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MOMENT RESISTING CONNECTIONS FOR SEISMIC APPLICATIONS
BROOK ROBAZZA AND SHILIN SUN
desired to have an open space uninterrupted by brace bays or walls.
This SFRS was the favoured seismic system in the USA prior to
the 1994 Northridge Earthquake [10]. This seismic event exposed
that the welded beam-to-column connections in these frames were
often the weak links in the seismic systems. It is therefore of
significant importance to continue the extensive study which has
already been performed on moment resisting connections for
seismic applications.
Abstract
A chain is only as strong as its weakest link. For structural design,
it can be restated as “a structure is as strong as its weakest
connection [7].” In other words, connection design is crucial to the
entire structural design. One of the specific structural design issues
is the behaviour of connections during earthquakes where ductile
response is expected from seismic force-resisting systems (SFRS).
Since steel is known as a ductile material with a high strength-toweight ratio, it is a desirable material to be used in earthquake
design as a means of reducing the damage to the structure. Steel
moment resisting connections have been proven to have excellent
performance as a part of seismic force-resisting systems [1]. Steel
moment frames which make extensive use of these connections
and are often used as the SFRS for many structures where it is
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Table of Contents
Abstract ........................................................................................... 1
Table of Contents ............................................................................ 1
List of Figures ................................................................................. 2
1.0 Introduction ............................................................................... 3
1.1 Bolting and Bolted Connections ........................................... 3
1.2 Welding and Welded Connection ......................................... 4
2.0 Steel Design for Seismic Applications ..................................... 5
2.1 Post-Northridge Earthquake Construction ............................ 6
2.1.1 Member Size of Connections ......................................... 6
2.1.2 Inelastic Behavior of Connections ................................. 6
2.1.3 Weld Procedure .............................................................. 6
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2.1.4 Shear Yielding of the Panel Zones................................. 6
2.1.5 Yield Stress of Structural Material ................................ 7
2.1.6 Stress Concentrations ..................................................... 7
2.1.7 After Northridge Earthquake Construction .................... 7
3.0 Objectives ................................................................................. 7
4.0 Plastic Design............................................................................ 8
4.1 Plastic Analysis ..................................................................... 8
4.2 Plastic Hinges........................................................................ 9
4.3 Prequalified Sections ............................................................ 9
4.4 Plastic Hinge Moment Calculation ..................................... 10
5.0 Investigated Moment Resisting Connections.......................... 11
5.1 Bolted Unstiffened End Plate Connection .......................... 11
5.2 Bolted Stiffened End Plate Connection .............................. 12
5.3 Reduced Beam Section Connection .................................... 13
6.0 Moment Resisting Connection Design ................................... 14
6.1 Canadian Code Governing Moment Connections for Seismic
Applications .............................................................................. 14
6.2 Design Procedure ................................................................ 14
6.2.1 Bolted Unstiffened End Plate Connection ....................... 15
6.2.1.1Bolt Diameter ............................................................. 15
6.2.1.2 End Plate Thickness .................................................. 16
6.2.1.3 Column Flange Thickness......................................... 16
6.2.1.4 Column Web Thickness ............................................ 17
6.2.1.5 Panel Zone Thickness ............................................... 17
6.2.1.6 Continuity Plates ....................................................... 18
6.2.1.7 Welded Joints ............................................................ 18
6.2.1.8 Bolted Unstiffened End Plate Connection – Summary
of Requirements and Limitations .......................................... 18
6.2.2 Bolted Stiffened End Plate Connection ........................... 19
6.2.2.1 Bolt Diameter ............................................................ 19
6.2.2.2 End Plate Thickness .................................................. 19
6.2.2.3 Column Flange Thickness......................................... 20
6.2.2.4 Column Web Thickness ............................................ 20
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6.2.2.5 Panel Zone Thickness ............................................... 20
6.2.2.6 Continuity Plates ....................................................... 21
6.2.2.7 Welded Joints ............................................................ 21
6.2.2.8 Bolted Stiffened End Plate Connection – Summary of
Requirements and Limitations .............................................. 21
6.2.3 Reduced Beam Section .................................................... 21
6.2.3.1 Flange Reduction Specifications............................... 21
6.2.3.2 Check Reduced Section Resistance to Applied Loads
............................................................................................... 21
6.2.3.3 Connection Shear ...................................................... 22
6.2.3.4 Panel Zone Thickness ............................................... 22
6.2.3.5 Continuity Plates ....................................................... 22
6.2.3.6 Weld Joint Details ..................................................... 22
6.2.3.7 Weld Access Holes ................................................... 22
6.2.3.8 Backing Bars ............................................................. 22
6.2.3.9 Welding Sequence for Bottom Beam Flange............ 23
6.2.4.0 Reduced Beam Section Connection – Summary of
Requirements and Limitations .............................................. 23
7.0 Changes between S16-09 and S16-01(R05) affecting
Connection Design ........................................................................ 23
8.0 Conclusion .............................................................................. 23
9.0 References ............................................................................... 24
10.0 Appendices ............................................................................ 23
List of Figures
Figure 1 Rivet Heading Procedure .................................................. 3
Figure 2 Anchor Bolts in Moment Connection............................... 4
Figure 3 Connection with Cover Plate ............................................ 7
Figure 4 Plastic Hinge Example [12] .............................................. 9
Figure 5 Plastic Hinge Location for an RBS Connection [20] ..... 10
Figure 6 Geometry of Bolted Unstiffened End Plate Connection [2]
............................................................................................... 12
Figure 7 Bolted Unstiffened End Plate Connection [10] .............. 12
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Figure 8 Geometry of a Bolted Stiffened End Plate Connection [2]
............................................................................................... 13
Figure 9 Bolted Stiffened End Plate Connection Example [13] ... 13
Figure 10 Geometry of Reduced Beam Section Connection [2] .. 13
Figure 11 Reduced Beam Section Example [10] .......................... 14
Figure 12 Critical Section Calculation Diagram ........................... 15
1.0 Introduction
Steel connections are constituted by individual steel structural
members. These members are usually connected by three types of
fasteners or connectors which are rivets, bolts or welds. The
structural members can be connected or assembled either in steel
shops or in the field at construction sites. Prior to the 1950s’, rivets
were used as standard fasteners for all types of steel structures, but
have gradually become obsolete because of the tedious procedure
to form rivet heads (Figure 1). To complete a rivet joint, the second
rivet head must be formed by either hammering or using a
hydraulic press which is a messy, noisy and hazardous process [7].
Figure 1 Rivet Heading Procedure
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1.1 Bolting and Bolted Connections
Bolts and welds are now used as standard connectors in fabricating
modern steel structures [7]. Bolting is the procedure to connect
steel pieces by mechanical means [16]. Compared to welded
connections, bolted connections are relatively easy to make and
inspect. Connection strength depends on the bolt types and the
connected steel pieces for bolts installed without pretension [16].
The bolt types are either common (ASTM Standard A307) or highstrength (ASTM Standards A325 and A490) bolts [16]. A307 Bolts
are manufactured based on the American Society for Testing and
Materials (ASTM) Specification A307, from low-carbon steel with
a minimum ultimate tensile strength of 415 MPa. This type of bolt
is usually square-head and used for minor connections. In most
bolted connection applications, high-strength bolts are used, which
are designated as A325 or A490 bolts. The A325 bolts are made
from carbon steel with a minimum tensile strength of 830 MPa,
according to the ASTM A325 specification, and then subjected to
heat treatment. The A490 bolts are made according to ASTM
A490 and are composed of high-strength alloy steel, with a
minimum tensile strength of 1040 MPa and are also heat-treated.
Both of these types of bolts are specified with their nominal
diameter in millimeters and the steel type used in their
manufacture. For example, M22 A325M stands for a 22 mm
diameter bolt made according to the specifications of ASTM A325.
The commonly used A325 bolt sizes are M16, M20, M22, M24,
M27, M30 and M36. A490 bolts are used in a connection with
heavier loads to reduce the number of fastener needed [7].
The bolted connections can be classified based on the types of
resultant force transferred. These connection types are concentric
connections (force transfer in tension and compression member),
eccentric connections (in reaction transferring brackets) and
moment resisting connections (in beam to column connections in
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frames) [9]. The moment resisting connections are more complex
to analyze than the first two types of bolted connections. This is
due to the fact that they involve moment transferred from the beam
to the column in moment resisting frames. When a connection is
designed to transfer moment from the beam to the column, the
connection is formed by connecting the flanges of the beam to the
column where the moment is transferred by the pair of tension and
compression forces in the top and bottom flange of the beam.
Meanwhile, shear force is usually transferred through moment
connections when the web of the beam is also connected to the
column. In general, this type of connection is used to connect beam
and column by the method of anchor bolts (See Figure 2). The
bolts in the connection are subjected to a combination of shear and
axial tension [9].
Figure 2 Anchor Bolts in Moment Connection
1.2 Welding and Welded Connection
Welding is a process of joining steel or metal parts by fusing or
melting them together at a joint to produce a continuous member.
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The steel members are usually referred to as base metal and the
material “fusing” them together is referred to as weld metal or the
filler. Some welding processes have been developed since it was
first used. The oldest welding method is believed to be forge
welding where blacksmiths joined iron and steel by heating and
hammering them [22]. It was the only welding method until the
end of the 19th century. Arc welding, oxyfuel welding and
resistance welding were developed late in the 20th century [23].
After Second World War, modern welding techniques were
developed to meet the demand for a reliable, fast and inexpensive
joining method [23]. The most popular one is the shielded metal
arc welding technique (SMAW) which uses shielding gas and
welding electrode (filler) [22]. It is used to make groove or butt
welds as well as fillet welds. Other similar welding techniques
include gas metal arc welding, submerged arc welding, flux-cored
arc welding, and electroslag welding. Later development include
laser beam welding, electron beam welding and robot welding. In
steel structural practice, 80% of the welded joints made are fillet
welds and only 15% are groove welds. Fillet welds are used to join
pieces in various positions such as tee (90°), skewed and lap joints.
Fillet welds are a partial penetration weld type which has fusion
through part of the thickness of the joining pieces. Groove welds
can be partial or full penetration welds. Full penetration groove
welds are required in the more important welding joints in moment
beam-column connections [8].
Compared to bolted connections, welded connections have many
advantages. Through the process of welding, a strong metallurgical
bond is created between two steel members so that they can form a
continuous and load bearing joint which enables direct transfer of
stress between members. The strength of welded connections
depends on the weldability of the base metal, the strength of the
weld metal and the process of welding [16].
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It also eliminates gusset and splice plates necessary for bolted
structures [9]. Without bolts and braces, the weight of the
connections can be simply minimized [9]. The absence of holes in
tension members improves the efficiency of the section.
Meanwhile, the cost of fabrication is less than bolted connections,
since the operations such as drilling and punching can be
eliminated. Due to the continuity of welded connections, stress
concentration effects are seldom discovered in welded connections.
Welded connections also have neat appearances even with
complicated shapes of structural members [7].
The disadvantages of welding are that it requires skilled welders as
well as inspectors [9]. Non-destructive inspections such as
magnetic particle, ultrasonic or radiographic methods, may be
carried out after welding to detect defects in welds in case the
cracks propagate from the defects under fatigue loading [9]. Field
welding may be problematic due to the location or environment
[9]. Large residual stresses may be increased by the distortion
developed in welded connections [16]. Therefore, it is important to
produce high quality welds with proper profiles, good penetration,
complete contact with base metal at all surfaces, no cracks,
porosity or inclusion [9]. However, the methods of connecting
steel structural members highly depend on the loads applied to the
structures and how to successfully achieve the design goals.
2.0 Steel Design for Seismic Applications
An earthquake is a sudden tremor or movement of the earth’s crust
that can cause the ground to shake. A violent earthquake can
trigger landslides, floods and cracks in the land. These can then
lead to large-scale destruction to life and property [6]. The most
basic property of earthquakes however is the ground accelerations
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imposed to masses on the surface of the crust. Buildings are prime
examples of these masses. As a result, structural designers and
engineers have been making efforts to improve earthquake
resistant structural design and to design structures that can
withstand earthquakes with no or minimum damage in most cases
[6]. It is difficult to design for earthquake loads, since they are
unpredictable with their amplitude, duration and frequency [6].
Buildings codes therefore present response spectrums and other
guidelines to analysis to provide a better approximation of these
seismic loads. Structures are primarily designed to carry vertical
loads, and therefore are prone to poor lateral load performance if
they are not designed for it. However, it is generally not
economically acceptable to design buildings to remain elastic in
regions where strong ground motions are anticipated. In these
regions, inelastic deformations must be accommodated as a means
of reducing the seismic demand during a moderate to high seismic
event. Therefore, the common seismic resistant design goal is
specified as “to ensure elastic behaviour under a moderate
earthquake which has a return period equal to the life of the
structure and prevent collapse under the extreme probable
earthquake” [6]. In other words, the designed structures can remain
elastic under moderate seismic events and experience damage, but
do not collapse, under severe seismic events.
Since steel is well known as a ductile material, and is strong under
compression and tension, it is widely used as an SFRS to resist
load reversals induced by cyclic motions under earthquake loading
[6]. However, the Northridge earthquake in 1994 was the starting
point where failure was discovered in the critical beam-to-column
connections. One year later, in the earthquake of Kobe, the beamto-column connections suffered severe damage and 10% of these
structures collapsed [1]. The failure of the beam-to-column
connections is because they prevented the formation of energy-
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dissipating plastic hinges and failed in a brittle manner. Following
the earthquakes in Northridge and in Kobe, the extensive damage
which occurred in many steel structures initiated intensive research
and testing undertaken by the Federal Emergency Management
Agency (FEMA) and a number of universities to produce better
seismic resistant steel frame design and construction [10]. A
number of improved beam-to-column connection design strategies
have been developed with the purpose of providing adequate
ductility response to seismic motion to prevent failure and collapse
[5]. From the research emerged an overall concept to avoid
damage to the columns of the structure. Therefore central to this
philosophy of considering the effect of yielding members on the
column is to prevent strong beam-weak column behaviour and
weak stories [10]. Weak story collapse is one of the most
undesirable failure mechanisms as it destabilizes the entire
structure above. It is thus essential to design efficient connections
so that ductile energy-dissipating plastic hinges can form at desired
locations to maintain continuity in force flow path and avoid
structural collapse [15].
However, the buildings damaged in the Northridge earthquake
employed W30 or larger beams connected to heavy W14 columns.
Therefore the initial testing was not representative of actual
practice. It appears that size plays a significant role in the behavior
of WSMF connections and that details that behave well for
connections using small sections do not necessarily behave as well
for larger sections.
2.1 Post-Northridge Earthquake Construction
2.1.3 Weld Procedure
An investigation conducted by FEMA addressed the factors which
contributed to the poor performance of Welded Steel MomentFrame (WSMF) buildings during the 1994 Northridge Earthquake.
The following sections examine the discoveries and notes
presented in FEMA 351 and FEMA 355.
Weld metal used to erect the previously used WSMF is low-notchtoughness weld metal which means it develops unstable brittle
fractures under high stress and strain demands. Meanwhile,
welding practice in many of the damaged structures was found to
be sub-standard such as inadequate fusion, failure to remove weld
backing. Compounding these factors led to many dangerous weld
connections and frequently resulted in brittle failure.
2.1.1 Member Size of Connections
Prior to the Northridge Earthquake, it was common that large
framing members were used even in relatively small buildings. The
initial testing of WSMF connections was conducted in the 1960s
and 1970s. All tested assemblies employed small-sized elements
which are W18 beams and light W12 and W14 column sections.
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2.1.2 Inelastic Behavior of Connections
Common steel design practice is to overdesign the connections so
the inelastic behavior occurs within the structural members.
However, typical detailing practice prior to the Northridge
earthquake relied on the large inelastic behavior occurring in the
beam-column connections, but the strength of ductility of any
connection is highly dependent on the quality of the workmanship
employed. Meanwhile, every process of making a connection such
as cutting, welding, and bolting, affects the behaviour of the
connection. Therefore, the probability of failure is very high in real
structures utilizing many moment resisting connections.
2.1.4 Shear Yielding of the Panel Zones
In the 1980s, some engineers believed that shear yielding of the
panel zones in a beam-column connection instead of flexural
hinging of the beam, was a more benign and desirable way to
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accommodate frame inelastic behavior. As a result, the building
code allowed weak panel zones in steel frames. However,
excessive yielding actually produces large secondary stresses at the
beam-flange-to-column-flange joint, which can exacerbate the
initiation of fractures [20].
2.1.5 Yield Stress of Structural Material
Between the 1980s and 1990s, new steel mills were replacing old
mills in the United States. The new mills used scrap-based steel
production which tends to produce steel with higher-strength.
However, since the steel is stronger than the strength defined in the
code, the designers designed the weak beams with the code
strength instead of the real strength. Therefore the sections do not
yield at expected stress levels. This leads to miscalculation of the
yield moment. It also thus caused the desired location of the plastic
hinge to be less predictable.
BROOK ROBAZZA AND SHILIN SUN
connection from the code. FEMA suggests each connection design
should be qualified by a program of prototype laboratory testing.
Between 1994 and 1996, the University of Texas discovered that
cover plate (See Figure 3) on the connections can encourage the
plastic behaviour to be within the beam elements. As a result,
cover plates were widely used during this period. In 1995, FEMA
267 published defined rating criteria for weld metal and standards
of design and fabrication of moment connections. Meanwhile,
FEMA then used these studies to test and certify different moment
connections including haunched connections, reduced-beamsection connections, vertical rib plate connections, side plate
connections and slotted web connections.
2.1.6 Stress Concentrations
Steel moment connections in moment frames usually experience
stress concentrations. In the design calculations of connection
capacity, it was presumed stresses were uniformly distributed
across beam flanges and the flexural stresses are carried primarily
by the flanges while shear stresses are carried primarily by the
web. However, it is common that the flange also carries significant
local bending and shear stress and the stresses are not uniformly
distributed within flange elements, which leads to cracking of
welds and initiates brittle fractures of weld metal. These brittle
fractures cause failure of the joint as well as fractures spreading
into the column.
2.1.7 After Northridge Earthquake Construction
After the 1994 Northridge earthquake, FEMA removed the
prequalified status of typical bolted-web welded-flange moment
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Figure 3 Connection with Cover Plate
3.0 Objectives
Some types of rigid connections have been prequalified by FEMA
for seismic applications [1]. The scope of this paper is to
investigate three types of moment resisting connections for seismic
applications by providing detailed design drawings, calculations
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and design procedures. A formatted spreadsheet as well is provided
for the design of practical examples from existing buildings. This
spreadsheet may be used to check not only the adequacy of a given
end plate, but also determines how much it is over or under
designed.
4.0 Plastic Design
Since it is economically impractical to design structures to resist all
seismic loads elastically, it is essential that plastic design be
utilized. This of course means that structures are designed to
undergo deformation and damage during large seismic events. The
main design philosophy of most building codes is thus intended to
protect for life safety by avoiding earthquake-induced collapse in
severe events, while permitting large amounts of structural and
nonstructural damage [12].
4.1 Plastic Analysis
Plastic deformation occurs after the specimen has been loaded past
its yield strength. As the specimen is loaded beyond its elastic
limit, the stress remains constant while the strain increases [17].
For structural members, when the moment capacity is exceeded by
the moment demand, a plastic hinge is formed which causes this
section of the member to rotate at the plastic moment capacity.
Plastic analysis is based on determining the maximum load that a
structure can withstand before the structure collapses. The final
collapse occurs when sufficient plastic hinging has occurred to
convert the structure into a mechanism [17].
In determining the collapse mechanism and the required plastic
moment diagram, there are a number of methods which may be
employed. The mechanism or the virtual work method is the
method to be studied here. This method incorporates a design
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procedure requiring the determination of the largest plastic
moment caused by a number of investigated failure mechanisms
[18]. This is because the method is a generalized one which is
based on upper bound theory of plastic design. This states that the
load corresponding to the assumed method of collapse must be
greater than or equal to the collapse load by definition. Therefore
the largest plastic moment calculated is associated with the
governing failure mechanism. The following is the procedural
steps for this method [18]:
Step 1: Assume a specific collapse mechanism.
Step 2: Calculate the amount of internal virtual work which is
defined as the sum of all the products of the plastic
moments and their corresponding internal virtual angle
changes. The internal virtual angle changes are computed
by designating any one angle as θ and calculating all
others in terms of θ and the geometry of the frame.
Step 3: Calculate the amount of external virtual work which is
defined as the sum of all the external loads times the
virtual distance through which they move at the collapse
mechanism. This distance is calculated by recognizing
that it is the product of the angle θ and the distance from
the angle change to the load.
Step 4: Equate external to internal work. The angle θ cancels out
and Mp can be solved for in terms of the loads and the
dimensions of the frame.
Step 5: Sufficient collapse mechanisms are tried so that the
designer is satisfied that the one with the largest Mp has
been found. This is done by drawing the plastic moment
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diagram for each trial mechanism to see that there are no
moments larger than the plastic moment
Figure 4 Plastic Hinge Example [12]
4.2 Plastic Hinges
Steel moment frames, when properly designed, are a perfect
example of what the building codes desire in terms of seismic
performance. By utilizing moment resisting connections, moment
frames are intended to experience inelastic behaviour
accommodated through the formation of plastic hinges at beamcolumn joints and column bases. These locations are ideal for the
structure in that they prevent undesirable failure mechanisms. It is
a basic principle of plastic design to prevent undesirable
mechanisms by forcing more desirable failure mechanisms to fail
first. This is essentially providing a “fuse” for the structure. Plastic
hinges provide this fuse. Plastic hinges allow for rotation which
can only occur when the bending moment reaches the plastic
moment capacity. The area around the section of maximum
bending moment is a region of localized plastic deformations form
in which the curvature is much larger than elsewhere [17].
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Plastic hinges form through flexural yielding of beams and
columns and shear yielding in the panel zones [12]. A plastic hinge
usually forms due to large cyclic plastic deformations imposed to
the steel section and results in local buckling (See Figure 3).
Severe local buckling causes strength loss and is the reason that
plastic hinges are preferred to occur in the beam rather than the
column. Yielding of the columns is normally discouraged or
prohibited for the reason that frame systems built with weakcolumn behaviour are known to have significantly larger inelastic
story drifts and local ductility demands than comparable frames
with yielding in the beams [19]. An alternative to having a plastic
hinge form in the beam is to have it form in the panel zone.
Research has found that there is high ductility in the panel zone but
high shear stress in the panel zone and high stresses in the column
flanges is found to possibly direct cracks that initiate at the welded
joint into the column [19]. Therefore, panel zone yielding can be
relied upon to provide plastic rotational capacity for seismic
resistant design, but total reliance on this ductility is generally in
appropriate [19]. The state of stress in the panel zone is also
extremely complex and deformations must be divided into axial,
flexural, and shear [21]. Excessive yielding actually produces large
secondary stresses at the beam-flange-to-column-flange joint and
exacerbates the initiation of fractures [19]. These factors all
contribute to the conclusion that flexural yielding is best
accommodated in the beams of the structure.
4.3 Prequalified Sections
Some moment resisting connections are prequalified by the AISC
to help reduce difficulties arising from the qualification process
which requires extensive full-scale testing [12]. All of the
connections investigated in this report are of this prequalified
variety. These connections have exhibited adequate performance
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reinforcement, and other connection conditions.
Calculated as:
𝐹𝑦 + 𝐹𝑢
𝐶𝑝𝑟 =
𝐹𝑦
based on extensive testing and from review by an expert review
panel. Therefore, for these connections a set of design procedures
have been developed such that adequate plastic design is utilized.
The plastic hinges locations for these prequalified connections are
valid for beams with gravity loads representing a small portion of
the total flexural demand and for conditions of strong columnweak beam behaviour. For frames in which gravity loading
produces significant flexural stresses in the members, or frames
that do not have strong-column, weak-beam configurations,
locations of the plastic hinge formation should be determined
based on a method of plastic analysis such as the mechanism or the
virtual work method described earlier in Section 4.1.
A value of 1.2 may be used for all cases, except
where otherwise noted in the individual
connection design procedure provided for
prequalified sections
RyFy = expected yield stress of the beam sections
as defined in Clause 27.1.7 of S16-01
Ry = 1.1 according to Clause 27.1.7 and RyFy need
not be less than 460 MPa for HSS section or 385
MPa for other sections.
Ze = the effective plastic modulus of the beam
section at the location of the plastic hinge. For
Bolted End Plate Connections, Ze = the plastic
modulus of the unreduced beam section, Zb
4.4 Plastic Hinge Moment Calculation
FEMA 351 gives a detailed analysis of how to calculate the plastic
hinge moment as well as the plastic hinge location, where this
moment is attained. The analysis was based on data collected from
tests on the prequalified connections. A similar, slightly modified
analysis for calculating the probable plastic hinge moment is
stipulated by CISC and presented below. The plastic hinge location
for each connection is provided in Section 6. See Figure 4 for a
diagram indicating the plastic hinge location.
For fully restrained connections designed to develop plastic
hinging in the beam or girder, the probabilistic plastic moment at
the location of the plastic hinge should be determined as:
𝑀𝑝𝑟 = 𝐶𝑝𝑟 𝑅𝑦 𝑍𝑒 𝐹𝑦
where:
Mpr = probable peak plastic hinge moment
Cpr = a factor to account for the effects of strain
hardening, local restraint, additional
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Figure 5 Plastic Hinge Location for an RBS Connection [20]
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5.0 Investigated Moment Resisting
Connections
In order to provide a highly ductile response and reliable
performance for beam-to-column connections, two key concepts
were developed: strengthening the connection or weakening the
beam framing into the connection or both [5]. These concepts are
recognize the fact that a strong column-weak beam configuration is
difficult to design due to large column size requirements since
beams in typical moment frame buildings will be deeper than 350
mm to support dead and live loads alone [10]. Large columns with
moment capacity exceeding that of the beams will usually result in
thick column flanges which are prone to welding problems [10]. In
addition, columns with depths greater than 350 mm present
problems when attempting to accommodate Canadian drift
limitations [10]. For these reasons, the given two concepts are
employed.
Reinforcing the connection is designed to provide a beam-tocolumn connection which is stronger than the beam section. By
providing this strong connection, the plastic hinge is forced away
from the face of the column. This causes large stresses and
inelastic strains to be developed in the beam with enough distance
from the column to prevent unacceptable stress and strain levels
arising in the column. This method of beam-to-column connection
is exhibited in the below Bolted Unstiffened End Plate and Bolted
Stiffened End Plate connections.
Weakening the beam framing into the connection is an alternative
method which provides similar benefits to reinforcing connections.
The idea of weakening relies on the fact that portions of the beam
flange are trimmed away in a region adjacent to the beam-to-
UNIVERSITY OF BRITISH COLUMBIA
BROOK ROBAZZA AND SHILIN SUN
column connection [5]. This method produces a “ductile fuse”
which forces yielding to occur within the reduced section of the
beam where large inelastic strains can be sustained [5]. By
providing this measure, limited stress and strain is developed near
or at the column. This beam-to-column method is best described
by the Reduced Beam Section connection.
Each connection type has its respective advantages and
disadvantages. Bolted End Plate connections aid in reducing the
drift in moment frames but also increase the cost, and with large
bolted end plate connections, can result in very large welds and
higher degrees of restraint [5]. Reduced Beam Section connections
often have a lower cost than Bolted End Plate connections but also
have increased the elastic drift due to the reduction in beam
stiffness [1]. As always when considering cost however, one must
also consider a number of factors varying from craftsmen wages
and geographical location to weather climate.
The following connections are believed to cover most practical
applications in Canada according to the Canadian Institute of Steel
Construction [1]. The first two connections represent the design
philosophy of reinforcing the connection while the last connection
investigated represents the design philosophy of weakening the
beam. All of these connections are prequalified and each has a
corresponding appendix in the report provided by CISC which
outlines a summary of the requirements and limitations stipulated
by the CISC
5.1 Bolted Unstiffened End Plate Connection
The Bolted Unstiffened End Plate (BUEP) connection utilizes an
extended end plate and is a common form of field-bolted moment
connection [1]. This connection is made by shop-welding the beam
to the end plate using a complete joint penetration groove (CJPG)
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welded joint for the beam flanges to the end plate and fillet welds
for the beam web to end plate joint [3]. After completion of the
shop-welding, the end plate is then field-bolted to the column
using eight bolts. See Figure 6 for the geometry of a bolted
unstiffened end plate connection.
Figure 7 Bolted Unstiffened End Plate Connection [10]
5.2 Bolted Stiffened End Plate Connection
Figure 6 Geometry of Bolted Unstiffened End Plate Connection [2]
The BUEP connection should be designed such that yielding may
be accommodated either as a combination of beam flexure and
panel zone yielding or solely due to beam flexure. The end plates,
bolts, and welds should be designed to prevent significant yielding
from occurring in these elements [1].The design procedure for the
proportioning of this type of connection is provided in Section 6.
See Figure 7 for an example of a bolted unstiffened end plate
connection.
UNIVERSITY OF BRITISH COLUMBIA
The Bolted Stiffened End Plate (BSEP) connection is identical to
the bolted unstiffened end plate connection but in this case the
outstanding flanges of the end plate at the top and bottom of the
beam are stiffened by a vertical fin plate that extends outward from
the beam flanges [1]. These stiffener plates are CJPG double-bevel
groove welded to the beam flanges and end plates [2]. See Figure 8
for the geometry of a bolted stiffened end plate connection.
As with the unstiffened variety of Bolted End Plate connection, the
BSEP connection should be designed such that yielding may occur
either as a combination of beam flexure and panel zone yielding or
due to beam flexure alone. The end plates, bolts, and welds should
be designed to prevent significant yielding from occurring in these
elements [1]. The design procedure for the proportioning of this
type of connection is provided in Section 6. See Figure 9 for an
example of a bolted stiffened end plate connection.
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weld metal to join the flanges of the beam to the column [1]. The
web joints for these connections may be either complete
penetration groove welds, or bolted or welded shear tabs [3]. When
the RBS connection is used in design it is important to take into
consideration the reduction in beam stiffness which occurs due to
the trimming of the flanges in the “dog-bone” region. Unless a
detailed analysis is conducted, a 7% to 9% increase in elastic drift
should be accounted for flange reductions of 40% and 50%
respectively [1]. See Figure 10 for the geometry of a reduced beam
section connection.
Figure 8 Geometry of a Bolted Stiffened End Plate Connection [2]
Figure 9 Bolted Stiffened End Plate Connection Example [13]
5.3 Reduced Beam Section Connection
The Reduced Beam Section (RBS) connection utilizes circular
radius cuts in both top and bottom flanges of the beam to reduce
the flange area over a length of the beam near the ends of the beam
span [3]. This connection uses no reinforcement other than the
UNIVERSITY OF BRITISH COLUMBIA
Figure 10 Geometry of Reduced Beam Section Connection [2]
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The RBS connection should be proportioned to allow yielding to
occur as either a combination of flexural yielding of the panel zone
and reduced section yielding or as the reduced section yielding
alone [1]. The beam-flange-to-column joints and the beam web
connection should be designed to prevent significant yielding from
occurring in these elements. The design procedure for the
proportioning of this type of connection is provided in Section 6.
BROOK ROBAZZA AND SHILIN SUN
these detailed design procedures for the three connections
investigated in this report.
6.1 Canadian Code Governing
Connections for Seismic Applications
Moment
The CISC Handbook contains detailed information regarding the
design and detailing of structural steel in SI metric units.
The Tenth Edition has been updated to reflect changes to CSA
S16-09 and the steel sectional data. This code uses the limit state
design procedure and factors up load demands while factoring
down material resistances. The factoring procedure is based on
probabilistic distributions of resistances and loads produced by
numerous experimental testing data in order to provide an
acceptable level of safety in a given structure. The code is intended
to be used in conjunction with the NBCC 2010. Clause 27 of CSA
Standard S16-09 is used for seismic design and stipulates the
performance criteria for beam-to-column connections for Ductile
(Type D) and Moderately Ductile (MD) Moment-Resisting
Frames.
6.2 Design Procedure
Figure 11 Reduced Beam Section Example [10]
6.0 Moment Resisting Connection Design
The general design objectives of the three types of moment
connection studied is to provide large, stable, plastic rotational
capacity with an aim to mobilize at least one ductile element
precluding any undesirable failure modes [1]. According to the
Canadian Institute of Steel Construction (CISC) [1], the design
procedures of moment connections include the following:
In order to help engineers design moment resisting connections,
the CISC provides detailed design procedures for prequalified
moment resisting connections. The following section will present
UNIVERSITY OF BRITISH COLUMBIA
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
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Identify the undesirable failure modes and all yielding
mechanisms
Determine the probably peak capacity of the primary
yielding mechanism and the onset of yielding in some cases
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Proportion the connection to ensure that the nominal
resistances against each of the undesirable failure modes at
least equal the probably peak capacity of the primary
yielding mechanism
In order to achieve the most desirable sequence of yielding
for connections in Type D and Type MD fames, each
connection should be proportioned to prevent yielding of
any secondary yielding mechanism prior to the onset of
yielding of the primary yielding mechanism.
Clause 27.2.5 stipulates code limitations on beam-to-column joints
and connections. The beam-to-column joint shall maintain strength
at the column face of at least the plastic moment resistance of the
beam, Mpb, through a minimum interstorey drift angle of 0.04
radians under cyclic loading according to Clause 27.2.5.1. In the
case of reduced beam section connections or when local buckling
limits the flexural strength of the beam, the beam need only
achieve 0.8Mpb at the column face when an interstorey drift angle
of 0.04 radians in developed under cyclic loading. Clause 27.2.5.2
states that the factored resistance of the beam web-to-column
connection shall equal or exceed the effects of gravity loads
combined with shears induced by moments of 1.1RyZFy acting at
plastic hinge locations. If the plastic hinge is to be located in the
column, which will not be the case for these moment-resisting
connections, Clause 27.2.8 for protected zones must be considered
as well as Clause 27.2.3. When the plastic hinge is designed to
occur in the beam, Clause 27.2.2 must come into effect. This
clause stipulates that the beam sections be Class 1 sections, be
laterally braced in accordance with Clause 13.7(b) and the forces
acting on other members and connections due to plastic hinging
shall be calculated using 1.1Ry times the nominal flexural
resistance, ZFy. It is important to note here that the following
design procedures are done assuming that the beams and columns
UNIVERSITY OF BRITISH COLUMBIA
BROOK ROBAZZA AND SHILIN SUN
of the connection have already been checked to have adequate
resistance to the design forces. These design procedures are thus
only for designing a moment-resisting connection to perform is a
desirable manner.
Figure 12 Critical Section Calculation Diagram
6.2.1 Bolted Unstiffened End Plate Connection
The following design procedure for the BEUP connection has been
laid out with guidance from CISC [1]. The procedure examines the
restrictions on the proportioning of the connection in no particular
order but they must all be satisfied in order to obtain a safe
connection. Guidelines for continuity plates and welded joint
design are presented as well.
6.2.1.1 Bolt Diameter
Bolt diameter must be large enough and of the correct type to resist
Mcf and Vcf. The following conditions check whether the bolt
configuration is appropriate.
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In order to prevent a Mode 1: Bolt Tension failure in BUEP
connections, the following equation must be satisfied. Otherwise
the bolt size or type must be reconsidered in order to resist Mcf.
𝑀𝑐𝑓
0.75𝐴𝑏 𝐹𝑢 ≥ 2(𝑑
1 +𝑑2 )
𝑡𝑝 ≥
𝑀𝑐𝑓
𝑏
𝑏𝑝 𝑑
√
1
2
1
𝑝 1
0.8𝐹𝑦𝑝 {(𝑑𝑏 − 𝑝𝑡 ) [ 2 (𝑝 + 𝑠 ) + (𝑝𝑓 + 𝑠) 𝑔] + 2 (𝑝𝑏 + 2)}
𝑓
𝑓
where:
where:
Fu = Specified minimum tensile strength (MPa)
Ab = bolt area based on nominal diameter
Mcf is defined in Figure 6
d1 and d2 are defined in Figure 6
To prevent a Mode 2: Bolt Shear failure in BUEP connections, the
following equation must be satisfied. Bear in mind here that the
bolt threads should not intercept the shear plane.
𝑠 = √𝑏𝑝 𝑔
See Figure 6 for the geometry definitions above.
To prevent a Mode 4: End Plate Shear failure, end plate shear
yielding should be precluded by selecting an end plate thickness
satisfying following condition:
𝑡𝑝 ≥
3𝐴𝑏 (0.5𝐹𝑢 ) ≥ 𝑉𝑐𝑓
where:
Vcf = is defined in Figure 12
Ab and Fu are defined above
6.2.1.2 End Plate Thickness
The end plates which are used in the BUEP connection should be
composed of CSA G40.21 300W or ASTM A36 steel. For both of
these types of steel, 𝐹𝑦𝑝 should be taken as 250 MPa. The
following restrictions are imposed to the configuration of the end
plate thickness 𝑡𝑝 .
BROOK ROBAZZA AND SHILIN SUN
𝑀𝑐𝑓
1.1𝐹𝑦𝑝 𝑏𝑝 (𝑑𝑏 − 𝑡𝑏 )
6.2.1.3 Column Flange Thickness
The BUEP connection configuration is dependent on the column
flange thickness. The following conditions check whether a
continuity plate is required for the connection depending on the
thickness of the column flange.
To prevent a Mode 5a: Beam Flange Tension Effect on Column
Flange failure in BUEP connections without continuity plates the
following equation must be satisfied. If the column flange is
thinner than is required for this condition, then continuity plates
should be provided.
To prevent a Mode 3: End Plate Flexure failure, end plate yielding
must be precluded by ensuring that the end plate thickness
satisfies:
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𝑀𝑐𝑓
(
)𝐶
𝑑
√ 𝑏 − 𝑡𝑏 1
𝑡𝑐 ≥
2𝐹𝑦𝑐 𝑐
𝐶2 =
𝑏𝑐 − 𝑔
2
𝑠=√
where:
c is defined in Figure 8
𝑔
𝐶1 = − 𝑘1
2
k1 = distance from the centerline of the column web
to flange toe of fillet as provided in Table 4.1
and in Part 6 of the CISC Handbook 10th
Edition
To prevent a Mode 5b: Beam Flange Tension Effect on Column
Flange failure in connections with continuity plates, the following
equation must be satisfied. If the column flanges do not satisfy the
below equation, either a new connection must be considered or a
new column section must be selected.
BROOK ROBAZZA AND SHILIN SUN
𝐶1 𝐶2
(2𝑏𝑐 − 4𝑘1 )
𝐶2 + 2𝐶1
6.2.1.4 Column Web Thickness
In the BUEP connection, continuity plates must be provided if the
column web thickness does not satisfy the following condition.
This prevents against Mode 6: Beam Flange Compression Effect
on Column without Continuity Plates failure. Otherwise continuity
plates must be provided.
𝑤𝑐 ≥
𝑀𝑐𝑓
(𝑑𝑏 − 𝑡𝑏 )(6𝑘𝑒 + 2𝑡𝑝 + 𝑡𝑏 )𝐹𝑦𝑐
where:
𝑘𝑒 = the k-distance of the column section for
engineering design.
𝑀𝑐𝑓
√ 2(𝑑𝑏 − 𝑡𝑏 )
𝑡𝑐 ≥
0.8𝐹𝑦𝑐 𝑌𝑐
where:
𝑐
1
2
4 2
𝑌𝑐 = ( + 𝑠) ( + ) + (𝐶2 + 𝐶1 ) ( + )
2
𝐶2 𝐶1
𝑐 𝑠
𝑔
𝐶1 = − 𝑘1
2
UNIVERSITY OF BRITISH COLUMBIA
6.2.1.5 Panel Zone Thickness
Connections for Type D and Type MD frames should be
proportioned to allow yielding to occur either as a combination of
beam flexure or panel zone yielding or beam flexure alone. Beam
flexure yielding is the primary yielding mechanism and panel zone
yielding the secondary yielding mechanism. In order to prevent
Mode 7: Panel Zone Shear failure it is necessary for one-sided
connections to have a panel zone thickness satisfying the following
condition.
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ℎ − 𝑑𝑏
)
ℎ
𝑤′ ≥
0.9(0.6𝑅𝑦𝑐 𝐹𝑦𝑐 𝑑𝑐 )(𝑑𝑏 − 𝑡𝑏 )
𝐶𝑦 𝑀𝑐 (
1) For two-sided connections, the thickness of the continuity
plate should be at least equal to the thicker of the beam
flanges.
2) For one-sided connections, the continuity plate thickness
should be at least one-half of the thickness of the beam
flange.
where:
h = average storey height of the stories above and
below the beam-to-column intersection, except:
a) where the column below has a pinned base, h = sum of the
storey height below and one-half the storey height above
and
b) for top level connections, h = the storey height but twice
the storey height where the column has a pinned base,
𝑆𝑒
𝐶𝑦 =
𝐶𝑝𝑟 𝑍𝑒
𝑀𝑐 is defined in Figure 12
𝑅𝑦𝑐 𝐹𝑦𝑐 = the probable yield stress of the column section in
accordance with 27.1.7 of S16-01 and
𝑆𝑒 = effective elastic modulus of beam section at the plastic
hinge location
6.2.1.6 Continuity Plates
Continuity plates may be required in some BUEP connections
when the column flange and/or web thickness is inadequate to
provide resistance against undesirable failure modes. These failure
modes are corresponding to limitations just described. Continuity
plates are also referred to as panel zone horizontal stiffeners and
column transverse stiffeners. When using continuity plates it is
important to consider that increasing the column strength may be
more economical than adding the continuity plates. This is due to
high labour costs associated with welding in many countries
including Canada. The following requirements must be met for the
design.
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BROOK ROBAZZA AND SHILIN SUN
6.2.1.7 Welded Joints
The beam flange-to-plate joints should be complete-penetrationgroove-welded joints (CJPG). This type of joint has special
requirements as access holes are not permitted. One sequence for
welding each single-bevel T-joint is the following:
a) install an 8-millimetre fillet on the inner flange face,
serving as backing,
b) gouge root of backing to sound metal, then
c) complete the groove weld in a horizontal or flat position
The beam web-to-end plate joint should be either a fillet CJPG
welded joint or a fillet-welded joint. A fillet-welded web should
have welds on both sides of the web and the welded web
connection should be proportioned to resist the more severe load
effect of the following two effects:
a) flexural yielding capacity of the web,
b) 2Vcf, where Vcf is defined in Figure 12
6.2.1.8 Bolted Unstiffened End Plate Connection –
Summary of Requirements and Limitations
See Appendix A for the summary of requirements and limitations
for the Bolted Unstiffened End Plate Connection.
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6.2.2 Bolted Stiffened End Plate Connection
The following design procedure for the BSEP connection has been
laid out with guidance from CISC [1]. The procedure examines the
restrictions imposed on the proportioning of the connection in no
particular order but they must all be satisfied for a safe connection.
Guidelines for continuity plates and welded joint design are
presented as well.
To prevent a Mode 2: Bolt Shear failure in BSEP connections, the
following equation must be satisfied. Like the BUEP connection,
the bolt threads should not intercept the shear plane.
6𝐴𝑏 (0.5𝐹𝑢 ) ≥ 𝑉𝑐𝑓
where:
Vcf = is defined in Figure 12
Ab and Fu are defined above
6.2.2.1 Bolt Diameter
Bolt diameter must be large enough and of the correct type to resist
Mcf and Vcf. The following conditions check whether the bolt
configuration is appropriate.
In order to prevent a Mode 1: Bolt Tension failure in BSEP
connections, the following equations must be satisfied. Otherwise
the bolt size or type must be reconsidered in order to resist Mcf.
0.75𝐴𝑏 𝐹𝑢 ≥
𝑀𝑐𝑓
3.4(𝑑2 + 𝑑3 )
2.58
3.25 × 10−6 𝑝𝑓 0.591 𝑃𝑐𝑓
1.91 0.327 0.965
𝑡𝑝0.895 𝑑𝑏𝑡
𝑡𝑠
𝑏𝑝
The end plates which are used in the BSEP connection should be
composed of CSA G40.21 300W or ASTM A36 steel. For both of
these types of steel, 𝐹𝑦𝑝 should be taken as 250 MPa. The
following restrictions are imposed on the configuration of the end
plate thickness 𝑡𝑝 .
+ 𝑇𝑏
where:
Fu = Specified minimum tensile strength (MPa)
Ab = bolt area based on nominal diameter
Mcf is defined in Figure 12
d1 and d2 are defined in Figure 8
dbt = bolt diameter
𝑀𝑐𝑓
𝑃𝑐𝑓 =
𝑑𝑏 − 𝑡𝑏
UNIVERSITY OF BRITISH COLUMBIA
6.2.2.2 End Plate Thickness
To prevent a Mode 3: End Plate Flexure failure mechanism, end
plate yielding must be precluded by ensuring that the end plate
thickness is at least equal to the larger of the following equations.
and
0.75𝐴𝑏 𝐹𝑢 ≥
BROOK ROBAZZA AND SHILIN SUN
𝑡𝑝 ≥
0.9
154 × 10−6 𝑝𝑓0.9 𝑔0.6 𝑃𝑐𝑓
0.9 0.1 0.7
𝑑𝑏𝑡
𝑡𝑠 𝑏𝑝
and
𝑡𝑝 ≥
267 × 10−6 𝑝𝑓0.25 𝑔0.15 𝑃𝑐𝑓
0.7 0.15 0.3
𝑑𝑏𝑡
𝑡𝑠 𝑏𝑝
where:
𝑑𝑏𝑡 is the bolt diameter
To prevent a Mode 4: End Plate Shear failure, end plate shear
yielding should be precluded by the use of an appropriate stiffener
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plate. These plates should be clipped at the intersection of the end
plate and the beam flange.
below equation, either a new connection must be considered or a
new column section must be selected.
6.2.2.3 Column Flange Thickness
The BSEP connection configuration is also dependent on the
column flange thickness. The following conditions check whether
a continuity plate is required for the connection depending on the
thickness of the column flange.
𝑀𝑐𝑓
√ 2(𝑑𝑏 − 𝑡𝑏 )
𝑡𝑐 ≥
0.8𝐹𝑦𝑐 𝑌𝑐
where:
𝑐
1
2
4 2
𝑌𝑐 = ( + 𝑠) ( + ) + (𝐶2 + 𝐶1 ) ( + )
2
𝐶2 𝐶1
𝑐 𝑠
𝑔
𝐶1 = − 𝑘1
2
𝑏𝑐 − 𝑔
𝐶2 =
2
To prevent a Mode 5: Beam Flange Tension Effect on Column
Flange failure on BSEP connections without continuity plates the
following equation must be satisfied. If the column flange is
thinner than is required for this condition, then continuity plates
should be provided.
𝑡𝑐 ≥ √
𝛼𝑚 𝑃𝑐𝑓 𝐶3
0.9𝐹𝑦𝑐 (3.5𝑝𝑏 + 𝑐)
𝑠=√
where:
1⁄
3
𝐴𝑓
∝𝑚 = 𝐶𝑎 ( )
𝐴𝑤
𝐶1 =
1⁄
4
𝐶3
( )
𝑑𝑏𝑡
𝑔 𝑑𝑏𝑡
−
− 𝑘1
2
4
and
Ca = 1.45 for A325 bolts and 1.48 for A490 bolts
Af = Area of beam tension flange
Aw = Area of beam web, clear of flanges
BROOK ROBAZZA AND SHILIN SUN
𝐶1 𝐶2
(2𝑏𝑐 − 4𝑘1 )
𝐶2 + 2𝐶1
6.2.2.4 Column Web Thickness
The BSEP connection has the same requirements on column web
thickness as the BUEP connection. Like the BUEP connection, if
the BSEP connection does not meet the requirements provided in
Section 6.2.1.4, continuity plates must then be provided.
6.2.2.5 Panel Zone Thickness
The panel zone requirements for the BSEP connection are the same
as those for the BUEP connection described in Section 6.2.1.5.
To prevent a Mode 5: Beam Flange Tension Effect on Column
Flange failure on connections with continuity plates, the following
equation must be satisfied. If the column flanges do not satisfy the
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6.2.2.6 Continuity Plates
6.2.3.1 Flange Reduction Specifications
The requirements on the continuity plates for the BSEP connection
are also the sameare identical to the BUEP connection listed in
When selecting the length and location of the beam flange
reduction, the following limits must be imposed:
are the same imposed on the BUEP connection as described in
Section 6.2.1.6.
0.5𝑏 ≤ 𝑎 ≤ 0.75𝑏
0.65𝑑 ≤ 𝑠 ≤ 0.85𝑑
6.2.2.7 Welded Joints
The welded joint specifications for the BSEP connection are the
same specifications imposed on the BUEP connection and
described in Section 6.2.1.7.
0.20𝑏 ≤ 𝑐 ≤ 0.25𝑏
where:
a and s are shown in Figure 10
b = beam flange width
d = beam flange depth
c = depth of cut
6.2.2.8 Bolted Stiffened End Plate Connection –
Summary of Requirements and Limitations
As can be noted from the above requirements, many of the
restrictions imposed on the BSEP connection are the same as
BUEP connection but there are a number of important differences.
See Appendix B for the summary of requirements and limitations
for the Bolted Stiffened End Plate Connection.
6.2.3 Reduced Beam Section
The following design procedure for the RBS connection has been
laid out with guidance from the CISC [1]. The procedure examines
the restrictions on the proportioning of the connection in no
particular order but they must all be satisfied for a safe connection.
Guidelines for continuity plates and welded joint design are
presented as well. The welded joint design includes the weld
details for the weld access holes, backing bars, and the welding
sequence for the bottom beam flange.
In addition, the reduced flange width should be equal to or less
than 14.6 times the flange thickness of the beam.
6.2.3.2 Check Reduced Section Resistance to
Applied Loads
Determine the factored resistance of the reduced section and check
its adequacy for the effect of the factored loads using Ze calculated
using c = 0.2b and be = 0.6b.
To preclude flexure failure of the connection the following
equation must be satisfied. Otherwise, increase c and recheck
conditions for the connection. The maximum value of c is 0.25b.
𝑀𝑐𝑓 ≤ 𝑅𝑦 𝐹𝑦 𝑍𝑏
where:
RyFy = 385 MPa and
Zb is the plastic section modulus of the gross beam section.
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bb = unreduced beam flange width
tb = beam flange thickness
RybFyb = the probable yield stress of the beam section
RycFyc = the probable yield stress of the column section
6.2.3.3 Connection Shear
The shear force at the column face should be calculated according
to the following equation.
2𝑀𝑐𝑓
𝑉𝑐𝑓 =
+ 𝑉𝑔
𝐿 − 𝑑𝑐
where:
Vg = shear due to companion gravity loads factored
L = length of span
For field-bolted shear tabs, the shear tab connection should be
designed for the previously defined shear force. The resistance
may be taken as nominal for bearing-type connections. This type
of connection needs to be connected to the column using either a
CPJ groove weld or full-depth fillet welds of 0.75 times the tab
plate thickness on each side of the tab plate.
6.2.3.4 Panel Zone Thickness
To prevent panel zone shear for the RBS connection, the panel
zone thickness should be proportioned the same as with bolted end
plate connections and outlined in Section 6.2.1.5.
6.2.3.5 Continuity Plates
Continuity plates are required if the column flange thickness does
not satisfy at least one of the following equations.
𝑅𝑦𝑏 𝐹𝑦𝑏
𝑡𝑐 ≥ 0.4√1.8𝑏𝑝 𝑡𝑏
𝑅𝑦𝑐 𝐹𝑦𝑐
and
𝑡𝑐 ≥
where:
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𝑏𝑏
6
BROOK ROBAZZA AND SHILIN SUN
6.2.3.6 Weld Joint Details
There are guidelines stipulated by the CISC for the use of weld
access holes, backing bars, and the welding sequence for the
bottom beam flange. These guidelines ensure that the welded
portions of the connections conform to a desired level of
performance.
6.2.3.7 Weld Access Holes
The weld access holes used in the connection design must meet the
dimensional requirements provided by FEMA-350 in Figure 3-5.
The access holes should be finished such that the surface
roughness is at most 13μm and there are no notches or gouges. The
welding should be done using controlled hydrogen electrodes that
meet H16 designator requirements. Following completion of the
welding, the area should be ground smooth and flush to meet the
contour and finish requirements for the access holes. To ensure
that the welds are free of cracks, weld access holes should be
inspected by means of magnetic particle testing or liquid penetrant
testing.
6.2.3.8 Backing Bars
Backing bars used when welding the bottom beam flange to the
column should be removed after welding. After the Northridge
earthquake it was found that there were failures stemming from the
failure to remove backing bars from the bottom beam flange region
and thus this new practice was acquired. The CISC provides
solutions for removing backing bars such as air carbon cutting,
grinding, chipping, or thermal cutting. A reinforcing fillet weld
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should be provided in these cases. This weld should have a
minimum leg size of the maximum of 8 mm or the root opening
plus 2 mm.
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6.2.3.9 Welding Sequence for Bottom Beam Flange
not yield at the levels expected. This fact is unconservative when
sizing adjacent members. This is a change from S16-01(R05)
which has a RyFy = 385 MPa for HSS sections. Since HSS sections
are not usually utilized in moment resisting connections however,
this change is of little consequence.
When the less desirable field welds are required, the CISC
recommends the following sequence to be used:
8.0 Conclusion
1) Weld start-stops directly under the beam web should be
avoided.
2) Each layer should be completed across the full width of the
flange before the next layer begins.
3) The start-stops of any two consecutive layers should not
fall on the same side of the beam web.
6.2.4.0 Reduced Beam Section Connection –
Summary of Requirements and Limitations
See Appendix C for the summary of requirements and limitations
for the Reduced Beam Section Connection.
7.0 Changes between S16-09 and S1601(R05) affecting Connection Design
There are few changes from the steel code upgrade from S1601(R05) to S16-09 which are specifically affecting connection
design. There are a number of changes which affect the beam and
column sections connecting into the connection but these only
indirectly affect the connection design. The only change that has
direct affect on the calculations is a change in Clause 27.1.5.1
which affects the calculation of the probable plastic bending
moment in the form of the Ry factor. For HSS members Ry has
increased to 1.37 (therefore a RyFy = 460 MPa) which reflects the
cold working of the plates producing higher strength steel that does
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Moment resisting connection design has been recognized as a
crucial part of designing an adequate seismic force-resisting
system. Although researchers and practicing engineers have made
substantial revisions to the seismic design of moment resisting
connections after the Northridge and Kobe earthquakes, there still
remains some question about the adequacy of the revisions and the
new types of connections being used. As could be seen from the
connections that failed during these two major seismic events,
testing is not always representative of true engineering practice. In
order to prove that these connections are truly adequate, further
analysis must be undertaken to study the performance of these
prequalified moment resisting connections as well as the procedure
of their design. Since the three connections investigated in this
report are only a few of the numerous connections available, the
reader should look into FEMA 350 which provides many more
connection types and corresponding design procedures.
The procedure used for the design is implemented into a flexible
tool that has been developed into a spreadsheet program. This
program helps the designer design connections by not only
determining the strength and adequacy of a proposed connection
configuration but also provides an estimation of the efficiency of
the connection. Three types of common prequalified moment
resisting connections are chosen which are representative of
Canadian construction. The spreadsheet tool incorporates each of
these connection type design procedures to enable the designer to
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MOMENT RESISTING CONNECTIONS FOR SEISMIC APPLICATIONS
consider several types of connection in their design. This attached
design tool provides the designer with substantial assistance in the
design of moment resisting connections for seismic applications.
9.0 References
[1] Canadian Institute of Steel Construction (2008). “Moment
Connections for Seismic Applications.” Lakeside Group Inc.
Ontario, Canada.
[2] Federal Emergency Management Agency (2000). “FEMA 350
3-48 Moment Resisting Connections.” Retrieved from
www.nehrp.gov. SAC Joint Venture. California, U.S.A.
[3] Federal Emergency Management Agency (2000).
“Recommended Seismic Design Criteria for New Steel MomentFrame Building: FEMA 350.” SAC Joint Venture. California,
U.S.A.
[4] Wilkinson, S., Hurdman, G., & Crowther, A. (2006). “A
Moment Resisting Connection for Earthquake Resistant
Structures.” Journal of Constructional Steel Research Volume 62.
BROOK ROBAZZA AND SHILIN SUN
[7] Rizkalla, S. H., & Thadani, B. N. (1983). “Structural Steel
Design.” Cantext Publications. Manitoba, Canada.
[8] Stiemer, S. F. (2011) “Connection Design and Cost Estimate.”
University of British Columbia, British Columbia, Canada.
[9] Satishkumar, S. R. (2005) “Design Steel Structure I.” NPTEL
Chennai, India.
[10] Metten, A. (2010). “Seismic Resisting Elements – Moment
Frames and Other Systems.” Bush, Bohlman& Partners. British
Columbia, Canada.
[11] Eliasova, M., Gomes, F. C. T., & Wald, F. (2002). “Stiffened
End Plates in Structural Steel Connections.”
[12] Adan, S. M., & Hamburger, R. O., (2010). “Steel Special
Moment Frames.” A Joint Publication of NCSEA, CASE and SEI.
[13] Steel Photo Gallery (2011). Retrieved from_http://www.meritsteel.com/PhotoGallery.html
[14] Canada Institute of Steel Construction (2010). “Steel
Handbook – Tenth Edition.” CISC Publications
[5] Pachoumis, D. T., Galoussis, C. N., & Christitsas A. D. (2009).
“Reduced Beam Section Moment Connections Subjected to Cyclic
Loading: Experimental Analysis and FEM Simulation.”
Engineering Structures Volume 31.
[15] Sehgal, V. K., Setia, S., & Murty, C. V. (2009). “Seismic
Behavior of Weak-axis Beam-column Connections.” CRC Press
Online.
[6] Institute for Steel Development & Growth (2006). “Earthquake
Resistant Design of Steel Structures.” Institute for Steel
Development & Growth. Kolkata, India.
[16] Unsworth, J. F. (2010). “Design of Connection for Steel
Members.” CRC Press Online.
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MOMENT RESISTING CONNECTIONS FOR SEISMIC APPLICATIONS
[17] Stiemer, S. F. (2011)“Plastic Design with Computers,”
retrieved from_http://www.sigi.ca/engineering/notes/plastic_
design_with_computers.pdf.
BROOK ROBAZZA AND SHILIN SUN
investigated connections. Similar tables of requirements and
limitations are also provided by FEMA 350.
[18] Stiemer, S. F. (2011)“Plastic Methods,” retrieved
from_http://www.sigi.ca/engineering/notes/plastic_
mathods.pdf.
[19] Federal Emergency Management Agency (FEMA)-355D,
(2000). “State of the Art Report on Connection Performance”
retrieved from_http://www.nehrp.gov/pdf/fema355d.pdf.
Appendix A: Bolted Unstiffened End Plate Connection –
Summary of Requirements and Limitations:
[20] Federal Emergency Management Agency (FEMA)-351,
(2000), “Recommended Seismic Evaluation and Upgrade Criteria
for Existing Welded Steel Moment-Frame Buildings” retrieved
from_http://www.nehrp.gov/pdf/fema355d.pdf.
[21] Charney, F. A., Downs, W. M., (2004). “Modeling Procedures
for Panel Zone Deformations in Moment Resisting Frames.”
Department of Civil and Environmental Engineering, Virginia
Tech, Blacksburg, Virginia, U.S.A.
[22] Cary, Howard B; Scott C. Helzer (2005). Modern Welding
Technology. Upper Saddle River, New Jersey, U.S.A.
[23] ASM International (2003). Trends in Welding Research.
Materials Park, Ohio: ASM International.
10.0 Appendices
The following appendices are tables from the CISC guidelines for
the connections investigated in this report. These tables provide a
summary of the requirements and limitations that are impost on the
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Appendix B: Bolted Stiffened End Plate Connection –
Summary of Requirements and Limitations:
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Appendix C: Reduced Beam Section Connection – Summary of
Requirements and Limitations:
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