78
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6.
COMPARISON OF PREDICTED AND
MEASURED UNDRAINED DEFORMATIONS AND STRESSES
6.1
YIELDING
Figure 6-1 indicates the predicted yielded zones beneath the embankment at Sta. 246. Elements whose yield factor, q/qyield, is greater than 0.95 and 0.90 are also shown.
At the full height of the embankment, NC clay and the lowest
part of OC clay
(where OCR < 1.4) is at or near yield under
the entire embankment.
This is agreater width of yielding
than that indicated by earlier finite element analyses with
a bilinear stress-strain relation (D'Appolonia et. al., 1971).
However, yielding over the full depth
of NC clay agrees
with the earlieranalyses. This is encouraging, since bilinearanalyses have been reported as more effective for predicting the performance of normally consolidated clays (Simon, 1972).
6.2
HORIZONTAL DEFLECTION
Figure 6-2 plots the measured and predicted hori-
zontal deflection
at the end of construction
(CD 620) for
the five inclinometers between the centerline and 225' R.
Since moduli were based on CKoUDSS data, it is not surprising
that the best agreement occurs in areas where DSS conditions
95
might be expected ( I-4 and I-5, 95'R and 160'R). Ladd and
Edgers (1972) show that values of E/Fvc,
for triaxial com-
pression and PSA tests are several times larger than those
from DSS tests, for NC clay.
In any event, considering that predicted deflections
commonly exceed measured deflections by a factor
of three
(Poulos, 1970), the close agreement at these location is
remarkable. Since the inclinometers were initially read at
the beginning of Stage 2 filling, they do not indicate lateral deformations due to Stage 1 filling. Therefore, in
order to provide a proper comparison, the indicated predicted deflections are incremental deflections from the
beginning of Stage 2 to the end ofconstruction. The predicted
deflections
at the end of Stage
1 and Stage
3 fil-
ling are tabulated in Table 6-1.
6.3
VERTICAL SETTLEMENT
Figures 6-3 and 6-4 compare the settlements measured
at the end of construction with predicted undrained settlements
for the centerline
and 95' R., at Sta. 246 and 245
respectively. For Sta. 245 the comparison is made at the
end of load step 3 (CD 329, fill El. +15) since measurements
were stopped shortly afterwards.
At Sta. 246, as for
the case of horizontal
deflec-
tions, measured values represent only that settlement due to
96
Stage 2 and 3 filling. Therefore, the appropriate incremental predicted settlements are shown. In order to reduce the
effects of consolidation, the measured values are the totals
of incremental settlements measured during filling only
(i.e. the settlement between the end of Stage 2 and the beginning of Stage 3 has been removed).
In the typical construction situation, rather small
increments
of fill are placed relatively
slowly.
In such a
case it is obviously impossible to accurately measure undrained settlements, since consolidation will always occur
during construction. This means that only a subjective comparison of predicted and measured settlements can be made.
All that can be stated is that the measured settlements
should be greater than predicted settlements at the end of
construction.
Measured horizontal deflections are much less affected, if at all, by consolidation
during
loading,
as dis-
cussed in Chapter 4. Therefore, predicted undrained settlements were corrected based on a comparison of predicted and
measured horizontal deflections. The precedure is shown by
example in Figure 6-5. At different elevations a correction
factor, CF, is computed from the areas to the left of the
measured and predicted deflection vs. elevation curves. The
predicted initial settlements are then multiplied by CF to
produce a corrected prediction. These latter values were
97
considered to be the actual initial settlements.
Original predicted settlements are tabulated in Table 6-2. Table 6-3 presents the computed CF values and the
resulting corrected predicted settlements for Sta. 246. For
each location, CF
was based on the nearest inclinometer. At
the centerline, it was necessary to use I-3 (45'R.) data,
since horizontal deflection was prevented there for FEECON.
The corrected initial settlements shown in Figure
6-3 suggest that at Sta. 246 there has been significant
consolidation settlement at the centerline by the end of
construction.
In addition,
at 95' R., there is a predicted
slight initial heave. This is due to filling above El. +16
moving away relative to 95' R. location. The effect is not
yet apparent for anearlier time (Figure 6-4) when filling at
CD 329 is still above that location. Additionally, heave of
the overlying sand layer was measured at 95' R. once fill
height
exceeded
El. +16
(Figure 4-5).
For comparison, initial settlement was calculated for
Sta. 246 at the centerline with the method of D'Appolonia et.
al. (1971), which might be used in practice. The average su
was taken as 1.7 ksf from CKoUDSS tests over the full length
of the clay. With Eu
1 2 00
su , b = 85', h = 145', an elastic
settlement Pe was computed from Davis and Poulos (v = 0.5)
to be 0.056 ft. Then, assuming
5
qu =
q/qu = 0.5, f = 0.5, H/B = 1.0, Pi
98
=
.1 su, average
0.073 ft.
q = 4.3ksf,
(D'Appolonia
et. al., 1971). This is less than half the corrected
prediction
at the top of the clay
FEECON
(0.20 ft.). D'Appolonia
et.al. suggest caution in using their method for non-homogeneous clays.
6.4
FOUNDATION STRESSES
Figure 6-6 shows the predicted final vertical effec-
tive stresses (drained line. For comparison,
vf) beneath the embankment center-
vf distributions for a number of me-
thods are portrayed. As one would expect, the FEECON prediction falls between the two extreme cases of one-dimensional
and Boussinesg strip load on a semi-infinite half-space.
Additionally, the FEECON prediction is somewhat lower than
the Davis & Poulos distribution for a strip on a homogeneous
elastic layer (based on Recker et. al., 1974). This is probably because FEECON can account for stress redistribution
during construction, as well as the non-homogeneous nature of the soils. All methods converge to the one-dimensional case at the top of the clay, as they should. The comparison indicates
that the FEECON predicted
vf values are rea-
sonable.
Figure 6-6 also shows the initial vertical effective
stress, based on FEECON predictions of Aav and the initial
excess pore pressure (Section 6-5). Assuming an undrained
loading, there is an initial apparent slight reduction
99
in av throughout the NC zone. However, this has probably
been eliminated by the end of construction due to some consolidation. FEECON stress:predictions are tabulated in Tables 6-4 to 6-6. These include vertical and horizontal
stresses, principal stresses and octahedral stresses.
6.5
PORE PRESSURES
Figures 6-7 to 6-9 present the maximum "measured" ex-
cess pore pressures beneath the embankment, and initial excess pore
pressures predicted by three different methods.
The maximum "measured" values are the peak values occurring
after Stage 2, plus the measured increase in pore pressure
due to Stage 3 filling. To determine these values, it was
occasionally necessary to use graphic interpolation, as discussed in Chapter 4.
All three methods use various stresses predicted by
FEECON. The methods include:
1) Au= Aoct
2)
Au=
Aav
3)
Au=
Aoct
k
+ aATOct (Henkel, 1960)
FEECON uses Henkel's equation to predict pore pres-sures. However, Henkel's a and k parameters were both input
as zero, so that FEECON pore
pressure output was generated
as Aaoct.
For the application of Henkel's method, his a parameter
100
was related to Skempton's A parameter at failure, Af
(D'Appolonia
et. al., 1971):
3 Af
a =
- 1
/2
This relation does not strictly apply outside the
yielded zone, but since so much of the foundation is near
yielding, it was chosen as a reasonable yet simple approximation.
Figure 6-10 plots both Skempton's and Henkel's parameters as a function of OCR on a log scale for the three
stress conditions: PSA, DSS, and PSP. The PSA and PSP Af
parameters are based on plane strain tests on resedimented
BBC (Ladd et. al., 1971). These curves were extrapolated to
OCR = 8, and the DSS curve was assumed to be an average of
the PSA and PSP curves. The equivalent Henkel a parameter
was then computed and plotted.
The curves in Figure 6-10 permit taking the stress
conditions into effect in a simple way. The foundation was
divided into zones where each of the three conditions (PSA,
DSS, PSP) might reasonably be expected to predominate. The
PSA zone was chosen
as that beneath the maximum height of
Fill (0 to 45' R.), DSS beneath the embankment slope (45'R.
to 140' R.), and PSP outside the toe (+ 140' R.). Pore pressures were then computedwithin each zone with the appropriate a parameter. Henkel's k parameter was assumed equal to 1.
101
Au = AGoct + aAToct
AaG
=
Aoct =
AT
A2
2
Ao 1 + Aa2 + Aa
3
Aol
+
3
A3
-2
1_
Oct
=
(Ao
- Aa
2) +
(Aal
Aa 3 ) 2 +
-
.
(Ao 2 -Aa
3)
2
Pore pressures predicted with Henkel's method are tabulated
in Table
6-7
Study of Figures 6-7 to 6-9 shows that for locations
beneath the embankment (0 to 95' R.) the modified Henkel
equation , as used in this study, most closely matches the
measured maximum pore pressures in the mid-third of the
clay (where consolidation effects are least). Beneath the
Au = A
embankment crest the relation
v
is quite good for
the NC region, but significantly high in the upper OC zone.
The change in octahedral stress, Aaoct, gives the lowest values for initial pore pressures beneath the embankment, and
apparently underestimates undrained excess pore pressures.
Considering that the Davis and Poulos Aav distribu(Figure 6-6),
tion is somewhat greater than the FEECON Av
it may be that the use of their relation for bulk stress, Ao
(equal to Aoct)
might be a fair approximation for excess
pore pressure if a linear distribution equal to A
at the
mid-plane is assumed. It should be noted that the jogs in the
red noccirn
oroZino;>
L1W4.1W--
l rotc ;t~Uc amp 1rnr'h,
--
102
1n
ra ;nsaE
V r--i
ly due to the use of a finite number of discrete elements
in the analysis.
The pore pressures predicted with the modified Henkel relation were used as the initial excess pressures in
the consolidation analysis (Chapter 7).
103
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7.1
ANALYSIS OF CONSOLIDATION BEHAVIOR
PORE PRESSURE
Figure 7-1 shows the vertical distribution of excess
pore pressure beneath the embankment centerline at different items. The predicted initial excess pore pressures and
the artesian pore pressure are also shown. Table 7-1 tabulates the pore pressure
dissipation with time. It is appar-
ent there was significant dissipation by the end of loading
(CD 620), even at the center of the clay.
In order to determine how the rate of consolidation
varied within the clay, the dissipation of pore pressure was
computed at varying times. This was done both for the entire
thickness and for several layers within the clay. The excess
pore pressure was assumed to be linearly distributed between
the piezometers. Additionally, the artesian head was assumed
to be linearly
distributed
through
the full thickness
of
clay. The average degree of consolidation, U, was computed
from the total areas to the left of the appropriate curves:
Am - Aa
U = 1 - Ai - Aa
where Am is the area left of the measured pore pressure
curve at some time (computed by summing the trapezoids), Ai
is the area left of the initial curve
ft.- ft. water),
(a constant,
Ai = 6810
and Aa is the area left of the artesian
123
curve
(also a constant,
Aa = 338 ft. - ft. water).
This me-
thod is depicted and the computations tabulated in Table 7-2.
In addition, in a similar manner, the degree of consolidation was computed for several layers within the clay.
These layers were chosen to coincide with layers for which
valid differential settlement data exist:
Layer A
El. -10 to -21.5
B
-21.5 to -43.0
C
-43.0 to -93.1
D & E
-93.1 to -145
D
-93.1 to 120.7
E
-120.7 to -145
Layer D & E was subdivided into D and E in order to take into account rapid consolidation just above the till drainage
surface.
Figure 7-2 indicates the average and incremental pore
pressure dissipation with respect to time in days (log scale)
after an instantaneous loading. The time of instantaneous
loading,
to, was taken as CD 357. This is the average
of the
times at the beginning and end of construction, since there
was no dissipation at the centerline between Stage 1 and piezometer installation (Chapter 4). The equivalent loading
ramp is shown in Figure
3-1.
The incremental degree of consolidation was computed
identically to U, except that only the areas within the layer
124
boundaries were considered.
From Figure 7-2, the average degree of consolidation
at the end of loading was already 40%, and it reached 60%
about 4.7 years after the middle of the loading period (22
August, 1968). The two uppermost layers (A and B),where the
OCR is greater than 2.5, consolidated most rapidly. Pore
pressure in the bottom layer (E) above the till also dissipated fairly quickly. Layers C and D in the interior of the
clay, however, are consolidating much more slowly and at
about the same rate. Incremental consolidation data
bulated
7.2
are ta-
in Table 7.3.
CONSOLIDATION SETTLEMENT
7.2.1
General
In order to determine the degree of consolidation
from settlement data, it is necessary to either know the final consolidation settlement, Pcf, from field data or have
an accurate prediction of it. Fortunately there are valid
settlement data for the top 33 feet of clay, which has essentially reached Pcf. These data allow the computation of
a field recompression ratio (RR). There are also valid data
which permit the computation of a field virgin compression
ratio (CR), but it is not as good due to the great thickness
of clay involved
(52 feet, from El. -93 to -145).
There are essentially two methods of predicting Pcf
125
beneath an embankment. Both approximate the field consolidation curve by two straight lines. RR is the slope of the
recompression curve from some initial stress (vo or
to the maximum stress (vm),
vi)
and CR is the slope of the vir-
gin compression curve from avm to the final stress (vf).
The first method, called a one-dimensional (-D) prediction, assumes that the change in vertical effective stress
(Aav)equals the change in the vertical total stress (Aav)
This method uses a two-dimensional estimate of avf in conjunction with the one-dimensional parameters CR and RR. The
formula
is:
Pcf =
vm
H[ RR log
+ CR log °vf ]
Ovm
avo
where
vo is the in-situ vertical effective stress.
The second method, called a modified Skempton-Bjerrum
(MSB) prediction,
is based on the Skempton-Bjerrum
(1957)
concept as modified by Ladd (1971). This method also utilizes
vf with CR and RR. However, it accounts
a two-dimensional
for non-one-dimensional conditions by assuming that Avv
equals the change in pore pressure (Au). The initial vertical effective stress (vi)
avi =
vo + Av
and 1-D compression
Ovf) is assumed.
except that
is estimated by:
- Au
from Fii to avf
The formula
(instead of from avo to
is identical
vo is replaced by
126
vi
to the 1-D method
Pcf
=
H [ RR log
m
uvi
+ CR log 27f ]
jvm
All measured consolidation settlements were derived
by subtracting the corrected FEECON initial settlements from
total measured settlements. Initial stresses (vi)
are those
predicted from FEECON analyses for final stresses and pore
pressures:
vi =
7.2.2
V
+ AGV - Au
=
avf - Au
Laboratory Compression Parameters and Predicted Pcf
Initial predictions of
cf were performed by both
1-D and MSB methods using corrected laboratory CR values.
The average laboratory value of 0.024 was used for RR
throughout the clay. To compensate for sample disturbance,
the laboratory CR values were increased by 15% (Ladd, 1971).
The resulting values for CR were: El. -10 to -40, CR = 0.173;
El. -40 to -70
, CR = 0.219; El. -70 to -145, CR= 0.242. The com-
putations and results of both analyses are shown in Tables 7-4 and
7-5, and depicted in Figure 7-5. The 1-D method predicted a-settlement of the clay of 4,26 feet,.and the MSB method
a value of
4.59 feet.
7.2.3
Field Compression Parameters
Firgures 7-3 and 7-4 show the valid measured differ127
ential consolidation settlements at the centerline and 90'R.
The upper 33 feet of clay (Layers A and B) has reached PcfThis is slightly at odds with the pore pressure data, which
indicates that layer B is at 89% of Pcf. Final consolidation
of layer A is given by SR 1-2 = 0.26 feet
and layer B Pcf
(SR 2-3) = 0.43 feet. Since only recompression
occurs
in
both these layers (Figure 6-6), the field RR values can be
easily computed by the ratio of measured to initial predicted Pcf within these layers. This procedure and the results
are tabulated in Table 7-6.
For the 1-D case,
-21.5) and 0.039
field RR's are 0.034
(El. -10 to
(El. -21.5 to -43). These values
are 50%
greater than the laboratory value. The MSB field RR's are
0.061
( -10 to -21.5) and 0.066
(-21.5 to -43). These
are
greater than the 1-D values because a much smaller change in
stress (vi
to
vf instead of avo to
vf) must cause the same
measured Pcf. For analysis the field RR's for El. -21.5 to
-43 were also assumed to be the RR's below -43.
Table 7-7 presents the calculations and results for
field CR values. The analysis is restricted to the layer subjected only to virgin compression, layer D & E, El. -93.1 to
-145. CR's were computed fromthe changes in consolidation
settlement (APcf) and vertical effective stress (Aav) for
three time increments after the end of loading from the formula:
128
CR=
Ap
H Alog av
Layer D & E was divided into three layexs, and for
each the term (H Alog
v) was evaluated. The
v was computed
by subtracting excess pore pressures at each time from
vf.
The (H Alogav ) terms were then summed for layer D & E, and
with Ac
(SR5) over a time increment, CR for that increment
was computed.
The field CR for layer D & E is increasing with time
because the increase in
v is due to pore pressure dissipa-
tion, which is occurring more and more slowly (Figures 4-8
and 7-2). However, the average of field CR values for the
three time increments was chosen between El. -93.1 and -145
(CR = 0.391). This is 86% greater than the laboratory value
of 0.21. It was assumed that field CR values vary with depth
in the same way as laboratory values. This results in
the
following field CR values (which were used in further analyses): El. -10 to -40, CR = 0.279; El. -40 to -70, CR = 0.354;
El. -70 to -145, CR = 0.391.
7.2.4
Predicted Final Consolidation Settlement
Figure 7-5 indicates the predicted Pcf at the embank-
ment centerline. Both 1-D and MSB predictions were performed
with the appropriate field RR and CR values. Comparison of
the new predicted Pcf with the measured values in layers A
129
and B indicated better agreement with the 1-D method. Therefore, a composite prediction of 1-D above El. -70 and MSB
below El. -70 was chosen as the best estimate and used in
consolidation analysis. With the field RR and CR values, the
Pcf predicted at the top of the clay are:
-D, 7.66 feet;
MSB, 7.82 feet; composite, 7.82 feet. These predictions are
tabulated in Table 7-8 to 7-10.
It is apparent that the
cf values predicted with
field RR's and CR's are much greater than the predictions
based on laboratory RR's and CR's. This is chiefly due to
the very large field CR computed for layer D & E. The increase of 86% over laboratory values seems very large to be
entirely explained by disturbance (Ladd, 1971). It may be
that increased compressibility due to artesian leaching of the
marine clay is a factor, especially in this bottom layer.
7.2.4
Consolidation
Based on the composite prediction of Pcf, the average
and incremental degreesofconsolidation
for the clay were
computed. These values are tabulated in Tables 7-2 and 7-11,
and plotted in Figure 7-6 versus time in days since an instantaneous loading. A different t
was chosen for settlement
than for pore pressure. This was necessary since settlement
due only to Stages 2 and 3 was measured. The t
was chosen
as CD390. The average consolidation of the entire clay stra130
tum was given by SR1 at the top of the clay. Incremental
consolidation of the layers within the clay was computed
from differential settlements of the layers.
There were sufficient valid data to compute consolidation of layers D and E up to CD 958. Based on the ratios
of consolidation
in layers D and E to that in the whole
lay-
er D & E for these early times, consolidation values were
extrapolated for these two seperate layers.
Figure 7-6 indicates the same general relative rates
of consolidation for the average and different layers as
the pore pressure data. The two uppermost layers (A and B)
and the lowest (E) consolidate more quickly than the other
layers or the average of the whole thickness. In the two upper layers, A and B, settlement apparently proceeds more
quickly than the rate of pore pressure dissipation. On the
other hand, at any given time, all other layers and the average indicate significantly less settlement than pore pressure dissipation. However, the amount of change in consolidation for these other layers is about the same for settlement and pore pressure:
U%
Layer
u
CD2053
AU%
u
p
CD 620-2053
p
Average
60.0
28.4
19.1
15.6
C
43.9
21.7
21.9
17.0
D + E
58.3
20.4
17.9
14.3
131
7.3
7.3.1
FIELD COEFFICIENTS OF CONSOLIDATION
General
Coefficients of consolidation, Cv, were backfigured
from both the pore pressure and settlement data. In order
to account for the markedly different laboratory values of
cv for the OC (cvl) and NC (cv2) clay, the two layer system
was transformed to a single equivalent layer with cv =
This method was first proposed by Gray
v2.
and later expanded
by Leonards (1962)
For an upper layer H
and cvl and a lower layer H2
and cv2, the single equivalent layer is:
He = H2 + H1
v
and has one cv = cv2. For all analyses the ratio of cv2/cvl
was taken equal to the ratio of the average lab
values at
vo and
vf
v2 and cvl
(i.e. cv2/cvl = 1/3) as discussed
in
Chapter 2. Analyses were performed for both the full clay
thickness and a reduced thickness in an attempt to remove the
affect of the very rapid consolidation of Layer A. In addition, for both thicknesses, varying elevations were used for
the break between cvl and cv2 in the transformation to a
single equivalent layer
7.3.2
Full Clay Thickness
For analyses on the full thickness, the break between
132
Cvl and
v2 was located at El. -70, -43 and -21.5. This re-
sulted in three different single equivalent layer thicknesses: 110, 121, and 130 ft., respectively.
Field values
of
cv2 were then computed from the relation:
2
cV
Cv2
TvHd
t
The Davis and Poulos (1972) graph for two-dimensional consolidation, with permeable top and base, was used to
determine Tv. This value must be multiplied by 4, since
their Tv is related to full height rather than drainage
height. This graph and one for an impermeable base are reproduced in Figures 7-7 and 7-8. These graphs account for
lateral drainage to some degree, since they assume isotropic
permeability. A value of H/b = 2 was used, and U was taken
from the pore pressures over the full depth, and the consolidation settlement of SR-1.
For all analyses, both the total time (t) and incremental time (At) methods were used. In the At method, ATv and
At are substituted in the basic equation. Table 7-12 summarizes the computed field
v2's, and Table 7-13 and 7-14 pre-
sent the actual computations.
Table 7-12 indicates theat there is better agreement
between t and At values for the settlement data than for the
pore pressure data. For both t and At methods there is relatively poor agreement between pore pressure and settlement
133
cv2's, with pore pressure values being significantly greater. For the t method, pore pressure values exceed settlement values by a factor of 5.5. The At method gives much
better agreement, with pore pressure
vs
greater by a fac-
tor of 2.5. The settlement data gives good agreement with
laboratory
v2 of 0.093 ft2 /day, and for the At method with
the cv change at -43 , matches it exactly.
7.3.3
Reduced Clay Thickness
In an attempt to minimize
the effect
of the extreme-
ly rapid consolidation of layer A (110 to -21.5), analyses
were performed only on the clay below El. -21.5. It was necessary to compute U values considering only the consolidation in the clay below El. -21.5. With the change in cv at
El. -70 and -43, the single equivalent layers were 103.5 and
114.5 feet respectively.
Field cv2 values were then computed in the same way
as for the full thickness analyses. However, new values of
U were computed, based on the pore pressures only below El.
-21.5,
and on the
c of SR-2
(El. -21.5). Table 7-15 sum-
marizes the resultant values, and the computations are presented
in Tables 7-16 and 7-17.
In this instance, the settlement data give excellent
agreement between the t and At methods and with the laboratory cv2 values. The pore pressure data do not give as good
134
agreement between the t and At methods, although it is much
better than for the full thickness analyses. Pore pressure
cv values are much higher than settlement values. For the t
method, they are higher by a factor of 5.1, and for the At
method higher by a factor of 3.1.
7.3.4
Predicted Consolidation Settlements
Several field cv2 values were used to pedict
con-
solidation settlment versus time. The value which gave the
prediction closest to measured
c at the end of loading was
Cv2 = 0.236 ft2 /day. This value was derived from pore pressure data with a full thickness analysis using At and the cv
change at El. -43. The predicted and measured
c for four
times are shown in Figure 7-9.
The laboratory and settlement values were too low,but
the chosen value gives an excellent prediction at the end of
loading (CD 620). For the laboratory measured (Guertin, 1967)
ratio
of horizontal
to vertical
permeability
(kH/kV) of 1.67,
the initial effects of lateral drainage are slight (Ladd and
Wissa, 1970). These effects are somewhat accounted for by the
use of the Davis and Poulos charts (kH/kv
=
1). However, with
increasing time the discrepancy in the kH/kv ratio becomes
more important, and the predicted
c lags slightly behind
measured values.
Consolidation settlements are predicted by entering the
135
appropriate
U - Tv graph with Tv from the relation,
Tv
Cvt
Hd 2
The U thus derived is applied to the predicted Pcf to determine the settlement at time t.
Rather than determining a single U for the full layer
by using Cv2 and the transformed drainage height (Lacasse
and Ladd, 1973), a different approach was used. Based on the
lab ratio cvl/cv2 = 3, field cvl values were computed from
the field cv2 values. Tvl and Tv2 values were computed as
discussed above, and Davis and Poulos' chart for an impermeable base was used to determine U1 and U2 values. These were
applied
to the predicted
differential
Pcf in the cvl and
cv2 zones and the resultant values summed vertically. This
method more closely relates to the layered consolidation behavior.
As a result of the above considerations, it appears
that the best prediction is given by cvl = 0.71 ft 2 /day above
El. -43, where OCR > 2.5, and cv2 = 0.24 ft 2 /day below El.
-43. These values are more than 2.5 times as great as the laboratory
values
(cvl = 0.27, cv2 = 0.09).
136
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8.
CONCLUSIONS AND RECOMMENDATIONS
It is apparent from this study that the details of a
field instrumentation program are extremely important. Improper installation of a few settlement rods and failure to
monitor casing settlements resulted in the loss of critical
data, making a detailed analysis much more difficult.
When there is any possiblilty that casing settlement
can interfere with a sensor, the casing must be monitored
periodically. In addition, the interchangeable use of several instruments to monitor horizontal deflection must be avoided, since this can result in very erratic data. In this respect, the M.I.T. Beaver system appears to produce very high
quality deflection data.
In order to maximize the usable data in this expensive instrumentation program, it is essential to repair or
replace certain critical instruments. This should be accomplished at least two months prior to the removal of the surcharge at the test section.
Of greatestpriority is the repair of settlement rods
affected
by casing drag
(SR-4, 6, 10 and 11). It is recommen-
ded that these casings be jacked out at least one foot. It
may be necessary to electro-osmotically release the casings
from the clay. The procedure might provide valuable electroosmotic data for Boston Blue Clay as a bonus.
163
In addition, at least some of the inoperative piezometers must be replaced To reduce the expense, simple single
tube Geonor M206 devices can be used. As a minimum, the following piezometers should be replaced: P7, P17, P18, P22,
P24, P27, and P28.
Based on comparisons of predicted and measured performance, the use of CKoUDSS hyperbolic parameters in conjunction with the finite element program FEECON is highly effective. However, it is recommended that additional FEECON
analyses be performed for the model footing tests to investigate the effect of avc on shear modulus for normally consolidated clay. It is also recommended that FEECON predicted
and measured internal embankment stresses be studied.
Excess pore pressures beneath the embankment were
equal to the modified Henkel relation Au = Aaoct + aAToct
where the a parameter is related to Skempton's A parameter
at failure: a = 3Af-l
.
The choice of a is also dependent
on the stress system, whether PSA, DSS, or PSP. Good results
were obtained by assuming PSA conditions beneath the embankment crest, DSS conditions beneath the slope, and PSP conditions outside the toe.
In the region where DSS conditions are likely (beneath
the slope) FEECON horizontal deflections agree well with measured values. It would be advisable to perform additional
164
analyses with input parameters based on CKUPSA
and PSP da-
ta in the appropriate locations.
At the top of the clay below the centerline, the predicted initial settlement is 0.2 feet. The predicted final
consolidation settlement is 7.8 feet.
Field values
of RR are 1.4 to 2.7 times as great
as
laboratory values, Field values of CR are 1.8 to 1.9 times
as great as laboratory CR and field values of CV are 2.7
times greater than laboratory cv's. The pore pressure and
settlement data indicate significantly different values for
the degree of consolidation in the field. However, both
types of datashow
about the same change in degree of con-
solidation with time.
In view of the very high field value of CR, it is recommended that soluble salt analyses be performed on samples
from the clay stratum. These may indicate leaching due to
the artersian pressure, a possible cause of increased compressibility.
16A%
9.
1.
Bromwell,
L.G.,
REFERENCES
Ryan, C.R., and Toth, W.E.
(1971) "Re-
cording Inclinometer for Measuring Soil Movement",
Porceedings, 4th Pan-American Conference on Soil
Mechanics and Foundation Engineering, Puerto Rico,
Vol. II, pp. 333-343,
2.
1971.
D'Appolonia, D.J., Lambe, T.W., and Poulos, H.G. (1971)
Evaluation of Pore Pressures Beneath an Embankment", ASCE, JSMFD, Vol. 97, No. SMG, June, 1971,
pp. 881-897.
3.
D'Appolonia, D.J., Poulos, H.G., and Ladd, C.C. (1971),
"Initial Settlement of Structures on Clay", ASCE,
JSMFD, Vol. 97, No. SM10, October, 1971, pp. 13591377.
4.
Davis, E.H. and Christian, J.T. (1971) "Bearing Capacity
of Anisotropic Cohesive Soil", ASCE, JSMFD, Vol. 97,
No. SM5, May, 1971, pp. 753-769.
5.
Davis, E.H. and Poulos, H.G. (1972) "Rate of Settlement
Under Two- and Three-Dimensional Conditions",
GEOTECHNIQUE
6.
22, No. 1, 1972, pp. 95-114.
Duncan, J.M. and Chang C.Y. (1970), "Non-Linear Analysis
of Stress and Strain in Soils", ASCE, JSMFD, Vol.
96, No. SM5, pp. 571-582,
1970.
7.
Foott, R. and Ladd, C.C. (1973), "The Behavior of Atchafalaya Test Embankments During Construction", Dept.
of Civ. Eng. Research Report R73-27, M.I.T., 1973.
8.
Guertin, J.D. Jr., (1967), "Stability and Settlement Analyses of an Embankment Clay", S.M. Thesis. Dept. of
Civ. Eng., M.I.T.,
June, 1967.
9.
Kinner, E.B. and Ladd, C.C. (1970), "Load Deformation Behavior of Saturated Clays During Undrained Shear",
Dept. of Civ. Eng. Research Report R70-27, M.I.T.,
May, 1970.
10.
Kulhway, F.H., Duncan, J.M. and Seed, H.B. (1969), "Finite
Element Analysis of Stresses and Movements in Embankments During Construction", Dept of Civ. Eng. Report
TE69-4, Univ. of Calif., Berkeley,
166
1969.
11.
Lacasse, S.M. and I add, C.C. (1973), "Behavior of Embankments on ew Liskeard Varved Clay", Dept. of
Civ. Eng. Res(arch Report R73-44 (unpublished) M.I.T.
September,
19 3.
(1971), "Settlement Analyses for Cohesive
Soils", Dept. of Civ. Eng. Research Report R71-2,
M.I.T. August, 1971.
12.
Ladd, C.C.
13.
Ladd, C.C. and Edge rs, L. (1972), "Consolidated Undrained
Direct Simple Shear Tests on Saturated Clays", Dept.
of Civ. Eng. 1esearch Report R72-82, M.I.T., July,
1972.
14.
Lambe, T.W. and Whi tman, R.V. (1969), SOIL MECHANICS,
John Wiley anc Sons, New York, 1969, 533 p.
15.
Lysmer, J. and Dunc an J.M. (1972), "Stresses and Deflections in Founc ations and Pavements", Dept of Civ.
Eng., Univ. oi Calif., Berkeley, 1972.
16.
Mitchell, J.K. and Gardner, W.S. (1971), "Analysis of
Load-Bearing
I ills over Soft Subsoils", ASCE, JSMFD,
Vol. 97, SM 1] ,November, 1971, pp. 1549-1571.
17.
Park, Y.K.
(1974), "Determination of In-Situ Stress in
Soil by Hydra.lic Fracturing", S.M. Thesis, Dept.
of Civ. Eng., M.I.T., February,
18.
1974
Perloff, W.H., Bale di, G.Y. and Harr, M.E. (1967),
"Stress Distri bution Within and Under Long Elastic
Embankments", HRB, HRR, No. 181, pp. 12-29, 1967.
19.
Recker, K.L. (1973) , "Measured Deformations and Pore
Pressures Undo r an Embankment on Soft Clay", S.M.
Thesis, Dept. of Civ. Eng., M.I.T., June, 1973.
20.
Recker,
K.L., Dowdi ng, C.H. and Lambe, T.W. (1973),
"Prediction
ol
the Magnitude and Rate of Movement
of the Embanka ent Upon Removal of the Surcharge at
1-95", Dept. f Civ. Eng. Research Report R73-45,
M.I.T., August , 1973.
21.
Reed, D.E. (1973), "Generalized Subsurface Soil and Rock
Descriptions n the Boston Area", Internal Report,
Haley and Aldi ich, Inc., Cambridge, Mass., 1973,
(unpublished)
167
22.
Simon, R.M. (1972), "Analysis of Embankment Construction
by Finite Elements", S.M. Thesis, Dept. of Civ. Eng.
M.I.T., Spetember, 1972.
23.
Simon, R.M., Christian,
J.T., and Ladd, C.C.
(1974),
"Analysis of Undrained Behavior of Loads on Clay",
ASCE Specialy Conference on Anlysis and Design in
Geotechnical Engineering, Austin, Texas, June, 1974.
24.
Simon, R.M., Ladd, C.C. and Cristian,
J.T.
(1972),
"Finite Element Program FEECON for Undrained Deformation Analyses of Granular Embankments on Soft
Clay Foundations",
Dept. of Civ. Eng. Research
Re-
port R72-9, M.I.T., 1972.
25
Storch Engineers (1965) "Soils and Foundations Report for
Interstate Route 95, Revere - Saugus, Massachusetts"
prepared for Highway Engineers, Inc., Boston, Mass.,
1965.
26.
Wolfskill, L.A. and Soydemir, C. (1971), "Soil Instumentation for the -95 MIT-MDPW Test Embankment", Dept.
of Civ. Eng. Research Report R71-28, M.I.T., August,
1971.
168
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LIST OF SYMBOLS
wl
=
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N
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bouyant unit weight
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total unit weight
=
total normal stress
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initial horizontal effective stress
a
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maximum vertical effective stress
avm=
a1
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a2
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a3
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octahedral total normal stress
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octahedral shear stress
RR
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recompression ratio
CR
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virgin compression ratio
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OC
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overconsolidation ratio
C
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ratio of secondary compression
211
(OCR > 1)
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coefficient of consolidation
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undrained shear strength
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normalized undrained shear strength
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isotropically consolidated undrained
triaxial compression test
CKoUC
Ko - consolidated undrained triaxial
compression test
CKoUPSA
Ko - consolidated undrained plane
strain active test
CKoUPSP
Ko - consolidated undrained plain
strain passive test
CKoUDSS
Ko - consolidatedundrained direct simple
shear test
U, UU
unconfined and unconsolidated undrained
triaxial compression test
Ks
ratio S(H)/Su(V)
Ko
coefficient of lateral stress at rest
a/b
ratio of major to minor half-axis in
anisotropic strength ellipse
A
Skempton's pore pressure parameter
a, b
Henkle's pore pressure parameters
q0
initial shear stress
Up
settlement
Pc
consolidation settlement
Pi
pi
initial settlement
Pcf
final consolidation settlement
u
pore pressure
ue
excess pore pressure
U
average degree of consolidation
213
=
vertical strain
y
=
shear strain
E
=
Young's modulus
G
=
Shear modulus, also Poisson's ratio
in granular relationship
Rf
=
ultimate stress factor
K
=
Bulk modulus
F
=
stress dependency factor for
Poisson's ratio
d
=
strain dependency factor for
Poisson's ratio
SuH
=
undrained shear strength for max.
compressive stress horizontal
SuV
=
undrained shear strength for max.
compressive stress vertical
214