Geotechnical Engineering Guide: Piling, Soil Improvement & Support

F O U RT H E D I T I O N 2008 F O U RT H E D I T I O N 2008 A Guide to Practical GEOTECHNICAL ENGINEERING in Southern Africa FIRST EDITION January 1976 written and compiled by IH Braatvedt Pr Eng, BSc (Eng), MICE, FSAICE SECOND EDITION December 1986 revised and updated by JP Everett Pr Eng, BSc (Eng), FSAICE THIRD EDITION July 1995 re-written and updated by G Byrne Pr Eng, BSc (Eng), MSAICE JP Everett Pr Eng, BSc (Eng), FSAICE K Schwartz Pr Eng, BSc (Eng), GDE, FSAICE assisted by EA Friedlaender Pr Eng, BSc (Eng), MSAICE N Mackintosh NH Dip (Civ Eng), MSAICE C Wetter BSc (Eng) FOURTH EDITION December 2008 revised and updated by G Byrne Pr Eng, BSc (Eng), FSAICE AD Berry Pr Eng, BSc (Eng), MEng, MSAICE i THE PURPOSE OF THIS BOOK When Franki Africa first published ‘The Guide’ in 1976 the main purpose was to create a practical reference on all aspects of soil investigation and piling as carried out by the company in southern Africa at that time. Judging from the popularity of the first edition this objective was achieved and most design engineers in southern Africa have a copy on their bookshelves. The second edition was published in 1986, as an update of the first, and it was equally popular, followed by the Franki ‘Blue Book’, as the third edition printed in 1995 soon become known. This, the fourth edition, is an update of the third edition of the book as Franki has expanded its activities further into soil improvement with the addition of the highly versatile jet grouting technology, and has enhanced its range of piling technology with the introduction of the Full Displacement Screwpile in 2008. Franki has also extended its already wide range of equipment with the purchase of several new piling rigs, including the very successful and versatile Rotapiling rig as well as hydraulically operated piling rigs capable of installing piles of up to 2 metres in diameter and to depths in excess of 50 metres. In addition, several new techniques of anchoring and micropiling have been added to Franki’s extensive product line. Many of the standard load tables have been updated to reflect current practice and data sheets have been included for ease of reference. A major addition to this fourth edition is the inclusion of a section on marine engineering, an area of significant development for the company over the last ten years, having recently completed various quays and jetties in the Seychelles, Mauritius, Angola and Mozambique. The purpose of this book is essentially the same as the third edition, and it remains a reference book containing a wealth of practical information on geotechnical engineering. The contents of this book are presented in good faith. As in all geotechnical design, the methods and data presented in the book must be interpreted and used with a degree of knowledge, experience and judgement. Franki Africa (Pty) Ltd does not hold itself in any way responsible for any inaccuracies or errors in the book, or for any interpretation thereof by persons other than its own employees. The company acknowledges, with appreciation, the contribution by ARUP to the section on pile-cap design. Acknowledgement is also due to Graeme Wray of Franki, for scanning in the text from the third edition while on sick leave. The contribution from Peter Day of Messrs Jones and Wagener on Limit State Design is greatly appreciated. Design, layout, reproduction and print – VIVO Design Associates. ii FOREWORD by IAN BRAATVEDT Just over thirty years ago (1976) the original ‘Frankipile Guide to Piling and Foundation Systems’ was published. It became a standard text for those in the industry of foundation engineering in southern Africa and elsewhere. Every ten years since that time the company has updated and revised the book. This is a glowing reflection not only of the expanding range of products and services that the company offers, but also of the high level of geotechnical knowledge and experience that it possesses. One identifies this process of updating and revising initially with John Everett and latterly with Gavin Byrne and their editorial teams. The latest edition (4th) is a highly professional but still practical guide and reference to those who seek the most economical and safe solution to a geotechnical problem. When one looks back and recognises the ongoing process within Franki of improved skills, knowledge and professionalism, one may be entitled to reword the following definition of Geotechnical Engineering jokingly put forward by Professor Noel Symons thirty or more years ago: ‘ The art of using soils whose properties we do not really know, or understand, to form and support structures we cannot really analyse, so as to withstand forces which we cannot really assess, in such a way that the public does not really suspect.’ This book, the fourth edition of the original guide will help the practicing engineer to eliminate the words ‘not really’ and ‘cannot really’ in the above cynical definition of geotechnical engineering; and as the author of the original edition, it is with admiration and respect that I commend the use of this book. IAN BRAATVEDT Pr Eng Johannesburg, August 2007 iii Hillside Smelter Phase Two, Richards Bay, South Africa iv CONTENTS PAGE 1.0 FRANKI AFRICA (PTY) LIMITED 1 2.0 GEOTECHNICAL INVESTIGATION 2.1 Guide for Planning a Geotechnical Investigation 2.2 Field Investigation Techniques 2.3 Geotechnical Engineering Laboratory Services 2.4 Site Investigation for Piling 2.5 Site Investigation for Lateral Support 2.6 Site Investigation for Soil Improvement Applicable Norms 4 5 9 28 30 30 30 30 3.0 SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS 3.1 Notes on Soil Profiling 3.2 Notes on Rock Mass Description 3.3 Interpretation of Geotechnical Investigation and Laboratory / In-situ Testing Data 3.4 Permeability Applicable Norms 31 31 38 44 4.0 FACTORS INFLUENCING THE SELECTION OF A PILE TYPE 65 5.0 CLASSIFICATION OF PILING SYSTEMS 67 6.0 SUMMARY DETAILS OF PILING SYSTEMS 68 7.0 TECHNICAL DETAILS OF PILING SYSTEMS 7.1 Franki Driven Cast-in-situ Piles 7.2 Driven Tube Piles 7.3 Precast Piles 7.4 Steel H-Piles 7.5 Auger Piles 7.6 Underslurry Piles 7.7 Continuous Flight Auger (CFA) Piles 7.8 Full Displacement Screwpiles 7.9 Forum Bored Piles 7.10 Oscillator Piles 7.11 Rotapiles 7.12 Micropiles Applicable Norms 70 70 80 86 93 97 105 114 119 124 129 134 138 142 8.0 UNDERPINNING 8.1 Old Foundation removed and new Foundation provided 8.2 New Footing located under the existing one 8.3 Jack-piles under the existing Foundation 8.4 Piles alongside the existing Foundation 8.5 New piled Foundation and Column 8.6 Piles through existing Foundation 8.7 Underpinning using Micropiles 8.8 Underpinning using Jet Grouting 143 145 146 148 150 153 153 154 155 9.0 PILE LOAD AND INTEGRITY TESTING 9.1 Pile Load Testing 9.2 Integrity Testing of Piles 156 156 161 v 64 64 10.0 FACTORS INFLUENCING THE SELECTION OF A SOIL IMPROVEMENT SYSTEM 165 11.0 CLASSIFICATION OF SOIL IMPROVEMENT SYSTEMS 167 12.0 SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS 168 13.0 TECHNICAL DETAILS OF SOIL IMPROVEMENT SYSTEMS 13.1 Vibratory Compaction 13.2 Dynamic Compaction 13.3 Compaction Grouting 13.4 Vibratory Replacement 13.5 Dynamic Replacement 13.6 Driven Stone Columns 13.7 Accelerated Consolidation 13.8 Jet Grouting 13.9 Deep Soil Mixing 13.10 Cutter Soil Mixing (CSM) Applicable Norms 170 170 175 182 184 187 190 192 195 199 201 201 14.0 FACTORS INFLUENCING THE SELECTION OF A LATERAL SUPPORT SYSTEM 202 15.0 CLASSIFICATION OF LATERAL SUPPORT SYSTEMS 15.1 Embedded Walls 15.2 Reinforced Soils 204 204 206 16.0 SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS 208 17.0 TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS 17.1 Embedded Walls 17.2 Embedded Wall Support Systems 17.3 Reinforced Soils Applicable Norms 210 210 232 241 249 18.0 PROBLEM SOILS AND THEIR FOUNDATION SOLUTIONS 18.1 Expansive Soils 18.2 Collapsible Soils 18.3 Soft Clays 18.4 Dolomites 18.5 Dispersive Soils 18.6 Liquefiable Soils 250 251 256 259 261 263 263 19.0 ENVIRONMENTAL ENGINEERING 19.1 Groundwater Monitoring 19.2 Monitoring of Surface Water 19.3 Containment / Remediation 264 264 265 266 20.0 MARINE FOUNDATION ENGINEERING 20.1 Piling for Marine Structures 20.2 Earth Retention for Marine Structures 20.3 Soil Improvement for Marine Applications 20.4 Construction Methods for Quays and Jetties 20.5 Rehabilitation of Quays and Jetties Applicable Norms 268 268 272 282 284 287 287 vi 21.0 DESIGN AIDS: PILING 21.1 Pile Capacity to Resist Compressive Load 21.2 Pile Capacity to Resist Uplift Load 21.3 Pile Capacity to Resist Lateral Load 21.4 The Design of Piles for Heaving Subsoil Conditions 21.5 Factors of Safety 21.6 Limit State Design in Geotechnical Engineering 21.7 Analysis and Design of Pile Groups 21.8 Settlement of a Single Pile and Pile Groups 21.9 Structural Design of Pile Shafts Applicable Norms 21.10 Structural Design of Pile-caps 288 288 308 310 314 317 319 322 325 334 338 339 22.0 DESIGN AIDS: SOIL IMPROVEMENT 22.1 Soil Compaction 22.2 Soil Replacement 22.3 Accelerated Consolidation Applicable Norms 356 356 363 367 368 23.0 DESIGN AIDS: LATERAL SUPPORT 23.1 Geotechnical Design Parameters 23.2 Earth Pressures 23.3 Water Pressures and Surcharge Loads 23.4 Embedded Walls 23.5 Reinforced Soils 23.6 Factors of Safety 23.7 Movements Associated with Excavations Applicable Norms 369 369 371 374 376 384 389 390 391 24.0 QUALITY ASSURANCE AND SAFETY 24.1 Level 1 24.2 Level 2 24.3 Level 3 24.4 ISO 9001 Certification 392 392 393 393 395 25.0 REFERENCE INFORMATION 25.1 Normal Plant Clearance Requirements 25.2 Rig Dimensions 396 396 398 REFERENCES 411 INDEX 417 LIST OF SYMBOLS 422 vii Construction of Fishing Quay, Mahe, Seychelles viii 1.0 FRANKI AFRICA (PTY) LIMITED The South African company, formerly a key member of the worldwide Franki group, was started by Wally Rowland in 1946. The initial seeds had already been sown early in 1939 but the second World War broke out in September and Wally joined up with the South African forces. At the end of the war The Franki Piling Company of South Africa, as it was initially named, was registered and the first contract secured. This involved the installation of eight piles for a building in Paarden Eiland and a steam driven piling machine was used to install the piles which were standard Franki driven cast-in-situ piles. By 1952 Franki had branch offices in Johannesburg, Cape Town and Durban. In 1955 Wally Rowland returned to the UK to take up the position of Assistant Works Director with the British Franki company. Ian Braatvedt took over as Managing Director in 1961 and under his guidance the company grew steadily. Large contracts such as the Alusaf Bayside Smelter in Richards Bay, the Mondi Paper Mill in Durban and Iscor Steelworks in Newcastle were secured in the late sixties and early seventies and these really helped Franki to establish itself as the leading piling company in South Africa. In 1968 Franki started a soil investigation subsidiary which is known as Soiltech. Today it has a full complement of soil investigation and field testing equipment. It also plays a role in the environmental investigation field. The need to diversify into other geotechnical fields led to the formation of GeoFranki in 1987. GeoFranki's main areas of activity are lateral support, ground improvement, micropiling, grouting and cut-off walls. At the same time as the diversification into Geofranki, Franki initiated a full in-house geotechnical design capability. This key function was expanded and culminated in Franki achieving ISO 9001 accreditation as part of its full specialist geotechnical design and construct service. Significant recent advances in extending Franki’s geotechnical capabilities include the addition of Jet Grouting, the Franki Rotapile for difficult ground conditions (bouldery riverbeds and Dolomitic Karst conditions), Micropiles, self-boring anchors and most recently full displacement Screwpiles. These additions to Franki’s already extensive capability keep the company up to date with the latest international technology. In 1999 the worldwide Franki organisation disbanded and Franki became a locally owned organisation operating in sub-Saharan Africa and the Indian Ocean Islands. In 2004 the company changed its name to Franki Africa (Pty) Limited and presently has over fifty major production rigs and an employee complement of approximately 900. It has offices in the main centres of Johannesburg, Cape Town and Durban in South Africa and regional offices in Luanda (Angola), Port Louis (Mauritius), Maputo (Mozambique) and Dar es Salaam (Tanzania). 1 The following is a summary of the products and services which Franki Africa can presently offer its clients, and which are described in greater detail in this guide. .. .. .. .. .. .. .. .. .. . .. .. .. .. .. .. .. .. .. .. .. .. .. .. . . . FIELD INVESTIGATION Auger Trial Holes Test Pits Bulk and Undisturbed Soil Sampling Dynamic Cone Penetration Tests Cone Penetration Tests Piezocone Tests Rotary Core Drilling In-situ Testing Lugeon Tests Piezometer Installations Shelby and Piston Tube Sampling Core Orientation Rotary Percussion Drilling Plate Load Tests In-situ Density Tests Geophysical Techniques Groundwater Monitoring Well Installations Environmental Investigation UNDERPINNING Foundation Replacement New Foundation under Existing Foundation Jack-piles Piles Alongside Existing Foundation New Piled Foundation and Column Piles Through Existing Foundation Micropiles Jet Grouting SOIL IMPROVEMENT Vibratory Compaction Dynamic Compaction Compaction Grouting Vibratory Replacement Dynamic Replacement Driven Stone Columns Accelerated Consolidation Jet Grouting LABORATORY TESTING Franki have discontinued their in-house laboratory service, but will deliver samples and arrange all required laboratory testing with local accredited laboratories. .. .. .. .. .. .. LATERAL SUPPORT Steel and Concrete Sheet Pile Walls Steel Soldier Pile Walls Concrete Soldier Pile Walls Contiguous /Secant Pile Walls Diaphragm Walls Geonails Reticulated Micropiles Soil Dowelling Tie-back Anchors Self-drilling Anchors PILING Franki Cast-in-Situ Piles Driven Tube Piles Precast Piles Steel H-Piles Auger Piles Underslurry Piles Continuous Flight Auger (CFA) Piles Full Displacement Screwpiles Forum Bored Piles Oscillator Piles Rotapiles Micropiles MARINE STRUCTURES Piling for Marine Structures Soil Retention for Marine Structures Soil Improvement for Marine Applications Construction Methods for Quays and Jetties Rehabilitation of Quays and Jetties 2 Franki Africa has always adopted a policy of combining innovative design with many years of practical experience to provide the most economical solution to a geotechnical problem. The company thus maintains a strong design capability as well as its professionally run geotechnical contracting activities. With this considerable expertise and product range, Franki offers a turnkey solution including investigating the site, the full design and detailing of the foundation system and any lateral support requirements, pricing and contract documentation, execution of the work and final handover. The recent turnkey marine projects successfully completed by Franki demonstrate its ability to deliver worldclass design and construct contracts efficiently and economically. CNOI Drydock, Port Louis, Mauritius 3 2.0 GEOTECHNICAL INVESTIGATION Soiltech, the division of Franki Africa responsible for Geotechnical Investigations, was established in 1968, and offers a complete Geotechnical Service to consulting engineers and client organisations, as well as to the company. The importance of obtaining adequate and reliable knowledge of sub-surface conditions at a sufficiently early stage cannot be over-emphasized when considering: .. .. .. . The choice and design of an economical and technically sound foundation Possible delays and additional expense due to inadequate soils information Expensive foundation failures or overdesign Potential contractor's claims based on inaccurate and/or inadequate soils information Soiltech is able to offer a complete geotechnical investigation service comprising: Planning of the investigation Execution of the field-work and management of laboratory testing Interpretation and reporting The range of field-work and laboratory testing that Soiltech can offer is outlined below: .. .. .. .. .. .. .. .. .. .. FIELD INVESTIGATION AND IN-SITU TESTING Auger Trial Holes Test Pits Bulk and Undisturbed Soil Sampling Dynamic Cone Penetration Tests Cone Penetration Tests Piezocone Tests Rotary Core Drilling Standard Penetration Tests Vane Shear Tests Pressuremeter Tests Lugeon Tests Piezometer Installations Shelby and Piston Tube Sampling Core Orientation Rotary Percussion Drilling Vertical and Horizontal Plate Load Tests In-situ Density Tests Geophysical Techniques Well Installations Groundwater Monitoring and Sampling LABORATORY TESTING Soiltech have discontinued their in-house laboratory service, but will deliver samples and arrange all required laboratory testing with local accredited laboratories. All site investigations should be undertaken in accordance with the SAICE Draft Code of Practice for Site Investigations (2007). Another important aspect of Soiltech's activities lies in the environmental engineering field. This service provides for the collection of data with respect to potentially contaminated soils, surface and groundwaters. For further details refer to SECTION 19.0: ENVIRONMENTAL ENGINEERING. 4 2.1 GUIDE FOR PLANNING A GEOTECHNICAL INVESTIGATION Guidance on planning a geotechnical investigation is given in the SAICE Draft Code of Practice for Site Investigations (2007) and may embrace any combination of the following: .. . . . . . .. .. . . To assess the general suitability of the site for the proposed engineering works To enable an adequate and economical design to be prepared To foresee and provide against difficulties that may arise during construction due to ground and other local conditions To determine the causes of defects or failure in existing works and the remedial measures required To advise on the availability and suitability of local materials for construction purposes Taking the above objectives into consideration, the planning of a geotechnical investigation will be influenced by the following main factors: Depth of Investigation: Knowing the nature of the proposed engineering development will allow assessment of the likely foundation types and sizes which determine the magnitude of applied stress and depth of influence of the foundations or structure, and thus determines the required depth of investigation; Lateral Extent of the Investigation: A knowledge of the geology and geomorphology of the site influences the frequency of probing, along with structural sensitivity and importance of the proposed development; Critical Design Parameters: Soil stiffness/shear strength/permeability Access to and the remoteness of the site Site topography, vegetation and drainage Nature and proximity of adjacent developments Knowledge of previous geotechnical investigations or foundation installations carried out in the area. In particular, the opinions of local engineers, farmers and contractors Evidence of problem soil conditions (expansive or collapsible soils, dolomites, dispersive soils, soft clays) The cost of an adequate investigation is very low in comparison to the total cost of the project. The consequences of not providing sufficient, accurate, and reliable geotechnical information can have a significant effect on a project and can lead to delays and extras during construction, with associated costly claims. The planning of a geotechnical investigation should be carried out in a phased approach. Phase one is an initial investigation to determine the site geology and to define the problem. This is followed by phase two which is a far more extensive investigation in which the site geology is studied in greater detail and all the critical design parameters are determined. The phased approach will generally commence with a desk study and site reconnaissance, followed by the field-work and high quality laboratory testing. 5 Conditions vary from site to site resulting in a variety of techniques that have been developed to enable both the geotechnical engineer and specialist contractor to select the appropriate investigation procedures. An accurate description of the soil profile forms the basis of the geotechnical investigation. In some cases this may be all that is required. However, in the majority of investigations it will be necessary to supplement an accurate description of the soil profile with appropriate in-situ testing and sampling, and associated laboratory testing. Under appropriate conditions, particularly where the water-table is at depth, (which is applicable to large areas of the interior of southern Africa), the drilling of large diameter trial holes and / or the forming of test pits for visual inspection by a geotechnical engineer or engineering geologist, can be carried out. The advantages of this procedure are as follows: .. . . It allows for the soil profile to be examined in-situ in its natural state Good quality undisturbed block samples can be cut from the auger hole or test pit side-walls. Disturbed samples can also be taken from specific horizons identified during profiling In-situ testing such as hand shear vane tests and horizontal plate bearing tests can be executed within the trial holes or test pits The procedures adopted are fast and economical and provide for accurate and comprehensive evaluation of site geotechnical conditions Safety procedures when profiling and sampling in trial holes and test pits are extremely important. All investigation work using trial holes and test pits must be carried out in accordance with the SAICE Code of Practice for the Safety of Persons Working in Small Diameter Shafts and Test Pits for Civil Engineering Purposes (2007). For certain projects it may will be necessary to supplement the auger trial holes/test pits with additional investigation procedures. A variety of techniques are available. These could include dynamic cone penetration tests, rotary core drilling with associated sampling and in-situ testing techniques (standard penetration tests, vane shear tests, lugeon tests etc). In the coastal regions and on sites with a high water-table the use of trial holes and test pits is often not feasible due to the possible collapse of the side-walls. In these areas the two standard methods used in a geotechnical investigation are boreholes with Standard Penetration Tests and Cone Penetration Tests. These are supplemented where necessary with, amongst others, rotary core drilling, piezocone and vane shear tests, dynamic cone penetrometer tests and the recovery of undisturbed samples using the Shelby tube method or piston sampling. TABLES 2.1.1 and 2.1.2 are provided as a guide to assist in the planning of a geotechnical investigation. These tables give typical details with regard to the field and laboratory tests that could be carried out in stable soil profiles above the watertable (TABLE 2.1.1) and in saturated variable soils (TABLE 2.1.2). 6 TABLE 2.1.1 Guide to Planning a Soils Investigation in Stable Soil Profiles above the Water-table (usually Residual Soils or Cohesive Transported Soils) Parameter Field Test / Requirement Laboratory Test Description of the Soil Profile Auger trial holes Test pits Boreholes with SPT Seismic survey — Consistency of the Soil Profile In-situ tests (DPSH/CPT/SPT/CPTU) In-situ profiling of trial holes/ test pits Sand replacement tests Density of undisturbed samples (oedometer) Undrained Shear Strength Recover undisturbed samples from auger trial hole, test pit or borehole; Vane shear test in borehole or trial hole Undrained triaxial test Unconfined compression test Drained Shear Strength: Effective angle of friction - φ‘ Effective cohesion - c’ Recover undisturbed samples from auger trial hole, test pit or borehole Drained triaxial test Drained shear box test Undrained triaxial test with measurement of pore water pressure Modulus of Compressibility (Stiffness at appropriate strain level) Cross-hole jacking test Plate load test Pressuremeter test Small strain stiffness - SASW Dilatometer Oedometer test Triaxial test with local strain measurement Bender element Index Property Tests Recover disturbed samples from auger trial hole, test pit or borehole Grading analysis Atterberg limits Moisture content Permeability Recover undisturbed samples from auger trial hole, test pit or borehole, CPTU, Lugeon test Falling or constant head permeability Collapse Recover undisturbed samples from auger trial hole, test pit or borehole Double oedometer Collapse potential test Heave Recover undisturbed and /or disturbed samples from auger trial hole, test pit or borehole Double oedometer Swell under load test Index property test (disturbed sample) Level of Water-table Drill a trial hole or a borehole, leave for a period of time for the water-level to stabilise in the hole and then measure the level Degree of saturation Soil Suction Pressures Filter paper test — 7 TABLE 2.1.2 Guide to Planning a Soils Investigation in Saturated, Variable Soils (usually encountered in Coastal Areas or Adjacent to Watercourses) Parameter Field Test / Requirement Laboratory Test Description of the Soil Profile Boreholes with SPT and/or rotary drilled cores — Consistency of the Soil Profile Dynamic Cone Penetrometer (DPSH) Cone Penetrometer Test (CPT/CPTU) Boreholes with SPT — Undrained Shear Strength Recover undisturbed samples from borehole, Vane shear test in borehole, correlate with in-situ penetrometer tests Undrained triaxial test Unconfined compression test Drained Shear Strength: Effective angle of friction - φ‘ Effective cohesion - c’ Recover undisturbed samples from borehole, correlate with in-situ penetrometer tests (sandy soils only) Drained triaxial test Drained shear box test Undrained triaxial test with measurement of pore water pressure Modulus of Compressibility (Stiffness at appropriate strain level) Pressuremeter test, correlate with in-situ penetrometer tests, Small strain stiffness - SASW Oedometer test Triaxial test with local strain measurement Index Property Tests Recover disturbed samples from borehole Grading analysis Atterberg limits Moisture content Permeability Recover undisturbed samples from borehole, CPTU, Lugeon test Falling or constant head permeability Collapse Recover undisturbed samples from borehole Double oedometer Collapse potential test Heave Recover undisturbed and /or disturbed samples from borehole Double oedometer Swell under load test Index property test (disturbed sample) Drill a borehole and install piezometer — Level of Water-table 8 2.2 FIELD INVESTIGATION TECHNIQUES The field investigation techniques used by Soiltech to determine the geotechnical subsurface conditions are summarised on page 4, and are discussed in more detail below. 2.2.1 AUGER TRIAL HOLES The auger trial hole involves the drilling of a large diameter auger hole using a powerful auger machine. A qualified person is then lowered in stages down the hole by means of a small winch and is often able to profile the hole by visually inspecting the side-walls and the base. Furthermore, it is possible to cut large undisturbed samples from the side-walls or base of the hole for testing in the laboratory, as well as, carrying out cross-hole jacking tests and plate load tests, as described in SECTION 2.2.7 in the trial hole excavation. Bulk sampling for the purposes of evaluating the mineral content of materials on old dumps, or within soil and weathered profiles, can also be accomplished using this technique. Auger trial holes provide a very quick and economical method for obtaining reliable geotechnical information for a variety of engineering solutions and this method is favoured by most engineers and geologists. For the successful application of this technique it is essential that the side-walls of the trial holes remain stable during drilling and profiling. It is thus not suited to areas with a high water-table where the collapse of the side-walls is most likely to occur. It is possible to drill a large number of trial holes in a relatively short space of time which makes this an economical form of investigation. A minimum hole diameter of 750 mm is required for in-situ profiling purposes, but trial holes of up to 2 metres in diameter are possible. Depths of up to 36 metres can be drilled in suitable materials. The technique is ideally suited to sites with deeply weathered profiles. The auger trial hole can also penetrate into soft rock and even hard fractured rock. Under suitable site conditions approximately eight 750 mm diameter auger holes to a depth of 10 metres can be drilled within a normal working day. It is also possible to profile this number of holes within the same working day. The auger rig with its crew and ancilliary equipment is normally hired on a daily basis. Soiltech can arrange for the profiling of the holes by experienced qualified personnel from an independent geotechnical engineering firm should this be required. To facilitate the profiling of the trial holes, a tripod frame fitted with a winch is positioned over the trial hole, the winch being connected via a steel wire rope to a specially designed bosun's chair. All operations are carried out in accordance with the SAICE Code of Practice for the Safety of Persons Working in Small Diameter Shafts and Test Pits for Geotechnical Engineering Purposes (2007). Plastic sample bags, clingwrap, labels, tape measures and sampling tools form part of the standard equipment available on site. Under special site circumstances breathing apparatus and methanometers are made available on site. 9 Auger Rigs for Drilling Trial Holes Soiltech has a variety of truck mounted auger rigs available for drilling trial holes. PLATE 2.2.1.1 shows a Williams LLDH 120 rig used for drilling auger trial holes. Various track mounted auger rigs are also available for the drilling of auger holes of between 0.75 to 2 metres in diameter, with a depth range of between 16 and 48 metres for the newer machines. The smaller hole sizes are generally preferred as this improves hole stability. Rigs weigh between 24 to 45 tonnes and rig dimensions are approximately 3 metres wide and 12 metres long. Most rig dimensions are given in SECTION 25.2. 2.2.2 TEST PITS The use of test pits as an investigation technique offers the same advantages in terms of profiling and sampling as described for auger trial holes. Test pits are easily formed with a mechanical excavator, or by hand, and therefore have the advantage of being relatively inexpensive. The main disadvantages are that they are limited to depths of two to three metres and cannot be used in areas with a shallow water-table. Test pits are therefore most appropriate in areas with a relatively deep water-table where competent soils or rock are anticipated at a relatively shallow depth. They are often used to investigate areas where there is poor access for other types of equipment. In view of cost advantages, test pits are often used in a preliminary or first phase of an investigation. Where a competent soil or rock stratum occurs close to the ground surface, the profiling and sampling of test pits may provide sufficient information for design purposes and no other form of testing is required. If the excavation of test pits discloses a much deeper soil profile, then it is essential to follow up the first phase with additional investigation work. This is normally carried out using techniques which can reach greater depths, such as auger trial holes and boreholes. It is extremely important to follow the correct safety procedures when profiling and sampling in test pits, and the SAICE Code of Practice for the Safety of Persons Working in Small Diameter Shafts and Test Pits for Geotechnical Engineering Purposes (2007) should be strictly adhered to at all times. Experience has shown that test pits are far more prone to collapse than auger trial holes, due to the fact that a rectangular pit is less stable than a circular trial hole. Even highly experienced engineers and geologists find it difficult to assess the stability of a test pit and serious accidents have been reported. Where a deep foundation solution is anticipated from the first phase of the geotechnical investigation (desk study), the information provided from shallow test pits is insufficient for the design of a deep foundation solution. It is essential that trial holes and/or boreholes are carried out to facilitate an economical and successfull foundation design. 10 PLATE 2.2.1.1 A Williams LLDH 120 Rig for Drilling Auger Trial Holes 11 2.2.3 ROTARY DRILLING AND SAMPLING The rotary drilling technique is used to drill a borehole which is normally cased through the upper soil profile. Various methods for testing and sampling the soil during the drilling of the borehole are available and described later in this section. The Standard Penetration Test (SPT) was originally developed as a sampling technique but later progressed into a more versatile correlation tool as well. Once the borehole reaches strata of rock consistency, rotary core drilling is used to recover samples. Rotary Drilling The borehole is typically drilled through the upper soil layers using a casing fitted with a diamond / tungsten tipped casing shoe. A drilling fluid is used to remove the cuttings and flush them to the surface where they can be sampled. This technique for advancing the borehole is called wash boring and the samples are known as wash samples. The borehole is advanced in stages with samples taken at the various depths required. PLATES 2.2.3.1 and 2.2.3.2 show two types of rotary drilling rigs. When materials of rock consistency are encountered and wash boring is no longer effective, rotary core drilling is used to advance the borehole and recover core samples. The cores are drilled using a core barrel which is fitted with a diamond tipped or impregnated drill crown. The core barrel with drill crown is rotated by the drilling rig which also has the means to hydraulically crowd the drill stem. A drilling fluid is pumped through the core barrel to cool the drill bit and flush the cuttings to the surface. The conventional core barrel can recover a 1.5 metre length of core at a time. Once the core barrel is full, the drill stem with core barrel is withdrawn from the hole and the core sample is recovered and stored in a core box. Core boxes are marked with the depths drilled so that a visual inspection of the core box shows what percentage of core was recovered relative to the depth drilled. Cores are sometimes waxed to retain their natural moisture content. Unconfined Compressive Strength Tests (UCS) and Point Load Tests (PLT) are often carried out on rock cores to determine the strength of the rock. This is an important factor when carrying out a geotechnical investigation for a contract on which piles will be required to penetrate the rock, as the piling contractors need to know the hardness of the rock to be able to assess penetration rates at the time of tender. Heinz (1989) gives a detailed description of rotary core drilling techniques and equipment. Soiltech complies with the Standard Specifications for Sub-surface Investigations (CSRA, 1993) in carrying out rotary drilling operations. Sonic Drilling Sonic drilling can be used where there is risk of pneumatic fracture, such as in dam walls. Piezometer Installations Piezometers are installed in boreholes in order to provide information regarding the at rest levels of the water-table. In addition, groundwater pressure can be measured via more specialised piezometers ie hydraulic, electrical and pneumatic. In general piezometers are installed into pre-cleaned holes by lowering a selected porous tip to approximately 500 mm above the bottom. The tip is surrounded by a filter of graded, washed silica sand and sealed off with a bentonite plug. The remainder of the borehole is sealed by introducing cement/bentonite grout. 12 SAMPLING Shelby and Piston Tube Sampling This sampling technique is employed to obtain undisturbed material from soft and very soft cohesive soils. The Shelby tube, used to recover the samples, consists of a thin walled stainless steel tube with an internal diameter of approximately 75 mm. The leading edge of the tube is bevelled and crimped in such a way that the entry diameter is fractionally smaller than the body diameter. The tube is usually a half metre in length with the top end designed to fit into an adapter. The adapter has a one-way valve built into it to allow water to escape so as to prevent compression of the sample. The Shelby tube sampler is attached to the drill string, in place of a core barrel, and lowered to the base of the borehole where it is pressed into the soft material using the drill rig hydraulics. The sample and tube are then raised and the sample extruded on site. The sample should be sealed and packed so as to maintain the in-situ moisture content and to resist damage during normal handling and transport. Under certain conditions, where the material to be sampled cannot be successfully obtained via the conventional Shelby tube technique, piston sampling may be employed. In these instances either a floating or rod-mounted piston is located in such a manner that the piston rests on the top of the sample as it is pushed into the tube. The piston creates a vacuum which allows for retention of the sample within the tube. Core Orientation Such surveys are carried out where information is required regarding the spatial orientation of planar features, palaeontological studies, etc. The techniques employed include the following: . . Impression Core Orientation: This technique employs a hollow tube fixed to the base of the drill string filled with a suitable Plasticine material. The tube is lowered to the base of the pre-washed borehole and the orientator is pushed to seat onto the proud core break. The tube is withdrawn and the impression in the Plasticine matched with the bottom of the previous core run. Correct orientation is maintained during the raising and lowering of the drill string. Integral Core Orientation: This technique involves the drilling of a pilot hole (E size or similar) to 1.5 metres below the base of the main borehole using centralising bushes to centre the pilot hole in the main borehole. An orientated bar or pipe is placed in the pilot hole and cemented into position. The orientated bar is overdrilled once the cement has set. The technique can be employed in vertical or inclined holes, and is specifically used where highly fractured formations have been intersected, or the impression technique cannot be employed. 13 PLATE 2.2.3.1 Skid Mounted Rotary Core Drilling Rig PLATE 2.2.3.2 Trailer Mounted Rotary Core Drilling Rig 14 2.2.4 ROTARY PERCUSSION DRILLING There are two types of rotary percussion drilling equipment. The one is known as a top drive rig and consists of a drive-head which remains above the surface and is connected via drill rods to a drill bit. The drive-head rotates the drill string as well as imparts an impact force into the rods. The impact of the drill bit chips the rock and these chips are then air-flushed to the surface. The second is known as a ‘Down-the-Hole Hammer’ (DTH) and is very similar to the top drive rig above, only the impact force is generated by a down-the-hole hammer. This is a percussion hammer driven by air and imparts a rapid series of impacts to the drill bit which is part of the hammer. The rotation drive to the drill stem is provided by a top drive-head. The down-the-hole hammer is favoured for geotechnical investigation purposes because of greater versatility and sensitivity, particularly when recording penetration times. The standard procedure in terms of geotechnical investigation is for percussion chips to be collected at one metre intervals. During drilling operations the operator is required to keep a record of penetration time per metre, air loss, levels of water strikes, intersection of cavities and anything else that may be of specific interest to the logger of the borehole. A borehole log is compiled from the inspection of the chip samples, an evaluation of penetration times, and other relevant information supplied by the driller. The nature of the technique is such, that the compilation of the borehole log can be influenced by a variety of factors, which in turn can lead to inaccurate interpretation of the soil / rock conditions. Some of the more important of these factors are as follows: . .. The highly disturbed nature of the chip samples recovered and the possibility of contamination of these samples Total loss of sample in loose or soft layers Incorrect interpretation of the penetration rate in relation to the hardness of the material being penetrated Based on the above, in terms of geotechnical investigation, rotary percussion drilling should only be used to obtain a rough indication of the soil / rock profile, as it is subject to a large number of inaccuracies, which include to a large extent the experience of the driller and logger. However, the advantages of the rotary percussion technique are that it is relatively inexpensive when compared with rotary coring, being about one tenth of the cost of rotary coring. Drilling production is also fast when compared to rotary coring, with production rates of 80 to 100 metres per day possible. It is also one of the few techniques that can be used to economically penetrate boulder horizons, or layers of chert, which are often encountered in Dolomitic terrain. In southern Africa the technique has been successfully used as part of the overall geotechnical investigation procedures in Dolomitic areas, Wagener (1984). The technique is also used for the following applications: .. .. As probe holes to determine rock head depths As probe holes to determine the depth and extent of old mine workings To form boreholes for the conducting of in-situ tests (pressuremeter, lugeon tests) To form boreholes for the installation of geotechnical instrumentation (piezometers, extensometers, inclinometers, etc). 15 2.2.5 IN-SITU SHEAR STRENGTH TEST Vane Shear Test The vane shear test is routinely used to obtain undisturbed peak and remoulded undrained shear strength. The test consists of placing a four-bladed vane in the undisturbed soil and rotating it from the surface to determine the torsional force required to cause a cylindrical surface to be sheared by the vane. This force is then converted to a unit shearing resistance of the cylindrical surface as shown in FIGURE 2.2.5.1. A typical example of the equipment employed to apply torque to the steel rods from the surface is also shown in FIGURE 2.2.5.1. The steel rods are housed in a sleeve to prevent flexing and protect the rods. The four-bladed vane, which is connected to the base of the steel rods, is housed within a ‘torpedo’ attached to the base of the sleeve. For standard tests the height of the vane should be twice its diameter. The selection of the vane size is directly related to the consistency of the soil being tested, with larger vane sizes being used in softer soils. The test procedure is to advance the vane from the bottom of the torpedo in a single thrust to the depth at which the test is to be conducted. Once the vane is in position torque is applied at a slow rate, using the gear-driven surface equipment. The torsional force is measured and converted to unit shearing resistance in accordance with the following assumptions: . .. .. . . .. .. .. Penetration of the vane causes negligible disturbance, both in terms of changes in effective stress and shear distortion No drainage occurs before or during shear (undrained conditions prevail) The soil is isotropic and homogeneous The soil fails on a cylindrical shear surface The diameter of the shear surface is equal to the width of the blades At peak and remoulded strength, there is a uniform shear stress distribution across the shear surface There is no progressive failure, so that at maximum stress at all points, the shear surface is equal to the undrained shear strength The results of a vane shear test may be influenced by various factors eg: Type of soil and grain size, especially where a permeable fabric exists Strength and anisotropy Disturbance due to insertion of the vane Rate of rotation, or strain rate Time lapse between insertion of the vane and the beginning of the test Progressive / instantaneous failure of the soil around the vane It should be noted that the assumptions described above are not likely to all apply at the same time. The test is therefore best limited to cohesive, fine grained soil. Borehole Shear Testing In this test, a sleeve is placed against the side-walls of the drilled holes and the applied pressure measured while pulling from the surface. By repeating this at different pressures, either at constant strain, or at constant applied stress, the Mohr Coulomb parameters can be determined. 16 FIGURE 2.2.5.1 Vane Shear Apparatus 17 2.2.6 IN-SITU STIFFNESS TESTING Pressuremeter Test The pressuremeter test was originally developed by Menard (1956) and comprises a horizontal in-situ loading test carried out in a borehole by means of a cylindrical expandable probe. A major difference between categories of pressuremeter tests lies in the method of installation of the device in the ground. In accordance with Mair and Wood (1987), the following two broad categories of tests can be distinguished in terms of installation method: .. .. .. Menard type pressuremeter (MPM) test in which the device is installed in a borehole Self-boring pressuremeter (SBP) test in which the device bores its own way into the ground, usually from the bottom of a borehole The following parameters can be deduced from the results of the pressuremeter test: Deformation modulus (ie compressibility) Undrained shear strength for clays or weak rocks Effective angle of friction for sands In-situ total horizontal stress The degree of success in obtaining any of these parameters is dependent upon the type of test and the interpretation of the data. Consideration must also be given to possible anistropic soil behaviour, ie when the vertical and horizontal soil stiffness vary. For more details with regard to pressuremeter testing and its interpretation, reference should be made to Windle and Wroth (1977), Baquelin et al (1978), and Mair and Wood (1987). Dilatometer Test Various dilatometers are available, such as the Marchetti dilatometer, to measure lateral soil stiffness by inserting a flat vane into the soil to the required depth and applying load to a small circular plate. Monitoring the plate displacement allows the lateral soil stiffness to be measured. Plate Load Test A plate load test is usually carried out to determine the compressibility and occasionally the bearing capacity of soils and rocks. This test is convenient and provides a direct method of obtaining these parameters. It is often used in soils or rocks which cannot be sampled, or where the structure, joints etc, may control the engineering behaviour of the soil /rock mass. In its simplest form, the plate load test comprises a rigid plate placed on the surface of the soil to be tested. The load is provided by an hydraulic jack, using a Kentledge or an anchored beam as reaction. FIGURE 2.2.6.1 shows a typical test arrangement. The plates used must be rigid and generally vary in diameter from 200 mm to 1000 mm. 18 . . . The following procedures are adopted for the test: The test site is carefully levelled, and the plate bedded into the layer being tested using Plaster of Paris and /or bedding sand Load is applied to the plate using a hydraulic jack in a series of pre-determined steps. This application of load and the maximum load applied must be designed to conform with the anticipated structural loads Plate settlement is usually measured by means of dial gauges. In order to measure any tilt of the plate it is advisable to have four measuring points. The dial gauges are usually fixed to a beam supported by posts, bearing on the soil, some distance from the loaded area to avoid the readings being influenced by the settlement of the plate A variation to the standard test procedure can be implemented to allow the soil below the plate to be saturated at a specific load. The objective of this procedure is to allow the determination of any collapse properties associated with the material being tested. The widespread use of auger trial holes and test pits in southern Africa has led to the development of light and portable horizontal plate load equipment suitable for use in trial holes and test pits. By carrying out the tests in a horizontal direction, the necessary reaction is provided by the opposing faces of the trial hole or test pit. The bearing plates on either side are of equal size and the test procedure is essentially the same as that used for vertical plate load tests. The distance between the plates is measured and the movement of each plate is taken as half the total on the assumption that the two plates have moved equally. FIGURE 2.2.6.1 Example of Vertical Plate Load Test Arrangement 19 2.2.7 IN-SITU PENETROMETER TESTING Penetrometers are useful for determining the consistency of the soil below the ground. Over the years many and varied correlations have been developed for both strength and stiffness determination from penetration testing. These tests are generally more cost effective than many of the other mentioned tests and thus are very popular for routine data collection at a greater frequency than the more expensive tests. Standard Penetration Test (SPT) This process was standardised in the 1920's and 1930's into what is known as the Standard Penetration Test. In the execution of this test a standard 51 mm diameter split-spoon sampler, known as a Raymond Spoon, is driven into the soil at the bottom of a borehole. A free-fall hammer of 63.5 kg operating off a trip mechanism and falling through a height of 762 mm provides the driving force. The number of blows required to drive the sampler each 150 mm increment of a total of 450 mm penetration is recorded. The blow count for the first 150 mm increment is discarded and the sum of the blow counts for the second and third 150 mm increments is known as the SPT ‘N’ value. The standard penetration test has become accepted worldwide as a useful test in geotechnical investigation and foundation design. SPT results in boreholes give an empirical qualitative guide to the in-situ engineering properties of cohesive and cohesionless soils, and provide samples of the soil for classification purposes. The results of the SPT can be affected by incorrect drilling and sampling procedures some of which are given below (refer also to the Canadian Foundation Engineering Manual, 1985): .. .. .. . Inadequate cleaning of the bottom of the borehole Driving the spoon above the bottom of the casing Failure to maintain sufficient hydrostatic head in the borehole Not using the standard hammer drop or correct mass Freefall of the hammer is not obtained The tip of the spoon is damaged Not recording blow counts and penetration accurately It is thus extremely important that the drilling crew carrying out the tests is experienced in this type of work. Even then, it is advisable to carry out some CPT tests close to the borehole positions to check the correlation between the two. This will give an indication as to whether the SPT values are reliable. The relationship between the SPT ‘N’ value and engineering properties is empirical and some guidelines regarding the evaluation and interpretation of SPT ‘N’ values are given in SECTION 3.0: SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS. 20 Dynamic Probe Super Heavy (DPSH) Test In southern Africa considerable use is made of a local standard of the Dynamic Probe Super Heavy Test (ISSMFE Technical Committee on Penetration Testing, 1988), instead of the Standard Penetration Test (SPT). A 60° disposable cone, 50 mm in diameter, is fitted onto the bottom of an ‘E’ size rod, in place of the SPT sampler, and driven into the ground by a 63.5 kg hammer falling through 762 mm. The number of blows required to drive the cone through each successive 300 mm of penetration is continuously recorded. This provides an empirical indication of consistency. Once refusal depth is reached (more than 100 blows per 300 mm), the driving rods are withdrawn by 600 mm. The disposable cone remains at the base of the hole. The rods are then re-driven with the number of blows per 300 mm being recorded. These re-drive blow counts provide an indication of the skin friction, if any, acting on the drive rods, which can lead to falsely high penetration readings. Data collected from the DPSH test, including the re-drive figures, are presented on a report sheet in FIGURE 2.2.7.1. This test is very economical and can be rapidly performed. A major disadvantage of the test is that no soil sample is obtained. In many instances this disadvantage can be overcome by adopting a variation to the test procedure by fitting a Raymond splitspoon sampler to the ‘E’ rods, instead of the solid cone ie an SPT. This technique provides a continuous disturbed representative sample of the soil profile. Any blow counts recorded during this operation cannot however be correlated with those of the actual DPSH test. The DPSH rig is designed so that tests can be undertaken in areas that are not readily accessible, such as inside existing buildings, and in narrow passageways between buildings. PLATE 2.2.7.1 shows a typical DPSH test rig. PLATE 2.2.7.1 DPSH Test Rig 21 .. .. . The DPSH is used under the following conditions: . . . As economical supplementary data between boreholes on larger sites On sites with erratic profiles, alluvial, colluvial or lacustrine deposits; the test will locate softer areas Probing for rock or hard strata (although refusal is not necessarily rock level) In conjunction with a soil profile it will provide rough consistency readings which can be plotted graphically As the test closely approximates a driven pile, it is extensively employed for determining an estimate of skin friction and installation depths of driven cast-in-situ piles. In non-cohesive materials it is very reliable, but must be used with caution in cohesive soils due to rod adhesion. The test will also indicate pile driving conditions. Limitations of the test are: Driving refusal is frequently experienced on hard layers (such as very dense ferricretes or calcretes, boulder horizons) which may be underlain by soft soil horizons Differences in remoulding, caused by the small diameter cone on the one hand and the considerably larger piling tube on the other, can lead to erroneous prediction of pile installation depth Similar differences may occur when excessive pore water pressures are set up during the driving of a pile, whereas this does not occur with the DPSH test A graphic presentation of this data is given in FIGURE 2.2.7.1. The interpretation of the test results is generally associated with local experience. As a preliminary evaluation the blow counts can be taken as being roughly equivalent to the SPT ‘N’ value (see SECTION 3.3). In the interpretation however, it is essential to take into account the influence of the rod friction. 22 Depth (metres) DPSH blows / 300 mm FIGURE 2.2.7.1 Typical Results from a DPSH Test 23 Cone Penetration Test (CPT), (formerly the Dutch Cone Penetration Test) This method was initially developed in the Netherlands in the 1930's where it was first used as a means of determining the ultimate bearing capacity of driven piles founded in sand. Over the years the test has been called the Dutch sounding test, the Dutch probe, and the static cone penetration test. In terms of acceptable international standards (ISSMFE Technical Committee on Penetration Testing, 1988) it is now referred to as the Cone Penetration Test (CPT). In the CPT, a 60° cone with a cross-sectional area of 1000 mm2, usually equipped with a friction sleeve of the same diameter as the cone, and a surface area of 1.5 x 10 4 mm2 is pushed into the ground at a rate of 20 mm per second. Separate measurements of cone penetration resistance (point resistance), total penetration resistance, and the side friction resistance of the friction sleeve are made continuously throughout the test. The main advantages of the CPT are that the testing procedure is relatively simple and repeatable, and the test results can be used directly for design purposes. The CPT also gives a continuous record of soil resistance values throughout the depth of penetration. .. . .. . . The main limitations of the CPT are: Penetration depth limitations due to machine reaction capacity The technique is rarely effective in gravels and boulder horizons, and also not suited to weathered rock profiles No samples are recovered The data obtained from the cone penetration test may be employed to: Assist in evaluating the soil profile Interpolate ground conditions between control boreholes Evaluate engineering parameters of soils (relative density, shear strength, compressibility characteristics, liquefaction potential) Assess driveability, bearing capacity and settlement of piled foundations Mechanical cone penetrometers, Begeman (1965), have a telescopic action which requires an outer probe sleeve and inner rod. These mechanical cone penetrometers offer the advantage of low equipment cost and simplicity of operation. They do however have the disadvantages of a slow incremental procedure, limited accuracy in very soft soils, and labour intensive data handling and presentation. With the electrical cone penetrometer, the friction sleeve and cone point advance together as a single system. The point resistance and local side shear are recorded continuously with the use of built-in load-cells. An electrical cable located inside the rods connects the load-cells to recording equipment at ground surface. Electrical cones carry a high initial equipment cost and require skilled operators as well as adequate back-up for calibration and maintenance. They do however offer advantages over the mechanical penetrometer, such as, a more rapid procedure, higher accuracy and repeatability, automatic data logging, reduction and plotting. One of the important applications of the CPT is to evaluate variations of soil type within the soil profile. With mechanical and electrical cones, extensive use is made of what is known as the friction ratio as a means of soil classification, Jones (1974), Schmertmann (1975). This friction ratio is the ratio between sleeve friction and the point resistance, and is expressed as a percentage. 24 The most significant recent development in electric cone penetration testing is the development of the Piezocone penetrometer test (CPTU) which incorporates a pore water pressure sensor in the cone. This allows for the measurement of pore water pressure present in the soil during penetration. Pore water pressure measurements during cone penetration testing provide more details on the stratification, and has added a new dimension to the interpretation of certain geotechnical parameters, especially in loose or soft, fine grained soil deposits. This has resulted in CPTU testing becoming a prime tool for stratification logging of soil deposits, Jones and Rust (1982), Campanella and Robertson (1988). In addition, pore water pressure dissipation tests readily allow the determination of the coefficient of consolidation, c v. . .. Further advantages of the CPTU test over the conventional CPT are given by Campanella and Robertson (1988): The ability to distinguish between drained, partially drained, and undrained penetration conditions The ability to evaluate flow and consolidation characteristics The ability to assess equilibrium groundwater conditions Recent cone developments include the use of a seismic cone to measure small strain stiffness in the soil. A guide to the interpretation of the results of CPT and CPTU can be found in SECTION 3.0: SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS. Dynamic Probe Light (DPL) Test, commonly called DCP in southern Africa A local standard of the Dynamic Probe Light Test (ISSMFE Technical Committee on Penetration Testing, 1988) is used in many applications in southern Africa. A 20 mm diameter 60° cone is driven into the soil by an 8 kg weight dropped through 575 mm. The penetration resistance is expressed as millimetres per blow. The original test, van Vuuren (1969), was designed for the rapid determination of the California Bearing Ratio (CBR) to depths of about one metre for investigation into road pavement performance and design. Besides the original application in the field of pavement evaluation and design, the test has also been used as a rough guide in compaction control and for estimating soil conditions for the design of shallow footings. The main advantage of this type of equipment is that it is light, portable, inexpensive to operate and provides a continuous rough record of soil consistency over the depth tested. The disadvantages are that no sample is recovered, the nature of the equipment limits its depth capability to two metres below surface, and the equipment is unable to penetrate hard lenses or other obstructions, large gravel, boulders etc. The ease and low cost with which results can be obtained is therefore somewhat offset by the limitations of the test and the indirect approximation to soil conditions that it provides. A guide to the interpretation of the results of this test can be found in SECTION 3.0: SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS. 25 2.2.8 IN-SITU PERMEABILITY TESTS Lugeon Testing Lugeon testing (also known as water pressure or packer testing) is carried out to measure the permeability of the soil or rock at specific depths in a borehole. The equipment consists of two packers comprising steel tubes surrounded by inflatable rubber sleeves separated by a perforated length of steel tube. The spacing of the packers can be adjusted to the specific depth of soil or rock to be tested. The minimum length of a packer sleeve is 700 mm to ensure a watertight seal. The packer arrangement is connected via high pressure tubing to a suitable pump on the surface. Data collected from the system is obtained by flow meters and pressure gauges. The above arrangement is known as a double packer system. The system can also be adapted for so-called single packer tests, where testing is carried out between the packer and the bottom of the hole. The test consists of pumping water into the isolated zone of the borehole at three different pressures, in the following sequence: 1st 2nd 3rd 4th 5th 10 minutes at low pressure 10 minutes at medium pressure 10 minutes at high pressure 10 minutes at medium pressure 10 minutes at low pressure a b c b - repeated a - repeated The actual duration of each pressure stage is accurately timed. The pressures selected are dependent on the depth at which each test is carried out. The required pressures are maintained to an accuracy of 5% during each pressure stage. Piezometer Installation Piezometers are installed in boreholes in order to provide information regarding the ‘at rest’ levels of the groundwater table. In addition, groundwater pressure can be measured via more specialised piezometers ie hydraulic, electrical or pneumatic. In general, piezometers are installed into pre-cleaned holes by lowering a selected porous tip to approximately 500 mm above the bottom. The tip is surrounded by a filter of graded, washed silica sand and sealed off with a bentonite plug. The remainder of the borehole is sealed by introducing cement / bentonite grout. 2.2.9 IN-SITU DENSITY TESTS In-situ density tests are mainly used for compaction control in roads and earthworks construction. In certain instances the determination of in-situ density may form part of an overall geotechnical investigation field-work programme. Both the sand replacement method and nuclear methods are used for the determination of in-situ density. In the sand replacement method, the in-place dry density is determined by forming a hole in a layer and dividing the mass of the material removed from the hole by the volume of the hole, the latter being determined by filling the hole with a fine sand of known density. The disadvantage of this test is that the material removed from the hole needs to be dried to a constant mass, usually overnight in a suitable oven. This means that a period of at least 12 to 18 hours is required before results become available. The advantage of the test is that it gives an accurate value of in-situ dry density and in-situ moisture content. 26 Nuclear systems for the determination of wet density and moisture content have become popular in recent years. One of the main advantages of this test procedure is that results are immediately available. The disadvantage is that there are some potential inaccuracies associated with the results produced from this test. The inaccuracies are generally associated with the measurement of moisture content and can easily be overcome by taking a sample at each test position for the laboratory determination of moisture content. To a large extent this negates the advantages of having results available immediately. On most roads and earthworks contracts the results of nuclear gauge tests are generally only accepted as a control procedure after suitable calibration, with sand replacement tests, has been carried out. Soiltech is able to offer both sand replacement and nuclear gauge density tests. These tests are carried out in accordance with the procedures recommended in TMH 1 (1986). 2.2.10 GEOPHYSICAL TECHNIQUES Geophysical exploration is a form of field investigation in which a set of physical measurements relating to the underlying soil or rock strata is made at ground surface or in boreholes. The measurements indicate variations in space or time of certain physical properties of the soil / rock materials. Geophysics is therefore a blend of physics and geology, since the physical measurements are interpreted in terms of sub-surface geological conditions. The properties of soils / rock which are of significance in geophysical exploration are density, magnetic susceptibility, electrical conductivity, elasticity, and thermal conductivity. Changes in one or more of these properties can be measured by sufficiently sensitive instrumentation. .. The main advantages of geophysical techniques are: It is possible to carry out investigations of large areas rapidly and economically The techniques can be used to locate critical areas for further field investigation The disadvantage of the technique is that the results are dependent on the interpretation of physical measurements. These measurements are not in themselves geological or geotechnical parameters relative to the site sub-surface conditions. It is therefore essential that geophysics is carried out and interpreted in conjunction with a carefully planned drilling programme. The main application of geophysics in geotechnical investigations is the interpolation of sub-surface geological strata between carefully controlled drilling positions. The more common geophysical techniques used in geotechnical investigations are magnetics, gravity and resistivity. For more detailed information reference should be made to Griffiths and King (1965), Kleywegt and Enslin (1973), West and Dumbleton (1975), Darracott (1976), and Bullock (1978). 27 2.3 GEOTECHNICAL ENGINEERING LABORATORY SERVICES Standardised and consistent soil mechanics and materials testing often forms the basis for design and site quality control in geotechnical and materials engineering. A guide to testing procedures and requirements for the commonly specified soil mechanics and materials tests is presented in TABLE 2.3.1. All relevant road type materials testing is carried out in accordance with TMH 1 (1986). Soil mechanics testing is carried out in accordance with accepted published or international standards. Soiltech no longer provides geotechnical laboratory testing services, but will assist and/or arrange a laboratory testing programme with a suitable accredited laboratory on projects where Soiltech has undertaken the field-work portion of the geotechnical investigation. TABLE 2.3.1 Guide to Laboratory Procedures and Requirements Laboratory Test Parameter Determined Duration (days) Sample Requirements Triaxial Compression Test Unconsolidated Undrained Undrained shear strength (UU) of cohesive soils (cu) 3 Undisturbed: Good quality sealed block sample 300 x 200 x 150 mm thick. Shelby tube or piston sample Consolidated Undrained with pore water pressure measurements (CU) Effective shear strength parameters c’ or φ’ 5 to 7 Consolidated Drained Test (CD) Effective shear strength parameters c’ or φ’ 7 to 10 Disturbed or remoulded: 2 kg of representative sample Effective shear strength parameters c’ or φ‘ 4 Residual shear strength parameter φ‘ 5 to 7 Undisturbed: Good quality sealed block sample 300 x 200 x 150 mm thick. Shelby tube or piston sample Shear Box Test Drained Shear Box Disturbed or remoulded: 2 kg of representative sample Consolidation Tests Consolidation test soaked at 11 kPa, loaded to 1600 kPa and then rebounded/ unloaded Compressibility characteristics Double oedometer test for collapse Compressibility and collapse characteristics over full loading spectrum 7 Collapse potential test. Sample loaded to 200 kPa, saturated and rebounded Compressibility and collapse characteristics Collapse potential index 3 Double oedometer test for heave Swell characteristics over full loading spectrum 7 Swell under load test Swell characteristics at specified load 3 28 7 Undisturbed: Good quality sealed block sample 300 x 200 x 150 mm thick. Shelby tube or piston sample Disturbed or remoulded: 2 kg of representative sample TABLE 2.3.1 (continued) Guide to Laboratory Procedures and Requirements Laboratory Test Parameter Determined Duration (days) Sample Requirements Permeability Tests Falling Head or Constant Head Coefficient of permeability 3 for Undisturbed: Good quality sandy soils sealed block sample 7 to 10 for 300 x 200 x 150 mm thick. clayey soils Shelby tube or piston sample Disturbed or remoulded: 2 kg of representative sample Bulk Density Bulk density Dry density Moisture content 3 Good quality sealed block sample 300 x 200 x 150 mm thick Grading / Sieve Analysis Particle size distribution to 0.075 mm 3 2 kg sample of undisturbed or disturbed soil Hydrometer Particle size distribution from 0.075 to 0.002 mm Atterberg Limits Liquid limit, plastic limit, plasticity index Moisture Content Moisture content Maximum dry density and optimum moisture content under specified compactive effort 2 40 kg of representative sample Mod AASHTO moisture density curve. Plot of CBR versus dry density based on CBR at 3 compactive efforts (Mod AASHTO, Procter, NRB) 6 70 kg of representative sample Index Properties Moisture Density Relationship Mod AASHTO Procter California Bearing Ratio (CBR) 29 2.4 .. . .. . SITE INVESTIGATION FOR PILING Minimum requirements for a site investigation for piling are: Establishment of competent founding material and engineering properties thereof Presence or absence of obstructions, including depth of fill/ builders’ rubble Presence of water/ seepage and the risk of hole collapse / necessity of casing over all, or part of the hole depth Presence of cavities Presence of aggressive soil/ water Consistency of the soil profile, including penetrometer data 2.5 . . .. . SITE INVESTIGATION FOR LATERAL SUPPORT Minimum requirements for a site investigation for lateral support are: Establishment of the shear strength parameters of the material in front of, and behind the wall Establishment of the soil stiffness in front of, and behind the wall (penetrometer data is a minimum requirement) Presence of water-table / seepage Likelihood of anchor / nail hole collapse (casing requirement) Presence of obstructions to piles / anchors, including services and adjacent structures / basements etc 2.6 .. .. SITE INVESTIGATION FOR SOIL IMPROVEMENT Minimum requirements for a site investigation for soil improvement are: Establishment of the soil consistency variation in plan, and with depth Water-table depth Soil grading / compactibility (are clays and silts present ?) Proximity of surrounding buildings /crack survey beforehand Applicable Norms As the topic of Site Investigations is broad, not all aspects can be adequately covered in one book. The reader can get further details on this important topic at the following links / references: BS 5930: Code of Practice for Site Investigations (1999) SAICE: Code of Practice for the Safety of Persons Working in Small Diameter Shafts and Test Pits for Geotechnical Engineering Purposes (2007) SAICE: Draft Code of Practice for Site Investigations (2007) 30 3.0 SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS 3.1 NOTES ON SOIL PROFILING As indicated in SECTION 2.0: GEOTECHNICAL INVESTIGATION, an accurate description of the soil profile forms the basis of the geotechnical investigation for any engineering development. It is important that each layer is described in a consistent way to ensure accurate interpretation of the soil profile by those involved in the geotechnical design and construction process. The description of the soil in profile, based on the work of Jennings, Brink and Williams (1973), is related to the following: Designation Heading Example M Moisture Moist C Colour Reddish Brown C Consistency Stiff S Structure Intact S Soil Type Clay O Origin Residual Shale Moisture The moisture content is assessed as: DRY, SLIGHTLY MOIST, MOIST, VERY MOIST and WET. The assessment of the moisture content is dependent on the soil type. With a moisture content of approximately 20%, sand will probably be described as wet, whereas clay will probably be described as slightly moist. Colour Colour is important for description and correlation. Colour is described from the soil in profile as well as from a small sample of soil made into a creamy paste with water. A profile is MOTTLED when small exposures of different colours occur. A profile is BLOTCHED when larger exposures (75 mm and larger) of different colour occur. Colour charts obtainable from the South African Institution of Civil Engineers illustrate the main colours as well as variations in hue and lightness of each colour. These charts illustrate the following colours. 31 Blue: Dusky Pale Red: Dusky, Dark Pale, Light Green: Dusky Pale Grey: Dark Light Olive: Dark Light Orange: Brown: Dark Light Dark Reddish Light Reddish Dark Reddish Light Reddish Dark Yellowish Light Yellowish Yellow: Dark Light Consistency Consistency is a measure of the strength or density of the soil. Observations are based on the effort required to dig into the soil or to mould it with the fingers. The consistency of cohesive soils is based on the undrained shear strength and described as VERY SOFT, SOFT, FIRM, STIFF AND VERY STIFF. Consistency versus undrained shear strength guidelines are set out in SECTION 3.3. Non-cohesive soil consistency is based on the angle of shearing resistance of the soil and described as VERY LOOSE, LOOSE, MEDIUM DENSE, DENSE and VERY DENSE. Consistency versus angle of shearing resistance guidelines are given in SECTION 3.3. Structure The presence and type of discontinuities in the soil mass define the structure. Structural characteristics are generally related to cohesive soils in the following terms: INTACT Absence of fissures and joints, though tension cracks may occur in firm samples when broken with a pick FISSURED Presence of closed joints SLICKEN-SIDED Highly polished fissures, usually indicative of expansive soils SHATTERED Indicates fissures which have opened up and allowed entry of air, often associated with expansive soils MICROSHATTERED Shattering on a small scale with shattered fragments the size of sand grains. If well developed, the soil appears granular when cut, but the grains break down into clay and/or silt when wetted and rubbed. Indicates the presence of a highly expansive soil. LAMINATED, FOLIATED or STRATIFIED Indicates that the soils show the laminated, foliated or stratified structure of the parent rock or geological process from which they were derived. 32 Soil Type The soil type is described on the basis of the grain size of the individual particles. The basic grain size classes are given below. Most natural soils occur as a combination of these classes eg silty clay or gravelly sand. BOULDERS Fragments of rock > 200 mm GRAVEL COBBLES COARSE MEDIUM FINE 60mm - 200mm 20mm - 60mm 6.0mm - 20mm 2.0mm - 6.0mm The range and size of boulders and gravel, the shape, the proportion by volume of the matrix and the description of the matrix are important. SAND COARSE MEDIUM FINE 0.6mm - 2.0mm 0.2mm - 0.6mm 0.06mm - 0.2mm Sand particles are visible to the naked eye. SILTS 0.002mm - 0.06mm Silts are barely gritty between fingers and thumb when wet, but are gritty on tongue against teeth. Silts are not easily rolled into threads when moist. Silts exhibit dilatancy when moulded with water into a pat, (ie it increases its volume when shearing occurs, which is illustrated by the film of water on the surface being absorbed, if the pat is distorted). Silts dry moderately quickly and can be dusted off the fingers. Dry lumps possess cohesion but powder easily in the fingers. CLAY Particles less than 0.002mm Clay particles are flaky (not powdery) when broken and will soften with the addition of water. They have a soapy or greasy feel when wetted and rubbed on the palm of the hand. Clay sticks to fingers and dries slowly. There is no dilatancy or grittiness on tongue against teeth. 33 .. .. Origin In any soil profile there are four basic categories of origin: Rock Residual Soil developed from Parent Bedrock Pedogenic Material Transported Soil In the southern African context, the demarcation between residual soils and overlying transported soils is often defined by the ‘pebble marker’. This horizon is generally characterised by a gravel layer overlying the residual soil. Rock Materials described as rock comprise igneous, metamorphic or sedimentary (not pedogenic) horizons with unconfined compressive strengths of the intact or unjointed material in excess of 1000 kPa. Residual Soil A residual soil is formed from in-situ decomposition of rock. Decomposition can be caused by chemical weathering or mechanical disintegration which is a function of potential evaporation (temperature, humidity, wind) and average annual precipitation. Pedogenic Material Pedogenic material is residual or transported soil that has become strongly cemented or partially replaced by one of the cementing agencies. Description Cementing Agency Ferricrete Iron Oxide Calcrete Calcium Carbonate Silcrete Silica Transported Soil This is soil which has been transported by a natural agency (water, wind, gravity) during relatively recent geological times (Pleistocene or Tertiary), and which has not undergone lithification into a sedimentary rock, or cementation into a pedogenic material. 34 Type Agency Source Resulting Soil Talus (scree and coarse colluvium) Gravity Rock Outcrops Unsorted Angular Gravel and Boulders Hillwash (fine colluvium) Run-off Acid Crystalline Basic Crystalline Arinaceous Sediment Argillaceous Sediment Clayey Sand Clay Sand Clay or Silt Alluvium (gully wash) Rivers, Streams and Gullies Various Rocks and Soils Boulders Gravels Sands Silts Clays Lacustrine Deposits Streams Terminating in Lake, Pan or Pool Various Rocks and soils Sand Silt Clay Estuarine Deposits Tidal Rivers and Waters Mixed Sand Silt Clay Littoral Deposits Waves Mixed Beach Sand Aeolian Deposits Wind Mixed Sand and Clayey Sand Sub-surface Water Condition The water-table is that level or those levels in the soil where the water in the pores of the soil occurs at atmospheric pressure, ie the level to which the water finds its own way in a borehole. The perched water-table is a table which is only present in the soil temporarily, often caused by heavy rains entering the soil and coming into contact with an impermiable soil layer. It will disappear and sometimes re-appear depending upon the seasons, or drainage conditions of the site. The permanent water-table is the water-table which persists throughout the seasons of the year, with only minor seasonal fluctuations of level. A typical soil profile and tabulation of the various soil symbols are given in FIGURES 3.1.1 and 3.1.2 respectively. 35 - FIGURE 3.1.1 Example of Typical Soil Profile 36 FIGURE 3.1.2 Typical Soil Symbols 37 3.2 NOTES ON ROCK MASS DESCRIPTION The accurate description of rock engineering conditions, like the requirements outlined for the soil profile, necessitates a detailed and practical method of describing samples of rock core retrieved from a rotary cored borehole. With these requirements in mind, the publication ‘Guidelines for Soil and Rock Logging in South Africa’, by the Core Logging Committee of the South African Section of The Association of Engineering Geologists (1994), has become the accepted norm for the description and interpretation of geological and rock engineering conditions in southern Africa. This method of logging rock cores was based on similar principles to those for soil profiling outlined by Jennings et al (1973), but due to the complexity of rock mass behaviour subject as it is to weathering and discontinuities, the soil profiling system was modified and adapted to provide an adequate rock mass description. The core log contains descriptions of the rock mass parameters, discontinuity surfaces, and the materials infilling these surfaces. The joint infill material has a major influence on the engineering properties of the rock mass and have to be taken into account in the design process. 3.2.1 DESCRIPTION OF PRIMARY ROCK MASS PARAMETERS Six basic rock mass parameters are used in the same way as soil descriptive parameters. These are tabulated and compared to the soil description below. Soil Description Rock Mass Description Moisture N /A Colour Colour Consistency Weathering Structure Fabric Description and Spacing of Discontinuities N /A Hardness Soil Type Rock Type Origin Stratigraphical Horizon Colour Colour is the basic and most easily identifiable characteristic, and colour variation is a primary indication of weathering. The colour of a rock mass is generally related to its mineralogy. Quartz and feldspar will give rise to light coloured rock while pyroxine and olivine give rise to dark coloured rock. Cores should be washed before logging and the colour recorded on wet, recently broken surfaces. The standard Munsell colour chart obtainable from SAICE should be used to describe the hue and the lightness of the colour. Where variable colour exists, the dominant colour of the rock mass should be given, followed by the secondary colour which usually exhibits a pattern such as: BANDED, STREAKED, BLOTCHED, MOTTLED, SPECKLED AND STAINED. Where inclusions, such as amygdales occur, they should be described together with their colour. 38 Weathering Weathering of a rock mass is a process of alteration by mechanical, chemical or biological action which significantly affects the behaviour of the rock material and the rock mass as a whole. The decomposition and the disintegration of a rock mass is given the term ‘weathering’ and the degree of weathering is given in TABLE 3.2.1. TABLE 3.2.1 Rock Mass Weathering Diagnostic Feature Grain Original Boundary Texture Condition Discoloration Extent Fracture Condition Surface Characteristics None Closed or Discoloured Unchanged Preserved Tight Slightly Weathered < 20% of Discoloured, fracture spacing may contain on both sides thin filling of fracture Partial Discolouration Preserved Tight Medium Weathered > 20% of Discoloured, Partial to complete Preserved fracture spacing may contain discolouration, on both sides thin filling not friable, except of fracture poorly cemented rocks Descriptive Term Unweathered Partial Opening Highly Weathered Throughout — Friable and Possibly Pitted Mainly Partial Preserved Separation Completely Weathered Throughout — Resembles Soil Partly Complete Preserved Separation For detailed definition of the five degrees of weathering, reference should be made to the original publication. Fabric Fabric describes the structural and textural features of the intact rock material. Texture of the rock mass is governed by the size and arrangement of the individual grains. TABLE 3.2.2 gives recommendations on grain size terminology. TABLE 3.2.2 Rock Mass Fabric Description Size in mm Recognition Equivalent Soil Type Very fine grained < 0.06 Individual grains cannot be seen with a hand lens Clays and Silts Fine Grained 0.06 - 0.2 Just visible as individual grains under hand lens Fine Sand Medium Grained 0.2 - 0.6 Grains clearly visible under hand lens, just visible to the naked eye Medium Sand Coarse Grained 0.6 - 2.0 Grains clearly visible to naked eye Coarse Sand Very Coarse Grained > 2.0 Grains measurable Gravel 39 Micro-structural features such as foliation and banding give rise to anisotropic behaviour of the rock material for small scale features, and the rock mass gives rise to this behaviour for larger scale features. A spacing of 10 mm is given as the boundary between micro-structure and a discontinuity. TABLE 3.2.3 outlines the terms given to both micro-structure and a discontinuity surface. Discontinuity Surface Spacing This describes mechanical discontinuity, weakness planes or bedding planes. Two major categories of discontinuity are given by features characteristic of the origin, such as bedding, and features as a result of movement within the rock mass, such as joints. In the core log only the discontinuity spacing is given in the primary rock mass description. Any additional features are outlined separately. It is important to note that discontinuities caused by the drilling operation and the handling of the cores are not included in the description of the discontinuities. TABLES 3.2.3 (a) and (b) show how discontinuity surfaces are described. TABLE 3.2.3(a) Macro Features Descriptions for Structural Features: Bedding, Foliation or Flow Banding Spacing in mm Description for Joints, Faults or other Fractures Very Thickly (bedded, foliated or banded) < 1000 Very Widely (Fractured or Jointed) Thickly 300 - 1000 Widely Medium 100 - 300 Medium Thinly 30 - 100 Closely Very Thinly 10 - 30 Very Closely TABLE 3.2.3(b) Micro Features Description for Micro-structural Features: Lamination, Foliations or Cleavage Spacing in mm Intensely Laminated (foliated or cleaved) 3 - 10 Very Intensely <3 Rock Hardness Rock hardness is a measure of the strength of the rock material and plays a dominant role in the behaviour of structures in rock engineering, and in particular, structural foundations such as piles. The unconfined compressive strength is directly related to the hardness, and is graphically illustrated in FIGURE 3.3.10, and tabulated in TABLE 3.3.8. Rock Type and Stratigraphic Horizon Three basic rock types define the origin of a rock mass: Igneous, Metamorphic and Sedimentary. The basic mineralogy and texture observed in the core together with knowledge of the regional geology will enable the logger to name the rock type. The stratigraphic horizon which often identifies the behaviour and characteristics of the rock should precede the rock type. 40 3.2.2 DESCRIPTION OF DISCONTINUITY SURFACES The behaviour of rock masses is often governed by the nature and spacing of the discontinuity surfaces rather than the intact rock material properties. Recommendations with regard to the type and spacing of the discontinuity surface are given with the primary rock mass parameters. Descriptions of the nature of the discontinuity should incorporate the following: .. .. Separation of Fracture Walls Filling (the presence or absence of fill material within the discontinuity) Roughness or Nature of the Asperities on the Fractures Orientation of the Discontinuity TABLE 3.2.4 gives values for separation, filling and roughness of the discontinuity surface. TABLE 3.2.4 Nature of Discontinuity Surfaces Description of Separation of Fracture Walls Description Separation of Walls in mm Closed 0 Very Narrow 0 - 0.1 Narrow 0.1 - 1.0 Wide 1.0 - 5.0 Very Wide 5.0 - 25.0+ Terminology for Presence or Absence of Fracture Filling Materials Description Definition Clean No fracture filling material Stained Colouration of rock only; no recognisable filling material Filled Fracture filled with recognisable filling material Roughness Classification Classification Description Smooth Appears smooth and is essentially smooth to the touch; may be slicken-sided Slightly Rough Asperities on the fracture; surfaces are visible and can be distinctly felt Medium Rough Asperities are clearly visible and fracture surface feels abrasive Rough Large angular asperities can be seen; some ridge and high side angle steps evident Very Rough Near vertical steps and ridges occur on the fracture surface 41 The rock core descriptions together with drilling method, percentage core recovery, RQD (Rock Quality Designation) and fracture frequency, as well as, type of test and test result should be indicated on the borehole log. For the symbolic representation of various rock types reference should be made to FIGURE 3.2.1. The core log together with the drilling record are combined for the compilation of a borehole log. A typical log is given in FIGURE 3.2.2. FIGURE 3.2.1 Typical Rock Symbols 42 FIGURE 3.2.2 Typical Borehole Log 43 3.3 INTERPRETATION OF GEOTECHNICAL INVESTIGATION AND LABORATORY / IN-SITU TESTING DATA Of key interest to the engineer interpreting the information contained in a geotechnical investigation report, soil profile, or laboratory/in-situ test results, is the allocation of representative geotechnical design parameters to the soil or rock profile. It is useful to classify soils into various categories, as these sub-divisions are indicators of what problems, or patterns of behaviour, may be expected. Once the soil or rock types have been classified, the next most important design parameters required, are the soil /rock shear strength and stiffness values. 3.3.1 SOIL CLASSIFICATION The correct classification of the horizons encountered is essential to accurately evaluate and predict the engineering properties and behaviour of this horizon. The classification of the material as sand, clay or rock must first be carried out before strength or compressibility characteristics are assigned to it. There are several methods of classifying a soil. All of these methods have a broad classification based on grain size. The Unified or MIT classification systems are most commonly used, with the MIT classification given in TABLE 3.3.1, and the Unified system given in TABLE 3.3.2. The behaviour of the soil mass, and the properties and parameters assigned to it, will depend largely on whether it is classified as sand, clay or rock. As a rule of thumb, the finer the grain size, the poorer the engineering behaviour. With penetration testing such as the CPT test (outlined in SECTION 2.2.7) where no samples are recovered for grading and laboratory testing, methods of classifying soils based on the test results have been developed. The method proposed by Schmertmann (1967) outlined in FIGURE 3.3.1 and based on the CPT test results is commonly used. The friction ratio forms the basic guide as to whether the soil is cohesive or noncohesive, with a friction ratio typically higher than about 4% indicating a clay. Another method of classifying soils based on the results of the CPTU test, is given by Jones and Rust (1982), and is detailed in FIGURE 3.3.2. When engineering works are constructed using soil, the response of the different soil types to compaction and stabilisation is of great importance, and compaction characteristics can be predicted using the classification systems outlined above. For roads and earthworks the PRA system (Public Roads Administration after Allen, 1945) of classification is often used and reference should be made to it for these applications. This system classifies soils in terms of grain size, liquid limit and plasticity index and assigns a Group Index number to the soil which varies between 1 and 20. Soils with a Group Index less than 10 are predominantly coarse grained and have good sub-grade properties. Soils with a Group Index greater than 10 have poor sub-grade properties. Peats and highly organic soils are unsatisfactory as sub-grade material. 44 TABLE 3.3.1 Particle Size Classes Commonly Used in Engineering — Massachusetts Institute of Technology (MIT) Classification Grain Size (mm) Classification Individual Particles Visible Mineralogical Composition Identification Test Less than 0.002 Clay Electron Microscope Secondary Minerals (Clay Minerals and Fe-oxides) Feels Sticky; Soils Hands; Shiny When Wet 0.002 - 0.06 Silt Microscope Primary and Secondary Minerals Chalky feel on Teeth; When Dry Rubs Off Hands; Dilatant 0.06 - 2.0 Fine Sand Hand Lens Primary Minerals (mainly quartz) Gritty Feel on Teeth 2.0 - 6.0 Fine Gravel Naked Eye Rocks (sometimes vein quartz) Observed with Naked Eye 6.0 - 20 Medium Gravel Naked Eye Rocks Observed with Naked Eye 20 - 60 Coarse Gravel Naked Eye Rocks Observed with Naked Eye 60 - 200 Cobbles Naked Eye Rocks Observed with Naked Eye More than 200 Boulders Naked Eye Rocks Observed with Naked Eye 45 TABLE 3.3.2 Unified Soil Classification System (ASTM D-2487) after USAWES (1967) Highly Organic Soils Group Symbols Typical Names GW Well-graded gravels, gravel-sand mixtures, little or no fines GP Poorly graded gravels, gravel-sand mixtures little or no fines GM d u Silty gravels, gravel-sand-silt mixtures GC Clayey gravels, gravel-sand-clay mixtures SW Well-graded sands, gravelly sands, little or no fines SP Poorly graded sands, gravelly sands, little or no fines SM d u SC Clayey sands, sand-clay mixtures ML Inorganic silts and very fine sands, rock flour, silty or clayey fine sands or clayey sands with slight plasticity CL Inorganic clays of low medium plasticity, gravelly clays, sandy clays, silty clays, lean clays OL Organic silts and organic silty clays of low plasticity MH Inorganic silts, micaceous or diatomaceous line sandy or silty soils, elastic silts CH Inorganic clays of high plasticity, fat clays OH Organic clays of medium to high plasticity, organic silts Pt Peat and other highly organic soils Silty sands, sand-silt mixtures 46 Laboratory Classification Criteria D30 greater than 4: D10 (D30)2 between 1 & 3 cc = D10 x D30 cc = Determine percentage of sand and gravel from grain size curve. Depending on percentage of fines (fraction smaller than 0.075 mm sieve size), coarse grained soils are classified as follows: Less than 5 percent: GW, GP, SW, SP More than 12 percent: GM, GC, SM, SC 5 to 12 percent: Borderline cases requiring dual symbol CLEAN GRAVELS (little or no fines) GRAVELS with FINES (appreciable amount of fines) CLEAN SANDS (little or no fines) SILTS and CLAYS (liquid limit is greater than 50) SILTS and CLAYS (liquid limit is less than 50) SANDS with FINES (appreciable amount of fines) GRAVELS (more than half of coarse fraction is larger than 4.5 mm sieve size) SANDS (more than half of coarse fraction is smaller than 2.0 mm sieve size) Coarse Grained Soils Fine Grained Soils (more than half of material is smaller than 0.075 mm sieve size) (more than half of material is larger than 0.075 mm sieve size) Major Divisions Not meeting all gradation requirements for GW Atterberg Above ‘A’ line limits below with P.1 ‘A’ line or P.1 between less than 4 4 and 7 Atterberg are borderline limits below cases requiring use of ‘A’ line with P.1 greater than 7 dual symbols D60 greater than 4: D10 (D30)2 between 1 & 3 cc = D10 x D30 cc = Not meeting all gradation requirements for SW Atterberg Limits plotting limits above in hatched zone ‘A’ line or P.1 with P.1 beless than 4 tween 4 and 7 are borderline Atterberg limits above cases requiring use of ‘A’ line with P.1 greater than 7 dual symbols FIGURE 3.3.1 Soil Classification Based on CPT Test after Schmertmann (1967) FIGURE 3.3.2 Soil Classification Based on CPTU Test after Jones and Rust (1982) 47 The classification of the swelling potential for expansive soils based on clay content and plasticity index has been given by Williams and Donaldson (1973), and Seed (1978), in FIGURE 3.3.3. TABLE 3.3.3 gives approximate swell values of clays for the range of potential expansiveness after van der Merwe (1975). (a) (b) FIGURE 3.3.3 Activity of Expansive Soils given by (a) Williams et al (1973) and (b) Seed (1978) TABLE 3.3.3 Potential Expansiveness of Clays after van der Merwe (1975) Potential Expansiveness Heave: mm per m Very High > 80 High 40 Medium 20 Low 0 48 3.3.2 SOIL CONSISTENCY EVALUATION As the soil shear strength parameters vary with the soil density, it is important to be able to estimate the variation of the soil density with depth. This is sometimes done by weighing block samples, but most often the soil density is measured by correlation with penetrometer data, either using SPT or CPT rigs. Commonly used relationships are given below. The relationship between SPT and Relative Density of sands and gravels (DR) is: emax -- eo DR = e = 100 -- e max Where: (N1)60 (3.3a) 60 min (N1)60 is the SPT ‘N’ value corrected to 60% energy efficiency and overburden pressure The relationship between SPT and Relative Density of sands and gravels (DR) is: DR = 100 DR = 100 qtl for Normally Consolidated sands 300 qtl 300.OCR0.2 for Over-consolidated sands (3.3b) (3.3c) qc / pa 0.5 and pa = 100 kPa, and σ ’νo = overburden pressure q t / σ ’νo [ ] Where: qtl = 3.3.3 SOIL STRENGTH EVALUATION Both compressive and shear strength parameters can be assigned to soils and rocks. As the soil’s shear stiffness is about one third of its axial stiffness, it tends to fail in shear. Hence a soil’s compressive strength tends to be ignored and its shear strength takes primary focus. .. .. Strength parameters can be assigned to soils or rocks based on: Descriptions of the strength in the soil profile Index property tests Empirical relationships with penetration test results Direct measurement in-situ, or in the laboratory It should be noted, that the plane strain friction angle of soils is higher than the tri-axial friction angle, due to the confinement provided by the intermediate principal stress. CIRIA C580 Economic Design for Embedded Walls, suggests using a value 10% higher. Wroth (1984) derived an equation showing that the friction angle is 12.5% higher: 9 φ φps = — tri-axial 8 . . (3.3d) For the purposes of shear strength evaluation soils have been divided into two broad categories : Cohesionless soils (sands, silty sands, slightly clayey sands) for which the shear strength is assumed to be represented by drained conditions in terms of the angle of shearing resistance φ‘. Cohesive soils (clays, silty clays, sandy clays etc) for which the strength can be defined in terms of undrained φ‘ = 0 and c u equal to a finite value, and drained shear strength φ‘ equal to a finite value and c ‘ = 0 or a finite value. 49 Shear Strength Parameters from the In-situ Description of Soil Strength The in-situ description of soil strength can be used to obtain a rough estimate of φ‘ for a cohesionless soil and of the undrained cohesion, cu , of cohesive soil. The undrained cohesion cu is equal to half the Unconfined Compressive Strength (UCS). It is recommended that strength be described in accordance with Jennings et al (1973). Guidelines in this regard are given in TABLES 3.3.4 and 3.3.5 for cohesive and non-cohesive soil respectively. TABLE 3.3.4 Shear Strength Parameters for Slow Draining Cohesive Materials Consistency Rule of Thumb Field Identification UCS (kN/m2) Approx SPT ‘N’ S.1 Very Soft Easily moulded by fingers. Distinct heelmarks left on freshly exposed surface (Heelmark = approx: 150 kN/m2). Geologist’s pick can be easily pushed in up to its handle. < 20 <2 S.2 Soft Easily penetrated with thumb. Moulded with strong pressure. Faint heelmarks on freshly exposed surface. Geologist's pick can be pushed in up to 30 to 40mm (sharp end). 20 to 40 2 to 4 S.3 Firm Indent by thumb with effort. Very difficult to mould with fingers. Geologist’s pick (sharp end) can be pushed in up to 10mm. Slight penetration with handspade. 40 to 80 4 to 8 S.4 Stiff Penetration by thumbnail. Cannot be moulded with fingers. Geologist’s pick (sharp end) makes slight indentation when pushed in. Handpick required for excavation. 75 to 150 8 to 15 S.5 Very Stiff Indentation by thumbnail difficult. Slight indentation with blow of geologist’s pick (sharp end). Power tools required for excavation. 150 to 300 15 to 50 S.6 Hard Completely weathered extremely soft rock > 300 > 50 50 TABLE 3.3.5 Shear Strength Parameter for Quick Draining Non-cohesive Materials Consistency Rule of Thumb Field Identification Approx Approx Approx Typical φ‘ Dry CPT SPT (MPa) ‘N’ Density (kN/m3) Very Loose Almost no resistance to shovelling 0 to 2 0 to 4 Loose Easily penetrated with 12mm bar pushed by hand; small resistance to shovelling 2 to 4 4 to 10 28 to 30 14.5 to 16.0 Med Dense Easily penetrated with 12mm bar driven with 2kg hammer; considerable resistance to shovelling 4 to 9 11 to 30 30 to 35 16.0 to 17.5 Dense Hard penetration with 12 mm bar to 300mm driven with 2kg hammer; handpick required for excavation 9 to 12.5 31 to 50 35 to 40 17.5 to 19.25 Very Dense Penetration only up to 75mm with 12mm bar driven with 2kg hammer; power tools required for excavation > 12.5 > 50 26 to 28 < 14.5 40 to 50 > 19.25 Index Property Tests Many researchers have published typical ranges and correlations between the results of index property tests (Atterberg Limits and Particle Size Distribution) and the effective shear strength of soils (mainly cohesive soils with plasticity index greater than 7). The following procedures are recommended: . . The effective angle of friction can be obtained using the relationship given by Kenney (1959). This relationship is given in FIGURE 3.3.4. Although this relationship is for normally consolidated soils the effective angle of friction should not be much different for over-consolidated soils. If the nature of the soil is such that it is considered appropriate to use an effective cohesion greater than zero for analysis purposes then the cohesion values given in TABLE 3.3.6 should be used as a guide. The values given in TABLE 3.3.6 are for compacted soils and should be taken as the upper bound values for natural soils. It is also necessary to point out that for most soils caution must be exercised if an effective cohesion value greater than zero is to be used for design purposes. The values given in TABLE 3.3.6 for friction angle can also be used as a check on the values given in FIGURE 3.3.4. 51 φ‘ FIGURE 3.3.4 Plasticity Index versus sin φ ‘ after Kenney (1959) TABLE 3.3.6 after NAVFAC DM7 (1971) Group Symbol GW GC Soil Type Well-graded clean gravels, gravel-sand mixtures Typical Strength Max γd Optimum Characteristics (kN/m3) Moisture cu c‘ φ‘ tan (%) (kPa) (kPa) (deg) φ‘ 19.7 - 21.2 11 - 8 0 0 > 38 > 0.78 Clayey gravels, poorly 18.1 - 20.5 graded gravel-sand-clay 14 - 9 0 0 > 31 > 0.6 SM Silty sands, poorly graded sand-silt mix 17.3 - 19.7 16 - 11 50 5 34 0.67 SC Clayey sands, poorly graded sand-clays 16.5 - 19.7 19 - 11 75 10 31 0.6 CL Inorganic clays of low to medium plasticity 15.0 - 18.9 24 - 12 85 12 28 0.54 ML Inorganic silts and clayey silts 15.0 - 18.9 24 - 12 65 10 32 0.62 CH Inorganic clays of high plasticity 11.8 - 16.5 36 - 19 100 12 19 0.35 Estimating Effective Strength Parameters from UCS Testing An important link between UCS testing and effective strength parameters exists, and should be noted whenever UCS tests are undertaken. Photos of the ‘failure’ samples should always be taken, and the angle of the failure plane measured if clearly defined. As the angle of the failure plane is known to be at φ’+ 45/2 to the horizontal plane, the effective friction angle can easily be calculated. Once φ‘ is known the effective cohesion is calculated from: c’ = UCS (1 -- sin φ‘) 2 cos φ‘ After Capper and Cassie (1949). This is shown in FIGURE 3.3.5. 52 (3.3e) FIGURE 3.3.5 Relationship between UCS and Effective Strength Parameters 53 Shear Strength Parameters from In-situ and Penetration Tests Empirical relationships between soil shear strength (φ‘ ) for non-cohesive and c u for cohesive soils, and penetration or in-situ test values have been put forward by many authors in different parts of the world. For non-cohesive soils two approaches have been adopted where relationships have been developed (i) dependent on vertical effective stress and (ii) independent of vertical effective stress. FIGURES 3.3.6 (a) and (b) give values of φ‘ independent of Po‘ while FIGURES 3.3.7 (a) and (b) show values of φ‘ dependent on Po‘ for both SPT and CPT tests. φ‘ φ‘ (a) (b) FIGURE 3.3.6 φ ‘ Independent of Vertical Stress after (a) Peck et al (1974) and (b) Kahl et al (1968) φ‘ φ‘ (a) (b) FIGURE 3.3.7 φ ‘ Dependent of Vertical Stress after (a) Mitchell et al (1978) and (b) ESOPT (1974) NOTE: Tabulated values should be checked to see if the method requires correction for rod energy and overburden stress. Where: = SPT corrected for 100 kPa overburden pressure N1 N60 = SPT ‘N’ corrected to 60% of the theoretical energy (N1) 60 = SPT ‘N’ corrected for both 54 The undrained cohesion c u for sensitive and normally consolidated clays has been studied by several authors and correlations with SPT results show a wide scatter as outlined by De Mello (1973) and Navfac DM7 (1971) and shown in FIGURES 3.3.8 (a) and (b). If accurate values are required for design, direct correlation should be obtained during the geotechnical investigation, or measured on undisturbed samples in the laboratory. (a) (b) FIGURE 3.3.8 c u versus ‘N’ Correlation for Soft Sensitive Clays after (a) De Mello (1973) and (b) Navfac DM7 (1971) Stroud (1974) gives correlations of SPT versus undrained shear strength for stiff insensitive clays which are considered to be applicable to a wide range of residual and transported clay soils in southern Africa. These are shown in FIGURE 3.3.9. FIGURE 3.3.9 Relationship of SPT ‘N’ versus Undrained Shear Strength after Stroud (1974) 55 Undrained shear strength correlations to CPT cone resistance values (q c) for normally and over-consolidated clays are well covered in the literature and, like the SPT test, a reasonably wide scatter is evident and is dependent on whether the clay is sensitive, normally consolidated, or over-consolidated. The equation: q c = Nk cu + σνo (3.3f) governs the relationship of qc with undrained shear strength. For normally consolidated clay N k = 15, while for sensitive clays N k can be as low as 5. Undrained shear strength correlations to the Dynamic Probe Light (DPL) test (outlined in SECTION 2.2.7) have been given by Brink et al (1982) and are summarised in TABLE 3.3.7. TABLE 3.3.7 Undrained Shear Strength Correlations with DPL Test after Brink et al (1982) Sandy Materials Description SPT ‘N’ (blows per 300mm) Dynamic Probe Light (DPL) (mm per blow) Very Loose <5 > 75 Loose 5 - 10 30 - 75 Medium Dense 10 - 30 12.5 - 30 Dense 30 - 50 5 - 12.5 Very Dense > 50 2-5 Clayey Materials 3.3.3 Very Soft <2 > 110 Soft 2-4 55 - 110 Firm 4-8 30 - 55 Stiff 8 - 15 15 - 30 Very Stiff 15 - 30 7 - 15 ROCK STRENGTH CLASSIFICATION As the shear strength of rocks is often determined by the nature of the joints, and testing of this is difficult to achieve, the Unconfined Compressive Strength (UCS) of intact rock is used as the basis for foundation design with an allowance being made for the structure of the rock mass. TABLE 3.3.8 and FIGURE 3.3.11 provide the basis for rock strength classification from profile descriptions. Point load index tests are often used to evaluate rock strength from core or block samples of intact rock. Correlation of these results with UCS values are given in FIGURE 3.3.10. Direct measurement of the unconfined compressive strength can also be carried out in the laboratory on core samples of intact rock. 56 TABLE 3.3.8 Rock Strength Classification versus UCS Classification Field Test UCS (MPa) Very Soft Rock Can be peeled with a knife, material crumbles under firm blows with the sharp end of a geological pick 1 to 3 Soft Rock Can just be scraped with a knife, indentations of 2 to 4mm with firm blows of the pick point 3 to 10 Medium Hard Rock Cannot be scraped or peeled with a knife, handheld specimen breaks with firm blows of the pick point. Hard Rock Very Hard Rock 10 to 25 25 to 70 Point load tests must be carried out in order to distinguish between these classifications; these results may be verified by unconfined compressive strength tests on selected samples Extremely Hard Rock 70 to 200 > 200 FIGURE 3.3.10 Point Load Index Correlation with UCS after Bieniawski (1973) 57 FIGURE 3.3.11 Rock Strength Classifications versus UCS 58 3.3.4 SOIL STIFFNESS EVALUATION Whereas soil shear strength parameters are used mostly to determine the degree of safety against ultimate failure, the soil stiffness is important for determining the performance of the structure under serviceability conditions. Stiffness parameters are therefore very important when the engineer is to ascertain the magnitude of movements expected under service loading, or the risk of damage to adjacent structures where deep excavations are required. .. .. . Soil stiffness consists of both the shear stiffness and stiffness in compression. These can be determined from: Direct in-situ measurement such as plate load tests Measurements in the laboratory using for example the oedometer test Empirical correlation with penetration tests Direct shear tests Shear wave velocity measurement NOTE: Both the magnitude and frequency at which the load is applied, affects the measured stiffness. Compressibility characteristics of fine grained soils are divided into immediate (elastic) settlement and long term (consolidation) settlement for drained conditions. Compressibility Moduli from Direct Methods Soil modulus can be obtained directly from tests such as the plate load test. With such tests consideration must be given to the test procedure and its constraints. Taking these factors into consideration the compressibility modulus can then be calculated from the measured stresses and strains using Poulos and Davis (1974) for various loading geometries and soil conditions. Laboratory Testing Triaxial tests with local pore water pressure and local strain measurements to avoid bedding errors should be undertaken. Alternatively oedometer tests can be used. Empirical Correlation with Penetration Tests FIGURES 3.3.12 (a) and (b) and FIGURE 3.3.13, showing soil modulus versus the SPT and CPT test results respectively, give empirically correlated drained modulus values for non-cohesive materials. Webb (1970) has carried out extensive research on the compressibility of estuarine soils on the Natal coast and drained modulus values are given by the following equations using qc values from the CPT test as well as SPT ‘N‘ values: E v‘ = 2.5 (qc + 3.2) MPa or E v‘ = 537 (N + 15) MPa (3.3g) for fine to medium sands below the water-table and Ev‘ = 1.67 (qc + 1.6) MPa or Ev‘ = 358 (N + 5) MPa (3.3h) for clayey sands with PI < 15%. Relationships of SPT ‘N’ versus drained modulus for stiff over-consolidated clays has been presented by Stroud (1974) and are given in FIGURE 3.3.14. These values are particularly useful in obtaining compressibility characteristics of stiff residual soils in the southern African region. 59 There is a wide scatter in the correlation of compressibility with penetration test values for normally consolidated and sensitive clays, and values used should be regarded with caution. Drained soil modulus values for these soils show a wide scatter and the relationship of E v‘/N varies between 300 and 2000. Good correlations of compressibility of granular soils is given by Stroud (1989) using SPT, and Menzenbach (1967) using the CPT. Stroud’s correlations take the degree of loading into account. (a) (b) FIGURE 3.3.12 Drained Modulus for Sands after (a) Stroud (1989) and (b) Menzenbach (1967) 60 FIGURE 3.3.13 Drained Modulus for Non-cohesive Soils Based on CPT FIGURE 3.3.14 Drained Modulus for Stiff Cohesive Soils after Stroud (1974) 3.3.5 SMALL STRAIN STIFFNESS Recent research has highlighted the need to take cognizance of the extremely high stiffness response of soils at low strains. In the past the need for simple modelling led to the use of a single stiffness value in the entire analysis. With the advent of numerical modelling this constraint is no longer an issue and complex analyses can readily be performed. This has led to the need to measure the variation of soil modulus at differing strain levels or modulus degradation curves. 61 3.3.6 ROCK STIFFNESS EVALUATION The modulus of rock material is related to the unconfined compressive stress qa with the ratio E r /qa showing a wide scatter of between 100 and 1000. A ratio of 300 should be used for design where no direct correlation for the rock type has been obtained. A reduction factor is usually applied to the intact modulus to obtain the rock mass modulus since discontinuities and joint infilling can markedly affect the compressibility of the rock mass. FIGURE 3.3.15 shows the wide scatter of the E r /qa ratio value for various rock types and strengths. FIGURE 3.3.15 Correlation of Rock Modulus with UCS after Peck (1976) and Deere (1968) 62 3.3.7 SUB-GRADE MODULUS The vertical and horizontal sub-grade moduli are parameters commonly used to model the lateral restraint of piles and in pavement design, but should be used with caution for the design of geotechnical structures, since the parameter is area dependent. The sub-grade modulus proposed by Terzaghi (1955) is defined as the deflection produced by a unit applied pressure on a 300 mm square plate and is given the unit k N/m 2/m. Plate load tests are commonly carried out to determine vertical and horizontal sub-grade modulus values. TABLE 3.3.9 gives typical values of horizontal modulus of sub-grade reaction, k h , for cohesive soils. In cohesionless soils the horizontal sub-grade modulus increases with depth and is given by: k h = n h x Z/B, where n h is known as the coefficient of modulus variation, Z is the depth in metres and B the pile breadth in metres. Values of n h are given in TABLE 3.3.10. TABLE 3.3.9 Relationship of Modulus of Sub-grade Reaction (k h) to the Undrained Shear Strength (cu) of Stiff Over-consolidated Clay (300 x 300 square plate) Consistency Firm Stiff Very Stiff Undrained Shear Strength cu (kPa) 50 - 100 100 - 200 > 200 Range of k h.(MN/m 3) 18 - 36 36 - 72 > 72 27 54 100 Recommended k h(MN/m 3) TABLE 3.3.10 Factors for Calculating Coefficient of Modulus Variation (n h) for Cohesionless Soil (300 x 300 square plate) Relative Density Loose Medium Dense Dense nh for dry or moist soil (MN/m 3) (Terzaghi) 2.5 7.5 20 nh for submerged soil (MN/m 3) (Terzaghi) 1.4 5 12 nh for submerged soil (MN/m 3) (Reese et al) 5.3 16.3 34 The sub-grade modulus can be estimated from load tests by using the equation for settlement of a rigid plate: ρ= therefore: k = Where: ρ υ q r E = = = = = q ρ π (1 -- υ2).q.r 2E = 2E π (1 - υ2).r (3.3i) (3.3j) the average plate settlement Poisson’s ratio (between 0.25 and 0.5) the average plate pressure the radius of the plate and the secant modulus at the appropriate stress level It should be noted that for shallow foundations with isolated loadings, the method gives satisfactory results if the correct parameter is chosen. However, for uniform loading of raft and piled raft foundations, it should be avoided, Poulos (2002). 63 3.4 PERMEABILITY BS 8004 (1986) gives the following typical values: Applicable Norms As the topic of Geotechnical Classification and Parameter Designation is broad, not all aspects can be adequately covered in one book. The reader can get further details on this important topic at the following links / references: CIRIA R143: The Standard Penetration Test (SPT) – Methods and Use (1995) CIRIA C 562: Geophysics in Engineering Investigations (2002) SAIEG: Guidelines for Soil and Rock Logging in South Africa (1999) Jennings, J.E., Brink, A.B.A., Williams, A.A.B. (1973): Revised Guide to Soil Profiling for Civil Engineering Purposes in southern Africa, Trans. South African. Inst. Civil Eng., pp. 3 -12, 15 64 4.0 FACTORS INFLUENCING THE SELECTION OF A PILE TYPE Before the design engineer can consider what type of pile is best suited to a project the following basic information is needed as a minimum requirement. .. .. Detailed geotechnical information Structural details and loadings Allowable total and differential settlements Knowledge of the site and its environs Using this information, the engineer will need to consider the following points regarding the various piling systems so that the most suitable foundation system can be chosen for the project: .. .. .. .. . .. .. .. .. .. . .. STRUCTURAL Range of pile sizes to suit the loading Founding level to meet the pile load capacity Founding level to meet the settlement criteria Spacing of piles The allowable rake of the piles if required The ability to resist tension forces if required The ability to resist horizontal forces if required The clearance from existing buildings The durability of the pile shafts SOIL PROFILE If driven, whether driving will be easy, difficult or impossible If bored, whether temporary casings will be required If bored, the difficulty in penetrating to the required depth The presence of obstructions such as boulders The founding level to meet pile load capacity The founding depth to meet settlement criteria Very soft layers which can cause problems with cast-in-situ piles Rock-sockets Presence of groundwater, and at what level The presence of aggressive groundwater The potential for pile heave during installation ENVIRONMENTAL The effects of noise pollution caused by piling equipment The effects of vibration caused by pile equipment installation 65 .. .. .. .. CONTRACTUAL Access to and on the site for piling equipment Headroom clearance on site for piling equipment The cost of the piles The cost of the pile-caps and ground beams The installation risks associated with a particular pile solution The remoteness of the site The availability of skills and plant to install the piling system Adequate plant and people resources for large contracts Most of these points and others are covered in SECTION 6.0: SUMMARY DETAILS OF PILING SYSTEMS and in SECTION 7.0: TECHNICAL DETAILS OF PILING SYSTEMS. An initial selection can be made from SECTION 6.0, but this should be checked by reading the more detailed information given in SECTION 7.0. As one can see, there are a number of factors to consider, and the assessment of some of these will be difficult for someone not experienced in piling. Should there be any doubt you are welcome to contact your local Franki office for advice. An incorrect choice of pile type can be an expensive and a risky mistake, so it is advisable to make sure beforehand. 66 5.0 CLASSIFICATION OF PILING SYSTEMS There are many types of pile, some which are extensively used, and some that are seldom used. The suitability of the various pile types to the local soil conditions and the requirements of the local codes and specifications have a strong influence on what pile types are more popular in any one country. The availability of piling plant and equipment also has a strong influence. The following is a list of the pile types that have been found to satisfy the needs of the southern African market and which suit the local soil conditions. These have been classified into three main groups: Driven Cast-in-Situ, Driven Pre-formed and Bored Cast-in-Situ. DRIVEN CAST-IN-SITU Franki Pile Driven Tube Pile (7.1) (7.2) DRIVEN PRE-FORMED Precast Pile Steel H-Pile (7.3) (7.4) BORED CAST-IN-SITU Auger Pile Underslurry Pile Continuous Flight Auger (CFA) Pile Full Displacement Screwpile Forum Bored Pile Oscillator Pile Rotapile Micropile (7.5) (7.6) (7.7) (7.8) (7.9) (7.10) (7.11) (7.12) The numbers quoted on the right (in brackets) are the sub-section numbers for a detailed description of the individual pile types and the methods of installation, and are grouped under SECTION 7.0: TECHNICAL DETAILS OF PILING SYSTEMS. With most of the main piling systems there are some variations to the installation technique which can be carried out so as to achieve a specific requirement or to overcome installation difficulties. There is a full description of these in SECTION 7.0 as well as a list of potential problems that can be experienced with each of the pile types. SECTION 6.0: SUMMARY DETAILS OF PILING SYSTEMS provides a table in which the more important details are presented in a readily referenced format. This enables easy comparison between one system and another when evaluating a suitable pile type for a specific project. 67 6.0 Ref. SUMMARY DETAILS OF PILING SYSTEMS Pile Type Nom. Shaft Diameter (mm) Typical Working Load (kN) Max. Tension Load (kN) Max. Rake Max. Depth (metres) Cost per kN per Metre 75 150 250 350 450 1: 4 1: 4 1: 4 1: 4 1: 4 8 15 18 18 18 Med / High Medium Medium Med / Low Med / Low Determined by friction 1: 4 20 - 50 High DRIVEN CAST-IN-SITU 7.1 7.2 Franki Pile Mini Light Medium Heavy Super Heavy 250 or 300 250 - 450 355 500 - 650 410 700 - 900 520 1000 - 1500 610 1600 - 2000 Steel Tube Pile 300 - 1200 Up to 10 MPa on shaft DRIVEN PRE-FORMED 7.3 Precast Pile 250, 300 750 - 2000 Determined and 350 sq by friction 1: 4 Unlimited Low 7.4 Steel H-Pile Different Steel H sections Maximum Determined Stress by friction 165 MPa 1: 4 Unlimited High BORED CAST-IN-SITU 7.5 Auger Pile 300 - 2000 3 - 8 MPa on shaft Determined by friction 1: 4 50 Low 7.6 Underslurry Pile 600 - 1500 Up to 8 MPa on shaft Determined Vert. by friction only 50 Medium / High 7.7 CFA Pile 300 - 750 Up to 6 MPa on shaft Determined 1:10 by friction 25 Low 7.8 Full Displacement Screwpile 300 - 700 400 - 3000 Determined by friction 1: 6 25 Low 7.9 Forum Bored Pile 410 500 - 750 200 1: 6 12 High 7.10 Oscillator Pile 900 - 1500 Up to 10 MPa on shaft Determined by friction 1: 4 60 High 7.11 Rotapile 255 - 610 300 - 2500 Determined by friction 1: 8 40 High 7.12 Micropile Up to 300 1: 2 30 High Up to 1000 Determined by friction NOTE: Check SECTION 7.0 for more explicit information on typical working loads. 68 SUMMARY DETAILS OF PILING SYSTEMS continued Establish- Penetration Ability ment Ability to Handle Costs Boulders Noise Pollution Levels Vibration Levels if not Pre-drilled Site Area Required Normal Headroom Required (metres) Low Medium Medium Medium Medium Fair Good Good Good Good Fair Good Good Good Good Low Medium Medium Medium Medium Low Medium Medium High High Small Medium Medium Medium Medium 7.2 19.2 19.2 19.2 19.2 Medium Good Good Medium High Medium 20 Medium Medium Poor High High Medium 21 Medium Good Good High Fair Medium 21 Medium Good Good Low None Medium 15 to 30 High Good Fair Low None Large 15 to 30 Medium Fair Poor Low None Medium 30 Medium Fair Poor Low None Medium 30 Low Good Fair Low Fair Small 3 High Excellent Excellent Low None Large 30 Medium Excellent Excellent Fair Low Medium 20 Low Excellent Excellent Low None Small 8 69 7.0 TECHNICAL DETAILS OF PILING SYSTEMS 7.1 FRANKI DRIVEN CAST-IN-SITU PILES The Franki pile has been used extensively throughout southern Africa for the past 60 years and is still today one of the most popular pile types. With a wide range of pile sizes and the advantages of the enlarged base the Franki pile is suited to structures that vary from single storey residential buildings to multi-storey office blocks. There are also some interesting variations in the installation technique which have special applications. .. .. . . . POSITIVE FEATURES The Franki pile is often a very economical system There is an extensive range of pile sizes The Franki pile has an excellent load/deflection performance Noise levels are relatively low The Franki pile has excellent tension load capacity OTHER CONSIDERATIONS Vibration associated with the driving of the piling tube and the formation of the enlarged base Pile heave in saturated cohesive soil profiles The main feature of the Franki Pile is the enlarged base formed at the toe of the pile. In forming the enlarged base the end-bearing area is increased considerably. Furthermore, the displacement achieved when expelling the plug and forming the enlarged base compacts and preloads the soil surrounding the base. Thus the end-bearing of a Franki pile in sands develops at much lower base deflections than that of a bored pile as illustrated in FIGURE 7.1.1. FIGURE 7.1.1 Base Performance Comparison between a Franki Pile and a Bored Pile 70 PILE DETAILS Mini Light Identification Medium Heavy Super Heavy Nom. Diameter of Pile Shaft 250 or 300 (mm) 355 410 520 610 Nom. Pre-drilled Diameter (mm) 300 or 350 400 450 550 650 Typical Working Load (kN) 250 - 450 500 - 650 700 - 900 1000 1500 1600 2000 Maximum Depth (metres) Minimum Pile Spacing (mm) 8 15 18 18 18 650 or 750 900 1050 1300 1500 1:4 1:4 1:4 1:4 1:4 Maximum Rake Typical Main Bar Reinforcing 4 x 10 mm; 5 x 12 mm 6 x 12 mm 6 x 16 mm 6 x 20 mm 6 x 10 mm Maximum Main Bar Reinforcing 6 x 16 mm 6 x 20 mm 6 x 25 mm 6 x 32 mm 8 x 32 mm Typical Spiral Reinforcing at 150 mm Pitch 6 mm 6 mm 6 mm 6 mm 8 mm Nominal Cover to Reinforcing (mm) 25 35 40 50 50 Max. Tension Load (kN) 75 150 250 350 450 OD of Piling Tube (mm) 248 or 305 355 406 521 610 ID of Piling Tube (mm) 229 or 285 323 366 457 546 550 1600 3000 4500 6000 Typical Hammer Mass (kg) INSTALLATION TECHNIQUE The piling rig for installing the Franki pile has an engine, a winch, a mast, an openended piling tube and a long cylindrical drop hammer which is located within the bore of the piling tube, the latter being held and guided by the mast. PLATE 7.1.2 shows one of the rigs used to install the Franki pile. The first operation is to drive the piling tube into the ground. To be able to do this a plug of gravel or sand is formed inside the tube at its toe. This is achieved relatively easily by placing a measured quantity of gravel or sand in the tube while the tube is resting on the ground and then compacting this with short drops of the hammer. Once the plug is compacted the hammer drop is increased and the tube is driven into the ground by successive blows of the hammer falling on the plug. The plug arches in the tube thus drawing the tube into the ground while at the same time preventing the ingress of water and/or soil. The number of blows to penetrate each 250 mm is monitored so as to have a record of the driving resistance of the tube. This assists in deciding on a suitable founding level for the pile. 71 FIGURE 7.1.2 72 B C D E F G A. A plug of sand / stone is placed in the piling tube and compacted with the hammer. B. The tube is driven by applying blows of the drop hammer to the plug which arches in the tube and draws the tube into the ground. C. On reaching the founding level the tube is held by the extracting gear while the plug is expelled using blows of the hammer. D. Measured quantities of relatively dry concrete are expelled from the toe of the tube thus forming an enlarged base. E. The reinforcing cage is placed in the tube which is then filled with high slump concrete. F. The tube is extracted by means of the extraction gear. On deeper piles the concrete level may have to be topped up during extraction. G. The completed pile. A FRANKI PILE INSTALLATION SEQUENCE The tube is normally driven to a pre-determined depth at which stage the penetration rate for ten standard blows is checked. The amount of penetration is referred to as the ‘set’. If the set is equal or less than that calculated then the tube has been driven to an adequate depth. The plug is then driven out the end of the tube by successive blows of the hammer while the extraction winch is used to hold the tube back from further penetration into the ground. The expelling of the plug is followed by the formation of the enlarged base. This is achieved by gradually feeding zero slump concrete into the tube while at the same time continuing with measured blows of the hammer, the hammer expelling the concrete to form the enlarged base. The volume of concrete expelled and the number of blows are recorded and these are compared with the theoretically calculated energy required to provide the load capacity. The size of the enlarged base is increased until such stage as the required energy levels are met. The final operation is to use the hammer to ensure all concrete is out of the tube. The steel reinforcing cage is lowered into the tube and the tube is filled with high slump concrete. The tube is then withdrawn using the extraction winch and a system of reeved sheave blocks to gain a mechanical advantage. With deeper piles the tube has to be topped up with additional concrete during the extraction process. Once the tube is out of the ground, the cut-off level of the concrete is checked and adjusted if necessary. The position of the steel reinforcing cage at the head of the pile is also checked and adjusted. The piling tube is also washed out before commencing the next pile so as to remove any latence on the inside of the tube. The above sequence of operations is illustrated in FIGURE 7.1.2. THE ADVANTAGE OF THE ENLARGED BASE As mentioned earlier there are numerous advantages to be gained by having the enlarged base on a Franki pile. The first of these is the increase in load-bearing capacity. With the end-bearing component of a pile's capacity being the product of the allowable end-bearing stress and the area of the base of the pile, one can conclude that if the endbearing area is doubled then so is the end-bearing component. With a Franki pile it is possible to more than double the end-bearing area by enlarging the base. PLATE 7.1.1 illustrates a typical enlarged base formed in loose sandy conditions. This is not the only advantage related to the end-bearing. By enlarging the base the surrounding soil is compacted so the ultimate end-bearing stress is also increased. Another advantage is that the enlarged base has an excellent load / deflection performance at low base deflections as shown in FIGURE 7.1.1. This is due to the fact that the expansion of the base is a form of preloading of the soil surrounding the base. The main benefit that is gained from these features is that the founding level for the Franki pile is often at a considerably higher elevation than that necessary for other piling systems. This can and often does result in an economical piled foundation based on the use of the Franki pile while pile-cap deflections remain within acceptable limits. 73 Another major advantage of the enlarged base is the large tension loads that can be resisted by the Franki pile. The enlarged base forms an ideal positive anchorage and significant tension loads can be resisted. The Franki pile is thus often used for structures with tension pile loads such as transmission towers and chimney stacks. It is also ideally suited to founding structures in heaving clay where the enlarged base formed within a stable stratum is very effective in preventing uplift movement. To be able to resist these tension loads it is essential that the steel reinforcing is cast into the enlarged base. See Anchoring Reinforcement in base in the following section for more information on this procedure. PLATE 7.1.1 Enlarged Base Formed in Loose Sandy Conditions VARIATIONS IN INSTALLATION TECHNIQUE Anchoring Reinforcement in Base Piles are often required to take tension loads. Because the Franki pile has an enlarged base a considerable tension capacity can be generated provided the steel reinforcement is anchored into the base. To achieve this the following variation to the installation technique takes place after the plug has been expelled. Using a 20 mm slump concrete a 1.5 to 2.0 metre length of pile shaft is formed using the rammed shaft technique described hereunder. A slightly wetter mix is used and the concrete is extruded from the tube with the hammer as the tube is slowly withdrawn. The tube is then re-plugged with normal zero slump base concrete and driven back through the shaft. As a result of this the shaft concrete is forced sideways and an enlarged base is formed. The plug is then expelled and further enlargement of the base is achieved using the standard basing technique. The reinforcing cage is then placed into the tube and the shaft concreted in the normal way. Franki piles made in this manner can resist considerable tension loads and ultimate tension load tests in excess of 1000 kN have been successfully carried out. Rammed Shaft The rammed shaft has an uneven surface which results in about fifty percent greater shaft friction than that of the smooth standard shaft formed using high slump concrete. The only reason for using the rammed shaft would be to increase shaft friction and as the Franki pile is generally considered an end-bearing pile it is not widely used. 74 The rammed shaft is formed by expelling successive charges of zero slump concrete out of the tube as the latter is gradually withdrawn. Each measured concrete charge is placed in the tube and the hammer is lowered to rest on the concrete. The tube is then withdrawn about one diameter and the concrete is given one short test blow of the hammer. The tube is then withdrawn an additional amount depending on the resistance of the ground as determined by the test blow. The concrete is given another three to four blows of the hammer by which stage the toe of the hammer should be level with the toe of the tube. Another charge of concrete is lowered into the tube and the process is repeated until the shaft has been cast to the correct level. Because of the stiffness of the concrete mix pile cut-off levels well below ground level are possible with the rammed shaft method. It should be noted that the reinforcing cage is placed in the tube prior to the concreting of the shaft and that the hammer operates within the reinforcing cage. The cage has to be well made so as to prevent damage from the hammer. There is also limited control on the concrete cover with this type of shaft. Problems have been experienced when using this shaft technique in very soft soil profiles. The resistance to the concrete being expelled from the tube is very low with the result that it tends to flow out in one direction more than another. This can result in abnormally shaped piles and piles with no cover to the steel in places. Similar problems can occur where the piles have been pre-drilled and soft saturated soil fills the annulus when the piling tube is installed. Predrilling, Jetting and Coring In the interior of southern Africa the soils are normally not saturated and are often of dense or stiff consistency. It is thus very difficult and in many cases impossible to drive the piling tube into these soil profiles. To achieve the required penetration the pile position is first pre-drilled using an auger rig. The depth to which the pile is pre-drilled will depend on whether there is any tendency for the pre-drilled hole to collapse, the consistency of the soil profile and any tendency for pile heave to take place. It is standard practice to attempt to drive the tube beyond the pre-drilled depth so as to ensure that the pile's end-bearing is not affected by the pre-drilling. In the coastal areas where the soil profiles are generally saturated, pre-drilling is often not possible. In these circumstances jetting with a water-jet pipe can be used to assist the penetration of the tube through a dense stratum. A water-jet is typically a 100 mm pipe with a nozzle on the toe-end and connected to a high pressure high volume pump. The jet pipe is lowered down the side of the piling tube as the latter is driven into the ground. By keeping the jetting action close to the toe of the piling tube the driving resistance is considerably reduced. To keep the piling tube plumb it is common to use two jet pipes, one either side of the tube. It is standard practice to drive the tube two to three metres beyond the level at which jetting was discontinued so as to ensure full end-bearing capacity. The alternative to jetting in saturated soil profiles and in particular in saturated cohesive profiles is the coring technique which is simply removing the soil from the bore of the tube while the latter is top driven into the ground using a drop hammer. The process is messy and slow but on occasions, such as, to avoid pile heave in saturated cohesive profiles, it provides the best solution. 75 Conditioning the Soil In sandy soil profiles the mere act of driving the piling tube has a conditioning effect on the soil. As more piles are driven so the soil strength increases and the ground is said to ‘tighten up’. A greater degree of conditioning can also be achieved in softer soil profiles by first making a sand pile and then driving back through this. This achieves a double volume displacement and thus a greater degree of conditioning. Another way a sandy soil can be conditioned is through the process of forming the enlarged base. By expelling material from the toe of the tube a sandy soil can be compacted. The material expelled can be concrete, sand or gravel. If the latter, a concrete base is finally formed in the compacted zone in the normal way. Additional Depth One of the limitations with the Franki piling system is the limited depth to which the standard pile can be installed. This is 8 metres for the Mini, 15 metres for the Light and 18 metres for the Medium, Heavy and Super Heavy. If the water-table is not high these depths can be increased using an extension tube. This is merely an additional length of piling tube which is attached to the top of the normal piling tube and which enables additional depth to be achieved. An additional one to two metres for a Mini is common and up to 5 metres for the other sizes. Using an extension tube does however slow down the production rate so the cost per metre will rise. Comparison of costs with other piling systems will determine whether the use of extension tubes is an economical proposition. Because it is difficult to seal the joint between the extension tube and the main piling tube the use of this technique is not possible in areas with a high water-table. Permanent Liner to Pile Shaft In cases where the groundwater is polluted with chemicals that are harmful to concrete it is desirable to have a permanent liner which protects the shaft concrete from attack. Liners made out of steel or plastic can be incorporated into the shaft of the pile. These liners are normally fixed to the reinforcing cage and lifted up and lowered into position with the cage. The shaft is then concreted in the normal way. Precast Concrete Shaft A Franki pile can also be formed using a precast concrete shaft with the resulting pile being referred to as a Franki Precast Composite. A precast concrete shaft is a high quality product and the Franki Precast Composite pile is ideally suited to sites with aggressive groundwater conditions. The precast shaft is also smooth and in the formation of the pile there is an annulus of loose material created around the shaft. In the case of negative friction conditions where the ground surrounding the pile is settling and causing downdrag, these two factors assist considerably in reducing the downdrag force. 76 A small quantity of sand / cement grout is placed in the tube once the enlarged base has been formed. The precast shaft is lowered into the piling tube and penetrates into the grout to bear on the enlarged base. The piling tube is extracted and the gap around the precast shaft is filled with loose sand so as to provide some lateral support to the shaft. POTENTIAL PROBLEM AREAS Pile Heave This is the phenomenon in which a previously installed pile is lifted by upward movement of the soil surrounding it caused by the driving of an adjacent pile. It occurs in saturated clayey and silty soils, but does not occur in clean sand. It only occurs with driven displacement type piles and not with bored piles. Where pile heave lifts the whole pile it is generally thought that bearing capacity is not affected materially. In some cases, however, pile heave has been found to be detrimental to the bearing capacity of the pile. This is believed to be due to the separation of the shaft from the base due to insufficient tension transfer mainly caused by low bond figures in the green concrete. By test loading a pile that has heaved one can establish whether the heave has been detrimental to the pile's bearing capacity. If this proves negative then the problem can be ignored although any simple measures to reduce the heave are nevertheless advisable. Pre-drilling or coring a pile as discussed in the VARIATIONS IN INSTALLATION TECHNIQUE can largely reduce the amount of heave because displacement takes place over a reduced depth. If pre-drilling or coring is carried out for the full pile depth then the risk of shaft-base separation is largely eliminated. Heave may still take place, but this will lift the base of the pile and from experience is not usually detrimental to the pile's bearing capacity. Another simple measure that can reduce or eliminate pile heave is to leave the pile for three to four days before driving the pile immediately alongside. This time delay enables the bond between the reinforcing bars and the concrete to build up. Welding a shear key to the steel to improve the bond is a further possibility. When considering the use of a driven displacement pile in saturated cohesive soils the number and spacing of the piles in the group is an important consideration. A single tank base with a large number of piles in one group is a real potential problem and more suited to a bored pile solution. An open warehouse type structure with groups of two and three piles should not present a major problem. The spacing of the piles in the pile-cap should be increased if pile heave is a potential problem. 77 Vibration During Driving and Forming of the Enlarged Base The act of driving a pile causes a certain amount of vibration. In general the smaller the pile the smaller the energy applied and the less the vibration. Exceptions to this rule have occurred and are thought to occur when the frequency of driving is resonant with the natural frequency of the ground. Normally, however, the vibrations experienced with a Light pile would be much lower than those experienced with a Heavy. The vibration levels generated by the Franki system are generally not that severe. They can and have resulted in minor cracks forming in buildings and the extension of existing cracks. The discomfort of feeling the vibration is not normally a problem but the longer the contract the more sensitive people become. Measures such as pre-drilling, jetting and coring can be used to reduce the levels of vibration. The vibration experienced when forming the base is however always there, but generally of a lower level than when driving because the energy levels are lower. Contracts close to residential buildings should be avoided unless the piles are pre-drilled, jetted or cored. The more sensitive parts of city centres should also be avoided. Noise Pollution Noise levels are not much above that of the main engine noise. There is the thump from the hammer impact but this takes place inside the tube so is fairly muffled. The odd clang of a wire rope hitting the side of the mast does not seem to worry people. Noise is generally not a problem with the Franki system. Artesian Conditions The risk of artesian conditions is very low but they are known to occur. When a cast-insitu pile is formed through an artesian layer the groundwater which is under pressure tends to travel up the side of the pile and in so doing washes out the cement in the concrete. The amount of defective shaft resulting from this action will depend on how strong the water source is. In the one recorded case of artesian conditions the effect on the pile shafts was serious. Artesian conditions should be identified and reported by the drilling operator during the geotechnical investigation. This condition should then be fully investigated. A Franki pile with a precast shaft, or a permanent casing, or one of the pre-formed piling systems, could provide a better choice of pile type if artesian conditions are present. 78 PLATE 7.1.2 Franki Crawler Rigs at Bank City, Johannesburg CBD, South Africa 79 7.2 DRIVEN TUBE PILES This system is not used extensively mainly due to the high cost of the steel tubes. There are however situations where the positive features of the system outweigh the costs. Small pile sizes are commonly used for underpinning houses and light buildings with limited headroom and poor access. Medium pile sizes are commonly used for piling new column foundations within existing buildings or in difficult access areas. The larger sizes are used mostly for river bridge foundations and in marine construction. .. .. .. . .. POSITIVE FEATURES There is an extensive range of pile sizes The system can achieve considerable depths ( > 60 metres in a suitable profile) The pile is permanently cased and thus ideal for river and marine work The pile can be installed in limited headroom The pile can be installed in areas with very difficult access The shaft is cast in the dry so quality control is good Noise levels are not high OTHER CONSIDERATIONS It is a relatively expensive system There is vibration associated with the driving of the tube Steel piling tubes can be installed either open-ended or closed-ended. With the closedended technique the toe-end of the piling tube is sealed off with a steel plate so that there is full displacement during driving. It is this more common technique which is covered in this section. The use of open-ended steel tube piles is associated with temporary staging structures as well as marine work and is covered in the Marine Engineering section of this book. PILE DETAILS The working load shaft stress is generally in the range 8 to 10 MPa. Shaft stresses of up to 16 MPa have been used on deep piles where there is a significant friction component and the piles are driven onto a competent founding stratum. The strength of the shaft concrete has to be raised in line with the higher shaft stress and the contribution of the steel tube’s load capacity taken into account. 300 (mm) 400 (mm) Pile Diameter 500 600 750 (mm) (mm) (mm) 900 (mm) 1200 (mm) Typical Working Load (kN) 400 700 750 1250 1150 1750 1700 2750 2750 4250 3750 6500 6750 11500 Maximum Depth (metres) 20 40 50 50 50 50 50 Pile Spacing (mm) 750 1000 1250 1500 2000 2500 3000 Maximum Rake 1: 4 1: 4 1: 4 1: 4 1: 4 1: 4 1: 4 Typical Main Bar Reinforcing Typical Spiral Reinforcing at 150 mm Pitch 5 x12mm 5 x16mm 6 x16mm 6 x 20mm 6 x 25mm 8 x 25mm 14 x 25mm 6 mm 6 mm 6 mm 80 8 mm 8 mm 10 mm 10 mm STEEL PILING TUBES Either barrel type tubes or spirally manufactured tubes can be used provided the manufacturing process, and in particular the welding, has been carried out to high standards with stringent quality control. The tubes should be manufactured according to SABS 719 Grade B which is the equivalent of the American API 5L Grade B specification which is used worldwide. Before installation can commence the sections of piling tube have to be prepared and welded up. The lead section of tube has a steel plate or rock shoe welded to the toeend and a splicing band welded to the top end. Follower tubes need only the splicing band to be welded to one end. The wall thickness of the tube will be in the 4.5 to 16 mm range. The smaller diameters will most likely have a 5 or 6 mm wall thickness throughout, whereas the larger diameters will have a 10 - 16 mm wall at the toe reducing to 6 mm at the top. The splicing bands should be about 250 mm wide and should be made out of plate of the same thickness as the wall of the tube. If two tubes of different thickness are being welded together then the plate thickness should match the thicker of the two. Ideally there should not be more than a 2 mm difference in the wall thicknesses of two sections of tube to be welded together. The diameter of the end-plate should exceed that of the tube by the OD plus 12 mm. Alternatively, the plate can be welded on the inside of the tube in which case the friction component of the pile's bearing capacity will be greater. The end-plate can have a rock shoe type arrangement welded to it. Rock shoes should be used if a boulder layer has to be penetrated or if the pile is to be driven onto a sloping rock face. An effective rock shoe can be manufactured using a central pin and four heavy gussets welded to the end-plate as shown in FIGURE 7.2.1. FIGURE 7.2.1 Driven Tube Pile Details 81 The tubes can be either top or bottom driven during the installation process. If bottom driving is used there is obviously more stress placed on the welds, both circumferential as well as longitudinal. Bottom driving is more efficient however, and is often resorted to for the final drive, even if the initial penetration is achieved with a diesel or hydraulic hammer driving on a helmet on the top of the tube, or vibrated in with a heavy vibrator. INSTALLATION TECHNIQUE Bottom Driven The lead section of tube is lifted up and positioned in the leader of the piling rig. A leader guiding mechanism must be clamped to the tube and the leader adjusted for verticality or rake. With the toe of the tube resting on the ground on the pile position a measured quantity of semi-dry concrete is discharged into the tube. The initial quantity should fill the tube about 3 to 4 diameters. The hammer is a cylindrical drop hammer which operates within the bore of the piling tube. The plug of concrete is compacted using a few short drops of the hammer. The drop is then gradually increased but is normally not greater than 2.5 metres. The plug material becomes pulverised during driving and this lowers the efficiency of the blow of the hammer. For this reason, and to prevent the tube from splitting, the plug has to be continually refreshed by adding additional plug material throughout the driving operation. The next section of tube is welded on when the top of the leader tube is at a convenient height for welding. Further sections are welded on, as and when required. When the toe of the tube is approaching the founding stratum, measurements of the penetration rate are made and the set checked occasionally. When the required set has been achieved this signifies the completion of the driving and the hammer is removed from the tube. It is possible to inspect the internal bore of the tube using a light or a mirror and reflected sunlight if necessary . The reinforcing cage is then lowered into position and the shaft of the pile concreted using a high slump self-compacting mix. It is common practice to reinforce only the upper 12 metres of the pile shaft because of the permanent thick-walled casing. Top Driven If top driving is used, the tube is lifted into the leader of the piling rig as previously described, the guides are fixed, and the leader is adjusted for verticality or rake. The helmet and hammer are then lowered onto the head of the tube and alignment of the hammer and tube is checked. Driving commences with the operation of the hammer and is continued, save for welding on follower tubes, until the required founding stratum has been reached and an adequate set achieved. 82 Top driving of long closed-ended steel piling tubes with a thin wall is not as efficient as bottom driving in hard driving conditions, as a large proportion of the hammer's energy is absorbed by the tube itself. If the driving is particularly hard, or the piles are heavily loaded, then it may be necessary to resort to bottom driving to achieve the required depth or set. The efficiency of bottom driving is estimated to be between 15 and 25 percent. Once the tube is driven, it is a simple matter to place the reinforcing cage and concrete the shaft as described above. PLATE 7.2.1 shows driven tube piles being installed on a river bridge foundation. VARIATIONS IN INSTALLATION TECHNIQUE Pre-drilling and Jetting If hard driving is experienced this can be relieved to some extent by pre-drilling or jetting the pile through the dense layers. This relief can only be achieved over the top 15 metres if jetting is used, and 36 metres, if pre-drilling is adopted. POTENTIAL PROBLEM AREAS Split Pile Tubes There is no doubt that this is the main risk when installing this pile type. The splitting of the odd casing is to be expected and there are ways to remedy this. If a number of casings split however, then the problem can be very serious. The risk of splitting a casing should not be high provided the casings are manufactured to high standards of quality control, the appropriate casing thicknesses are used, and the correct driving procedures are adhered to. As the welds in the casing are potential weak areas it is sensible to keep the number of welds to a minimum. Tubes made up of long barrels are thus preferred over those made using the spiral technique. The standard of on-site welding must be high, and only skilled welders should be used. During driving it is most important to refresh the plug regularly. The impact of the hammer tends to enlarge the casing, so if the plug is not refreshed, the casing will continue to expand until it fractures. By refreshing the plug, the point of impact of the hammer is raised and another section of the tube comes under bursting stresses caused by driving. A sufficient length of heavier wall casing must be used at the toe of the pile so that the impact of the hammer stays within this section. As stated earlier the risk of casing fracture can be reduced if top driving is used. The efficiency of top driving is however much lower, and top driving is very noisy. 83 Pile Heave Being a displacement type pile, pile heave may be experienced in saturated cohesive soil profiles. Heave checks should be carried out and if heave is occurring then the concreting of the piles should be delayed until all the piles in a group have been driven. They should then all be re-driven so as to eliminate the heave and any negative effects it may have on the pile's bearing capacity. Vibration During Driving As large hammers are used on the larger pile sizes a considerable amount of vibration can occur with this pile type. The severity of the problem can be reduced by using short hammer drops, pre-drilling the pile positions, or jetting alongside and below the tube during the driving operation. Noise Pollution With bottom driving the noise levels are not high and are similar to that experienced with a Franki pile. Top driving, on the other hand, is noisy and should be avoided in built-up residential and commercial areas. 84 PLATE 7.2.1 Driven Tube Piles Being Installed on a River Bridge Foundation 85 7.3 PRECAST PILES The Precast pile was one of the first piling systems to be used in southern Africa, there being a record of the piling to the old Putt Bridge in Port Alfred (now demolished) where the precast piles were installed using a steam operated piling machine, back in 1908. Modern technology has introduced the jointing of precast piles which has overcome the original depth limitations and increased the use for this pile type. Precast piles have a wide use from bridges to commercial and industrial buildings. Because of the high noise levels associated with the driving of precast piles they are seldom used in heavily populated residential areas or in downtown city centres. .. . .. .. POSITIVE FEATURES Precast piles can provide an economical solution, especially in deeper soil profiles A precast pile shaft is a higher quality product than an in-situ shaft Installation is quick and control on site is good OTHER CONSIDERATIONS There is considerable noise associated with the driving of the precast pile There is vibration associated with the driving of the precast pile Damage or fracture of precast pile shafts can occur during driving Pile lengths have to be accurately assessed, well in advance, to avoid excessive wastage PILE DETAILS Pile Size 250 mm Square 300 mm Square 350 mm Square Typical Working Load (kN) 800 - 1000 1200 - 1500 1600 - 2000 Maximum Depth (metres) Unlimited Unlimited Unlimited Minimum Pile Spacing (mm) 750 900 1050 Maximum Rake 1: 4 1: 4 1: 4 Typical Main Bar Reinforcing 4 x20mm 8 x16 mm 8 x 20 mm Maximum Main Bar Reinforcing 4 x25mm 8 x 20 mm 8 x 25 mm Nominal Cover to Reinforcing (mm) 30 35 40 The above three sizes are the most common in jointed precast piles. If single length piles are to be used then the size can be made to suit the contract. The cost of the shutters will however be considerable, so this is only economical on a large contract. Common sizes for single length precast piles are 350 and 400 mm square and 350 by 400 mm rectangular. Single pile lengths of up to 16 metres in ordinary reinforced concrete are common. Longer pile lengths of up to 30 metres are possible using prestressing but this technology is not in common use in southern Africa. 86 PRECAST SHAFT MANUFACTURE The shafts of precast piles are normally cast in a factory and transported to the site as this is the most economical arrangement. On larger contracts especially in remote areas it may well be more economical to set up a casting yard on site. Before deciding on this the local aggregates should be tested so as to ensure that concrete of a sufficiently high quality can be produced. The moulds in which the shafts are cast are made of steel and have a slight taper on them so the shaft can be extracted without having to strip the mould. The moulds are sometimes designed to allow the passage of steam through them for accelerated curing. It is extremely important that the mechanical joints are very accurately aligned so that when sections are coupled together on site the shaft of the combined pile is straight. The moulds are fitted with devices for locating and aligning the mechanical joints and rock shoes. The concrete used in the manufacture of precast concrete pile shafts has to be of the highest quality and thus the aggregates used have to be likewise. The stone, in particular, has to be of good quality and clean, which may require washing the stone. As the design 28 day strength is 50 to 60 MPa the cement content is of the order of 450 kg per cubic metre. Additives are generally used to assist in obtaining these high strengths. The main bar reinforcing steel is high tensile with a mild steel spiral which is generally at 150 pitch except near the ends of the shaft unit where the spacing is reduced to cope with the higher bursting stresses. To ensure that the full moment resistance of the shaft is transmitted through the mechanical joint, splice bars which are threaded into the joint plates extend for the full bond length back from the joint plate. For handling the pile once it has been cast lifting lugs are cast into the pile at fifth points from each end. Piles are often steam cured by passing steam through the shutters while the pile bed is covered with a heat retaining blanket. The timing of the steaming process and the control of the temperatures must be carefully controlled in accordance with recommended steam curing procedure. Piles that are removed from the casting beds are stacked one upon the other in a holding yard while the final curing takes place. When stacking precast pile shafts it is important to support the piles on bearers at fifth points from the ends. All the bearers for a stack of precast shafts must be exactly in line with those of the bottom row or else moments are induced into the shafts and cracking can take place. The overall stability of a stack of precast shafts is also important. When transporting the pile shafts and when stacking on site it is also important to stack the piles correctly. When lifting the pile shaft for driving, a single lift point one quarter from the end is normally used. PLATE 7.3.1 shows a precast pile manufacturing operation. 87 MECHANICAL PILE JOINTS There are a number of internationally patented mechanical pile joints available and there are also a few locally developed joints. It is not the subject of this text to examine these in detail but the structural essentials of a good joint are: . . . . An ability to transfer the full shear force and bending moment of the concrete section without overstressing any component of the joint and its splice bars An ability to transfer the full axial compression and tension forces both static and dynamic without overstressing any component of the joint and its splice bars An ability to transfer the full axial compression force both static and dynamic without overstressing in the pile shaft concrete or causing spalling The faces of the joint must be at right angles to the longitudinal axis of the pile shaft to ensure that the jointed shafts have one common longitudinal axis. For this to be achieved the joint must have a device for accurate location in the shutter when casting the pile PLATE 7.3.2 shows a mechanical precast pile joint. An alternative form of mechanical joint is the welded joint. In this form the joint consists of two 25 mm thick mild steel end-plates one of which is chamfered to take the weld. The normal group of splice bars are threaded into the end-plates so as to transfer the moment. When installing the pile a single run of an automatic feed welding machine is normally sufficient to ensure an adequately strong joint. ROCK SHOES Precast piles that are required to penetrate cobble, or boulder layers, or required to found on hard rock or a sloping rock face, are normally fitted with rock shoes. Rock shoes can vary from a simple end-plate which is merely there to protect the concrete to a sophisticated shoe with a hardened steel point which is design to key into a sloping rock face. PLATE 7.3.3 shows a typical rock shoe. INSTALLATION TECHNIQUE The piles are delivered to site and stacked close to the pile positions. The lead pile section is picked up by the rig and located in the piling helmet which is fitted to the leader of the piling rig. The pile section together with the helmet is then lowered so the toe of the pile is on the pile position. The leader is then adjusted for verticality or rake and the driving hammer is lowered onto the piling helmet. The pile can be driven with a drop, diesel or hydraulic hammer. Once the pile is a metre in the ground the verticality or rake is checked again and final adjustments are made. For the remainder of the driving operation the technique is to follow the pile, as any attempt to try and correct the position and rake tolerances can result in cracking of the pile shaft through induced moments. 88 When the first section of pile shaft is in the ground the second section is lifted up and located in the helmet as was the first. The shaft with the helmet is lowered onto the first pile shaft assisted by a locating pin. The wedges that lock the two joint plates together are then driven home. The alignment of the leader is checked and driving recommenced. If additional lengths are required the process is repeated for the additional shaft lengths. The precast pile is traditionally driven to a set as are most driven piles. The set for long slender precast piles should be calculated using the Wave Equation method. The traditional Hiley formula for calculating pile sets can be used for short precast piles. The final set and temporary compression diagram should be recorded as part of the pile record. It is recommended that the set on some of the piles on a contract are checked 24 hours later so as to ensure that there is no increase in the set which can occur in certain clayey soil profiles. If this is found to be the case the pile must be redriven. In sandy soil profiles, the set taken 24 hours later is normally considerably less than that taken at the time of driving, due to the fact that the static friction is much greater than the dynamic and pile shaft resistance increases with time. PLATE 7.3.4 shows precast pile installation in progress. VARIATIONS IN INSTALLATION TECHNIQUE Pre-drilling and Jetting To obtain relief from hard driving and to assist in reducing the noise levels pre-drilling and jetting may be resorted to. Slipcoating of Pile Shafts Shell Slipcote is a form of bitumen which is used to coat the sides of pre-formed piles with the object of reducing the friction on the sides of the pile. It is used in negative friction situations where settlement of ground around a pile results in additional load being transferred to the pile by the fact that the friction is acting in reverse. Hence the term negative friction. The slipcoating of precast pile shafts is an involved process. The shafts have first to be coated with a priming coat which can be brushed or sprayed on. The shafts are then placed in steel moulds which allow a 10 mm gap all round the shaft. Spacers are used under the shaft and on the sides so as to ensure the 10 mm gap. The slipcoat bitumen is heated and poured into the moulds so that it fills the 10 mm gap. The mould when full also allows for a 10 mm thickness on the upper surface. Once the bitumen has cooled the pile shaft is lifted out of the mould and stored. When cool the slipcoat layer is quite stiff and is not wiped off during the driving of the pile. The product's ability to reduce friction is very impressive and negative drawdown can virtually be eliminated. 89 POTENTIAL PROBLEM AREAS Pile Heave Precast piles, being displacement type piles, can experience pile heave. If heave has been measured, then a selected number of piles that have heaved should be re-driven to check whether the set has increased from that of the initial drive. If it has, then all the piles should be re-driven. Noise Pollution The driving of precast piles is a noisy process and due consideration must be given to the potential for objections from the public and possible legal intervention. This should only present a problem in residential and downtown city areas. Vibration During Driving The vibration caused by driving can be a problem, but it is very difficult to forecast beforehand how severe the vibration will be. The general approach from a contracting point of view is to take the risk, which is relatively low, and sort the problem out in the unlikely event of one arising. Due cognizance should be taken however of the condition of surrounding buildings and whether these buildings are occupied by businesses whose operations are susceptible to vibration. Shaft Breakage A precast shaft has to withstand fairly severe driving stresses during the installation process. As a result it is not uncommon to have the occasional shaft fracture during driving. With well-made pile shafts and the correct driving technique the fracture rate should not be greater than one percent. Extremely hard driving over a considerable depth will increase the risk of shaft breakage. The affected pile is normally replaced with one or two additional piles using a revised pile layout. Sonic testing of precast piles is a simple and effective way for checking whether the shafts have been damaged or fractured during driving. See SECTION 9.0: PILE LOAD AND INTEGRITY TESTING for details of this form of testing. 90 PLATE 7.3.1 Precast Pile Manufacture PLATE 7.3.2 Precast Pile Joint PLATE 7.3.3 Precast Pile Rock Shoe 91 PLATE 7.3.4 Precast Pile Installation in Port Louis, Mauritius 92 7.4 STEEL H-PILES Steel H-Pile sections are rolled by steel mills in South Africa and abroad and are thus available for installation as piles. As the cost of steel is relatively high, it is not a widely used pile type due to the fact that it is not normally an economical solution. It has advantages in that the steel sections can be driven through soil profiles which have minor obstructions in the form of cobbles and small boulders as well as fill materials containing builders’ rubble. .. . .. .. POSITIVE FEATURES The pile has good penetrating ability The pile can be installed to considerable depth H-sections are ideal for use as soldier piles OTHER CONSIDERATIONS Steel H-sections are relatively expensive Noise levels are high if the pile is driven The end-bearing capacity is limited unless driven onto rock Corrosion protection needs careful consideration PILE DETAILS The H-section is rolled in two forms: Universal Columns and Universal Bearing Piles. Details of these sections are given in FIGURE 7.4.1. The working load shaft stress can be up to 125 MPa for Grade 43 steel and up to 165 MPa for Grade 50B steel. Using the higher stress the range of working loads varies between 500 to 3300 kN for the sections available. There are limited rollings of bearing pile sections in our region and availability of the required section should be checked with suppliers before being specified. INSTALLATION TECHNIQUE The first section of shaft is lifted up by the piling rig and located into the piling helmet which is then released from the hammer. The first shaft section together with the helmet is then lowered down onto the pile position and the hammer positioned to rest on the helmet. The leader of the piling rig is adjusted for verticality or rake. The pile can be driven with either a vibratory, drop, diesel or hydraulic hammer. Once the first shaft section has been driven the second one, which has been prepared for welding, is lifted up and then lowered to line up with the first section. Guiding lugs tack-welded to the first section assist in this operation. A butt-weld is then carried out. Once the initial weld has been made the piling rig can move off and carry on driving another pile while the full weld is completed. Driving is then continued and the cycle is repeated if further lengths are required. 93 B Y HIGHVELD UNIVERSAL BEARING PILES T D t d r X X Y Serial Size Weight Depth of Width of Web Flange per Section Section Thickness Thickness Metre D B t T (kg) (mm) (mm) (mm) (mm) (mm) Root Area of Moment Moment Rad Section of of r Inertia Inertia 2 (mm) (cm ) x- x y- y (cm4) (cm4) 305 x 305 110 79 308 299 310 306 15.4 11.1 15.4 11.1 15.2 15.2 140.0 100.4 23580 16400 7688 5290 254 x 254 85 63 254 247 260 256 14.3 10.6 14.3 10.6 12.7 12.7 108.1 79.7 12264 8775 4188 2971 203 x 203 54 204 207 11.3 11.3 10.2 68.4 4987 1683 B Y r d HIGHVELD UNIVERSAL COLUMNS X X T D t Y Serial Size Weight Depth of Width of Web Flange per Section Section Thickness Thickness Metre D B t T (kg) (mm) (mm) (mm) (mm) (mm) Root Area of Moment Moment Rad Section of of r Inertia Inertia 2 (mm) (cm ) x- x y- y (cm4) (cm4) 305 x 305 158 137 116 97 327.2 320.5 314.5 307.8 310.6 308.7 306.8 304.8 15.7 13.8 11.9 9.99 25.0 21.7 18.7 15.4 15.2 15.2 15.2 15.2 201.2 174.6 149.8 123.3 38740 32838 27601 22202 12524 10673 9006 7268 254 x 254 112 89 72 60 49 289.1 276.4 266.7 260.4 254.0 264.5 261.0 258.3 255.9 254.0 19.2 15.8 13.0 10.5 8.6 31.7 25.3 20.5 17.3 14.2 12.7 12.7 12.7 12.7 12.7 212.4 168.9 136.6 114.0 92.9 29914 22575 17510 14307 11360 9796 7519 5901 4849 3873 203 x 203 86 71 60 52 46 222.3 215.9 209.6 206.2 203.2 208.8 207.2 205.2 203.9 203.2 13.0 10.3 9.3 8.0 7.3 20.5 17.3 14.2 12.5 11.0 10.2 10.2 10.2 10.2 10.2 110.1 91.1 75.8 66.4 58.8 9462 7647 6088 5263 5464 3119 2536 2041 1770 1539 152 x 152 37 30 23 161.8 157.5 152.4 154.4 152.9 152.4 8.1 6.6 6.1 11.5 9.4 6.8 7.6 7.6 7.6 47.4 38.2 29.8 2218 1742 1263 709 558 403 FIGURE 7.4.1 Details of Universal Column and Bearing Pile Sections 94 The pile is driven to a set which has been calculated to provide adequate bearing capacity. The set should be checked after 24 hours so as to determine whether there has been any relaxation. If there has, the pile should be driven until the required set has been achieved and again checked after 24 hours. The head of the pile is finally trimmed to the correct level. Shear transfer cleats or lugs may need to be welded to the head of the pile to ensure transfer of load from the pile cap into the pile. VARIATIONS IN INSTALLATION TECHNIQUE Pre-drilling and Placing In the case of soldier piles the alternative to driving is to auger a hole and drop the steel H-Pile sections in. This is common practice in all soil profiles which are cohesive and not saturated, thus allowing the augering of the hole to take place without collapse. One of the main advantages of placing the H-Pile sections in pre-drilled holes is the fact that position and verticality tolerances attainable can be very stringent, whereas with driving this is not the case. Noise reduction is also an important factor. Having placed and fixed the H-Pile section in position the annulus around it is filled with concrete or grout. Below the excavation level this concrete is full strength, but above the excavation level a weak concrete or grout is used. The reason for this change is to facilitate the removal of the weak material from in front of the H-pile section on the excavated face thereby achieving a plane face to the excavation. Pre-drilling and Jetting To obtain relief from hard driving and to assist in reducing noise levels, pre-drilling or jetting can be resorted to. POTENTIAL PROBLEM AREAS Noise Pollution The driving impact of steel on steel is unfortunately one of the noisiest operations in piling. Even with a cushioning helmet on top of the H-Pile section there is still a high noise level. Certain measures have been used in Europe to reduce the noise levels and these involve enclosing the leader of the piling rig with light sheet-metal lined with sound absorbing material. This has been reported as being fairly successful. The use of hydraulic or vibratory hammers would also assist in reducing noise levels. The safest solution is to use the Steel H-Piles only on sites where noise pollution is not a problem. 95 Pile Heave Because there is limited displacement when driving steel H-Piles there is a very low risk of pile heave. If heave is measured then the piles should be re-driven. Vibration (During Driving) The vibration caused by driving can be a problem and it is very difficult to forecast beforehand how severe the vibration will be. Because of the limited displacement associated with steel H-Piles the risk is even lower than with a precast pile. The general approach from a contracting point of view is to take the risk which is relatively low and sort the problem out in the unlikely event of one arising. Due cognizance should be taken however of the condition of surrounding buildings and whether these buildings are occupied by businesses whose operations are susceptible to vibration. Bent Pile Shafts Because of the elastic properties of steel it has been observed that the shafts of steel H-Piles can be deflected and bent during driving, particularly if there are obstructions in the ground. Unfortunately there is no easy way in which to monitor whether bending has taken place and if so, to what degree, other than if the toe of the pile emerges from the ground surface! A pile load test programme will indicate whether the performance of the piles is acceptable, but this is an expensive exercise. A study of the founding levels and driving records may highlight discrepancies and suggest a limited number of piles for load testing. 96 7.5 AUGER PILES This pile type is very common in southern Africa as it is ideally suited to the partially saturated cohesive and residual soils found in large inland areas of this part of the continent. A number of the large industrial projects in South Africa such as the Sasol oil-from-coal plants, Eskom power stations and Iscor steelworks are founded on this pile type. There is a wide range of pile sizes available and thus the pile type is suited to both large and small structures, as well as large and small contracts. .. .. .. . . . . POSITIVE FEATURES In conditions suited to the use of the auger pile it provides an economical solution The system also provides an economical solution for heaving soil profiles Noise levels are low and limited to engine noise from the piling rig There is no vibration associated with auger piles There is a considerable range of pile sizes from 300 to 2000 mm diameter Auger piles can be installed to depths in excess of 50 metres using the latest hydraulic piling rigs The system can accommodate boulder layers with some limitations OTHER CONSIDERATIONS There is a risk that the side-walls of the pile excavations may collapse resulting in the use of costly temporary casings. This risk can be eliminated in most instances by carrying out an adequate geotechnical investigation using an auger rig to drill trial holes Below the water-table there is a risk of excessive water ingress into the pile excavation which can seriously inhibit progress, and can lead to a change in pile type The presence of boulders can be problematic for smaller diameter piles (less than 750 mm) NOTE: The ‘other considerations’ can and should be checked as part of the geotechnical investigation. This is normally done and these risks are thus largely eliminated. PILE DETAILS The range of auger piles sizes is more extensive than with any other pile type. Auger piles can be as small as 300 mm and as large as 2000 mm in diameter. The following sizes are the most common: 300, 350, 400, 450, 500, 600, 750, 900, 1050, 1200, 1350, 1500, 1650, 1800 and 2000 mm in diameter. Any other size within this range is possible, but a special flight may have to be manufactured for any non-standard size. The safe working loads of auger piles founded on competent material can be calculated using a shaft stress of 3 MPa for pile diameters less than 600 mm and 6 MPa for pile diameters of 600 mm and greater. The reason for the difference in the shaft stress is the fact that with piles of 600 mm diameter and greater, the base of the pile can be cleaned out carefully by hand and thus one can be sure that there is sound endbearing. Smaller sizes are normally cleaned out using a cleaning bucket but this is not 97 successful as the cleaning bucket will always leave a certain amount of loose material on the base of the pile. Because of this, the end-bearing is generally ignored in the calculation of the pile's capacity for diameters smaller than 600 mm. Hence the recommended lower design figure for the shaft stress for the smaller diameter piles. Piles that are socketed into rock can be designed for higher loads due to the increased capacity of the socket friction as well as end-bearing. The shaft stress for socketed piles can be increased to a maximum of 8 MPa provided the sockets are cleaned out and inspected to ensure the competency of the rock at founding level. PILE DETAILS Pile Diameter (mm) Working Load (kN) Minimum Pile Spacing (mm) Typical Main Bar Reinforcing 300 200 - 550 600 6 x 10 mm 350 300 - 750 700 5 x 12 mm 400 375 - 1000 800 6 x 12 mm 450 475 - 1250 900 5 x 16 mm 500 600 - 1500 1000 6 x 16 mm 600 1600 - 2250 1200 6 x 20 mm 750 2500 - 3500 1500 6 x 25 mm 900 3750 - 5000 1800 8 x 25 mm 1050 4750 - 7000 2100 10 x 25 mm 1200 6750 - 9000 2400 14 x 25 mm 1350 8500 - 11500 2700 12 x 32 mm 1500 10500 - 14000 3000 14 x 32 mm 1650 13000 - 17000 3300 16 x 32 mm 1800 15000 - 20000 3600 20 x 32 mm 2000 18500 - 25000 4000 24 x 32 mm INSTALLATION TECHNIQUE The drilling section of the auger rig consists of an engine powering a winch and an auger drive-head. The latter drives the drill stem which is known as a kelly, the kelly is in turn attached to the drill bit or auger flight. The rig has both a drilling speed and a spin-off speed, the latter being used to spin the spoil off the flight. The installation of the pile thus commences with the drilling of the hole. The auger rig is positioned over the pile position and the mast is checked for verticality or rake. Drilling is started by rotating the auger flight and allowing the flight to penetrate 98 into the soil. Once the auger flight is loaded with soil, rotation is stopped and the winch is used to withdraw the flight from the hole. Once clear of the hole the flight is spun and the spoil gets flung off clear of the hole. The flight is then lowered down the hole and the process is repeated. The excavated spoil is continually removed from around the hole during the drilling operation. Different types of drilling tools can be fitted to the kelly. These tools have been developed for penetrating the various types of soil strata that can be encountered. Tools for the penetration of rock are also available. The pile hole is drilled to the required depth which is often onto rock. With pile diameters 600 mm and greater the auger rig is then moved off the pile position and a small tripod rig with a safety winch is erected over the hole. A man is then lowered down the hole by means of the safety winch and he sets about cleaning the remaining loose soil from the bottom of the hole. This material is removed using the winch and a small spoil bucket. This operation is carried out in accordance with the SAICE Code of Practice for the Safety of Persons Working in Small Diameter Shafts and Test Pits for Civil Engineering Purposes (2007). With pile sizes less than 600 mm in diameter the holes cannot be cleaned out by hand. In these instances a cleaning bucket is used to remove as much of the loose material as possible after which anything remaining is compacted by a plate fixed to the end of the kelly bar. Once the cleaning operations has been completed, the hole is checked by a supervisor. The steel reinforcing cage fitted with cover spacers is then lowered down the hole. The shaft of the pile is concreted using a self-compacting concrete mix which is discharged into the hole via a short funnel. The purpose of the funnel is to ensure that the concrete flows down the centre of the reinforcing cage and not onto the cage as this leads to segregation as well as poor compaction. After the shaft has been concreted a poker vibrator is used to assist the compaction of the top three metres where there is insufficient head for self-compaction. Franki Africa owns a number of large auger rigs. Traditionally these rigs were mainly truck mounted, but are being replaced by the latest hydraulic crawler mounted rigs. These new rigs have significantly enhanced capability for both penetration and depth. Rig Type Max. Diam (mm) Max. Depth (metres) Torque (tonne metres) Rig Mass (kg) Soilmec RTAH 1500 32 12 30000 Soilmec R312 1500 38.5 12 35000 Bauer BG15 1500 40.7 15 47500 Bauer MGB25 2000 65 48* 55000 Soilmec R825 2500 77 25 90000 * with torque enhancement Photographs of auger rigs and auger pile methods are shown in PLATES 7.5.1 to 7.5.6. 99 PLATE 7.5.1 Modern Soilmec Hydraulic Auger Rig PLATE 7.5.2 Typical Auger Flight PLATE 7.5.3 Spinning Soil off the Auger Flight 100 VARIATIONS IN INSTALLATION TECHNIQUE Underreams An underream is an enlargement of the base of the pile. In certain circumstances it may be considered desirable to increase the pile's end-bearing area by forming an underream. With the auger system there are mechanical tools that can be attached to the kelly which will excavate the underream. This can only be achieved in materials that are not too dense or stiff, especially on the larger diameters. In the harder materials the underreams can be formed manually with the assistance of air tools. Underreams must be cleaned out by hand prior to concreting. By implication therefore, underreams are only used for piles with a diameter of 600 mm or greater. The alternative to an underream is a rock-socket, as this can be drilled more easily and quickly, and therefore the tendency is to use sockets as opposed to underreams. Piles to Resist Ground Heave There are large areas of southern Africa where the soils are described as active. The volume of active soils increases with the increase in moisture content and decreases with the decrease in moisture content. Increase in soil volume results in the ground surface rising, this is referred to as ground heave. When soil heaves in an area which has been piled there is a friction transfer between the soil and the pile which imparts an uplift force on the pile shaft. Where this force exceeds the axial compression load on the pile, the pile shaft will be in tension. If the tension force is not high, this can be resisted by the shaft reinforcement and the anchorage provided by the length of pile below the active zone. In circumstances where potentially very large tension forces can occur, a method to reduce the friction transfer to a level that can be resisted by the shaft reinforcement has been developed. This method involves the creation of an annulus around the shaft of the pile into which a low strength material is placed. The theory is that the low strength material is only capable of transferring low friction forces and thus the total friction transfer will be similarly reduced. The annulus is created by drilling a pile excavation which is nominally 200 mm larger in diameter than the required shaft diameter. A thin-walled metal liner of the required shaft diameter is then placed in the excavation thereby creating an annulus. This is then filled with a suitable low strength material such as vermiculite or polystyrene aggregate. The annulus is only created over the depth which is considered to be active. The remainder of the pile below the active zone is constructed in the normal way as this is designed to provide the anchorage against any tension force in the pile shaft which may still develop. See SECTION 18.0: PROBLEM SOILS AND THEIR FOUNDATION SOLUTIONS for more information on piled foundations in heaving soils. 101 POTENTIAL PROBLEMS Collapse of Side-walls If there is an area of the site where side-wall collapse occurs then the pile holes need to be temporarily cased through the collapse zone. To be able to do this and still maintain the correct size pile, drilling down to the top of the collapse zone has to be undertaken using a flight about 100 mm larger in diameter than the pile size. A steel casing which has a slightly smaller diameter is then lowered into the hole. A vibratory hammer is clamped to the top of the casing and it is driven down through the collapse zone. Using the size flight for the pile diameter the spoil inside the casing is drilled out and the remainder of the pile hole drilled. The pile is cleaned out and concreted in the manner described above, the concrete filling the temporary casing to a level above the cut-off level. The casing is finally extracted using the vibratory hammer and the concrete flows out of the casing so as to fill the larger diameter. As a number of additional plant items are required for handling the temporary casings their use increases the cost of the operation quite significantly and to an extent that may render the solution uneconomical when compared to other possible pile types. If only a small section of the site has the problem then the overall economics will still favour the Auger pile. This may not be the case if the whole site has a collapse problem. Water Ingress The occurrence of ingress of groundwater into a pile excavation is fairly common, particularly in the soft rock at the toe of the pile. This water seeps into the pile excavation, collecting at the bottom of the hole, and a decision has to be made as to whether the pile can be concreted successfully in these conditions. The answer depends on the rate of inflow. If the flow is limited, then a pump lowered to the bottom of the hole during the cleaning operation will remove the water, allowing cleaning to proceed. When this has been achieved, a large quantity of concrete is discharged rapidly into the hole via a concreting shute as soon as the pump is removed. The steel reinforcing cage is then lowered into the hole and bedded into the concrete. The remainder of the shaft is then concreted. During the short time it takes to remove the pump and discharge the first concrete a quantity of water will enter the hole. This will not affect the performance of the pile provided that the concrete that is discharged into the pile hole is concentrated in the centre and the concrete flow rate is high. This will limit the amount of water that is mixed with the concrete at the base of the pile. Some water will of course get mixed with the concrete during the concreting operation but this can be compensated for by using additional cement in the mix. 102 If the rate of inflow is too fast, then it is better to cast the pile under water. For this to be achieved, it will be necessary to temporary case the hole for the full depth, prior to concreting, to prevent material that may collapse off the side-walls, due to the presence of the water, from collecting at the bottom of the hole. The concreting operation is then carried out using a tremie pipe which is standard procedure in piling, and the casing is extracted. Another alternative in this situation would be to drill the piles under a head of bentonite as covered under SECTION 7.6: UNDERSLURRY PILES. This would eliminate the problem but would increase the cost. Many of the problems associated with collapse and ingress of groundwater can be overcome by modern piling rigs which can install temporary casings during the drilling operation. The diameter and depth to which a casing can be drilled into the ground is a function of the machine size and drive torque, with cased depths up to 20 metres achievable with larger machines. Boulders The problems presented by boulders depend on factors such as the size, hardness and concentration of the boulders, the type of matrix, and whether there is any water present in the boulder layer. If the size of the boulders is less than one third the pile diameter, the concentration is plus/minus two or three boulders per metre of depth and the matrix is soft or loose, the auger rig will drill through the layer albeit with a slower penetration rate. If the concentration increases to tightly packed, the penetration rate will decrease even further. If the matrix also changes to very stiff, or very dense, then penetration will be very slow, if at all possible. It will not be possible to drill out boulders larger than one half the pile diameter. In these cases personnel are lowered down the pile excavation and a sling is attached to the boulder by drilling a hole through it. A crane is then used to lift the boulder out of the excavation. If a large boulder is encountered then it will be necessary to split the boulder into smaller pieces before they can be removed. Piling in these conditions can be very slow and thus expensive, but the problem can be solved. Auger piling in boulders below the water-table can still be feasible provided the ingress of water can be handled by pumps to allow access to the boulder by personnel. If the ingress is too fast for this to happen then there is no alternative but to change to the Oscillator Piling System, or the Rotapile for smaller pile diameters, which are capable of handling this type of problem. For details of these systems see SECTION 7.10 and 7.11. 103 PLATE 7.5.4 Installing Temporary Casing PLATE 7.5.5 Hole Cleaning Operation PLATE 7.5.6 Exposed Auger Piles 104 7.6 UNDERSLURRY PILES This system is commonly used for heavily loaded structures where the soil profile is saturated over all or part of the profile resulting in unstable side-wall conditions for conventional auger piles. The term underslurry indicates that the pile is excavated under a head of bentonite slurry, which prevents collapse of the pile excavation. Both auger and grab excavation methods can be used, resulting in circular and rectangular cross-sections. These two methods are referred to as Auger Underslurry and Barettes respectively. .. .. .. .. .. . POSITIVE FEATURES Economical pile solution for heavy structures on saturated and unstable soil profiles Noise levels are low and limited to engine noise from the equipment There is no vibration associated with Underslurry piles A good range of pile sizes, from 600 to 1500 mm in diameter, is available Underslurry piles can be installed to depths in excess of 50 metres Barette type piles can be shaped OTHER CONSIDERATIONS High establishment costs Large site area required Suited to heavily loaded structures only Only vertical piles can be constructed The slurry contaminated spoil must be disposed of at an approved waste site PILE DETAILS Auger Underslurry Pile Diameter (mm) Working Load (kN) Minimum Pile Spacing (mm) Typical Main Bar Reinforcing 600 1700 - 2250 1500 6 x 20 mm 800 3000 - 4000 2000 6 x 25 mm 900 3800 - 5000 2250 8 x 25 mm 1000 4750 - 6250 2500 10 x 25 mm 1100 5750 - 7500 2750 12 x 25 mm 1200 6750 - 9000 3125 14 x 25 mm 1350 8500 - 11500 3375 12 x 32 mm 1500 10500 - 14000 3750 14 x 32 mm The above working loads are calculated using a shaft stress of 6.0 MPa. If the piles are socketed into bedrock, or there is considerable shaft friction, then the loads on the piles can be increased, but should not exceed 8 MPa on the pile shaft concrete. The pile's tension capacity is dependent on friction and the amount of reinforcement. The maximum depth to which auger underslurry piles can be installed is 50 metres and the piles should be vertical as it is not possible to control the drilling on the rake. 105 PILE DETAILS (continued) Barettes Pile Width (mm) Pile Length (mm) Shaft Area (m2) Working Load (kN) Min. Pile Spacing (mm) Typical Main Bar Reinforcing 600 2800 1.68 7500 - 13500 2250 14 x 25 mm 800 2800 2.24 10000 - 18000 2500 18 x 25 mm 1000 2800 2.80 12500 - 22400 2750 14 x 32 mm 1200 2800 3.36 15000 - 27000 3000 18 x 32 mm The above working loads are calculated using a shaft stress of approximately 4.5 MPa. Because of the grab's limited ability to penetrate materials of very soft rock consistency it is recommended that this be a maximum unless a socket is formed where the pile shaft stress can be increased to 8.0 MPa. The pile's tension capacity is dependent on friction and the amount of shaft reinforcement. The maximum depth to which Barettes can be installed is really unlimited, but a practical limit of 60 metres is suggested. Barettes cannot be installed on a rake. BENTONITE SLURRY Bentonite slurry is a mixture of bentonite and water. Bentonite is a form of Montmorillonite clay. It is mined in various parts of the world including southern Africa and is processed into a powder which is shipped in 40 kg sacks or bulk. Bentonite slurry is a mixture of approximately 5 percent of bentonite powder by mass with water. It has to be mixed in a specially designed mixer as the powder tends to float on water like talcum powder. The mixed slurry has a Specific Gravity of about 1.04. A bentonite slurry has various unusual qualities, the main one being that it is a thixotropic liquid. This means that the clay particles do not settle out of suspension with time. Sand and other heavy particles mixed with the bentonite slurry during the excavation process are also held in suspension and do not settle to the bottom of the excavation. The bentonite slurry when maintained at a positive head of at least 1.5 metres relative to the water-table builds up a layer of clay particles on the walls of the pile excavation. This layer which is normally 2 to 3 mm thick is known as the cake and is relatively impermeable. The cake coupled with the positive head is sufficient to prevent collapse of the side-walls of the pile excavation which is the main advantage of using bentonite slurry. Bentonite slurry is mixed on site in a mixing plant and stored in large tanks. It is fed to the pile excavation during the drilling operation and is pumped back to the storage tanks during the concreting operation. Sand becomes mixed with the slurry during the drilling and the sand content can be as high as 30 percent by weight. Prior to concreting this should be reduced to less than 3 percent. De-sanding is achieved by means of cyclones and vibrating sieves. 106 During the pile installation process the viscosity and the pH of the bentonite slurry may be altered by the drilling, the cement in the concrete and the groundwater. These properties are routinely checked during the piling operations and various chemicals are added to maintain the bentonite slurry within the specification. If the pH of the slurry rises above 12.5 the slurry will flocculate and lose its properties. This contaminated slurry has to be disposed of. INSTALLATION TECHNIQUE Auger Underslurry The auger rig is set up on the pile position and the mast is checked for verticality. A shallow excavation is formed after which a short length of steel casing is screwed into the ground by the auger machine. This casing is known as a starter casing and it is installed to protect the top of the pile excavation from collapse. The starter casing is filled with bentonite slurry after which the drilling of the pile excavation continues once the mast has been checked for position and verticality. During the drilling operation the excavation is kept full of bentonite slurry at all times. The drilling bucket and kelly displace bentonite slurry when they are lowered into the excavation, to overcome this, a surge tank is created around the head of the pile to contain the rise in level. On reaching the founding level the excavation is cleaned out as best as possible with a cleaning bucket attached to the kelly. Further cleaning of the base of the excavation is achieved using an airlift pump which suctions the full area of the base. This airlifting operation is also used to change the bentonite slurry over, with the contaminated slurry being pumped to the storage tanks for processing while clean bentonite is fed into the pile excavation. It is preferable to have clean bentonite in the excavation during concreting. On completion of the cleaning operation the steel reinforcing cage is lowered into the pile excavation. Wide concrete cover spacers are used to prevent the spacers penetrating into the soft side-walls of the excavation. The pile shaft is then concreted using standard tremie techniques. Measurements of the level of the concrete in the excavation are taken regularly to check that the tremie is immersed in the concrete at all times. The pile should be concreted a minimum of one metre above the cut-off level to achieve some degree of compaction and sound concrete at the head of the pile. The installation sequence is shown in diagrammatic form in FIGURE 7.6.1. Some aspects of Auger Underslurry piling operations are shown in PLATES 7.6.1 to 7.6.5. 107 FIGURE 7.6.1 108 B C D E F A. A short length of starter casing is drilled into the ground on the pile position. The starter casing is filled with bentonite slurry and drilling commences. B. The drilling continues until the founding level is reached, the level of the slurry in the excavation being maintained at all times. C. An airlift pump is lowered into the excavation and is used to clean the base of the pile and to pump the contaminated slurry to storage tanks for treatment. D. The reinforcing cage is placed in position and the tremie pipe is inserted into the pile shaft. E. The concreting of the pile shaft is carried out in one continuous operation and displaced bentonite slurry is pumped to storage tanks. F. The completed pile after the starter casing has been extracted. A AUGER UNDERSLURRY PILE INSTALLATION SEQUENCE PLATE 7.6.1 Drilling under Bentonite Slurry PLATE 7.6.2 Bentonite Slurry Mixing, Storage and De-sanding Equipment 109 Barettes A Barette is constructed in a very similar manner to that described for the Underslurry pile, except that the excavation is carried out using a grab and not an auger rig. The grab is normally of the cable or hydraulic type and excavates a rectangular hole. The width of the hole can be varied by changing the jaws on the grab. The depth of excavation is virtually unlimited. To achieve penetration into material of rock consistency, chisels can be used to break up the rock for removal by the grab. This is a costly exercise however, and normally the piles are founded on the rock and not socketed into rock. Excavations using a grab can be carried out in multiple passes in the same manner as that for a diaphragm wall. In this way the shape of the excavation can be extended and altered. These larger units are referred to as load-bearing panels and can be arranged to form part of a diaphragm wall as well as performing a load-bearing function. The grab for excavating Barettes is shown in PLATE 7.6.3. VARIATIONS IN INSTALLATION TECHNIQUE Integrity Testing As the whole pile installation is carried out under bentonite slurry some form of integrity testing is considered desirable. Rotary core drilling can be used to check the contact between the concrete and the rock. The integrity of the pile shaft itself can be checked with either nuclear or sonic methods as described in SECTION 9.0: PILE LOAD AND INTEGRITY TESTING. To enable these tests to be carried out, three or four small diameter steel tubes are cast into the pile. These tubes are normally fixed to the reinforcing cage and lowered with the cage into the pile bore. Composite Steel Pile Shafts In Europe a modern building technique is sometimes used to accelerate the construction of multi-storey buildings with basements, as well as to reduce movements associated with basement excavation. In essence this technique calls for the construction of the basement of the building from the ground floor downwards while at the same time constructing the building itself from the ground floor upwards. The saving in time with this form of construction is considerable. If the building is founded on piles then a variation of the Underslurry system is used to make this modern building system possible. This variation involves the use of large steel column sections as pile shafts over the depth of the basement. The advantage gained by this variation is that brackets, cleats and other connections can be welded to the columns to allow the floor-beams to be joined to the columns. Each floor level is completed as excavation proceeds downwards to form the bracing between the surrounding basement walls. 110 These steel columns are cast into the concrete pile shaft with suitable shear connectors. Special adjustable guiding devices have been developed for the accurate positioning of the steel column sections in the excavation. This is necessary so as to ensure that the pre-fabricated steelwork fits at each floor level. This system has not been used in southern Africa to date, but the expertise is available within Franki Africa should there be a suitable project where the savings in construction time are meaningful to the client. POTENTIAL PROBLEMS Collapse of the Pile Shaft Excavation There is a risk of collapse of the side-walls of the excavation but this can be reduced to a minimum by observing correct procedures and ensuring that the bentonite slurry itself is within the specification and that the differential head is not less than 1.5 metres at any stage. A minor localized collapse can occur occasionally but this should not be detrimental to the pile. A major collapse of a pile shaft excavation is serious and the quick fix is to backfill the excavation as quickly as possible. The cause of the collapse should be determined before any further piling work is carried out. Additional care should be exercised with long panel excavations, excavations in extremely soft clays and silts, and excavations carried out in close proximity to existing loaded foundations. Delays Before Concreting Ideally a pile should be concreted the same day as its excavation is completed. If there are delays before the concrete is placed the bentonite cake tends to increase in thickness and this has been found to be detrimental to the pile's friction capacity. If a delay is unavoidable then the pile excavation should be reamed out again, using the excavation tools, prior to concreting the pile. Tremie Concreting This is an operation which needs a large amount of experience on the part of the contractor for it to run smoothly and produce a sound pile shaft concrete. Unfortunately even the best trained crews can run into problems, especially if the concrete itself does not have the required workability. The most likely problem is a blockage in the tremie. This requires the tremie to be removed and cleaned out. If only a small amount of concrete has been discharged into the pile the best plan of action is to remove the steel reinforcing cage and permanent liner and drill out the wet concrete. Once cleaned out the steel cage can be installed and the pile concreted. 111 If a large amount of concrete has already been placed when the blockage occurs the tremie should be removed and cleaned out. A watertight end-cap should be placed on the toe of the tremie and the latter lowered until it penetrates the wet concrete by at least two metres. In this operation, as with all tremie operations, it is vital that the joints in the tremie pipe are absolutely watertight. This can be checked by shining reflected sunlight down the tremie pipe using a mirror. The tremie is then filled with concrete and raised slightly at which time the end-cap comes off due to the weight of the concrete. Thereafter the concreting operation can proceed as normal. Restarting a tremie operation is a tricky process. For this reason the end result should be checked using one of the integrity testing methods. The alternative is to abandon the pile and install a replacement pile. Concrete Bleeding For the concrete to flow through the tremie it has to have a slump of about 200 mm. To obtain this slump without having excess water in the mix is extremely difficult if not impossible with the available aggregates and admixtures. On deep piles the excess water in the concrete bleeds out and makes its way up through the concrete to the surface. The channel or run it forms in so doing is normally found near the centre of the pile where the tremie tends to form a zone of excess mortar. The concrete in this zone often shows signs of excess water and may have had some of the cement washed out of it. This is a problem which cannot be completely eliminated with deep piles cast using a tremie but certain measures can be taken to ensure it is reduced to a minimum. By adopting suitable techniques the percentage area of the pile shaft affected in this manner can be kept low so that the overall load-bearing capacity of the pile is not affected. Before commencing the contract, the gradings of the available aggregates should be determined and trial mixes designed and tested to arrive at the optimum mix design. During the contract there should be adequate site control to ensure the mix is stringently adhered to and in particular that no additional water is added to the mixer trucks while they are on site. 112 PLATE 7.6.3 Barette Grab Excavation PLATE 7.6.4 Testing of Bentonite Slurry PLATE 7.6.5 Typical Underslurry Contract 113 7.7 CONTINUOUS FLIGHT AUGER (CFA) PILES This piling system is a fast and economical one which has no vibration and limited noise levels associated with it. Unfortunately it also has some limiting considerations and it is problems associated with these that have reduced the popularity of the system. .. .. . . .. . . POSITIVE FEATURES High productions levels are attainable in suitable soil conditions The system is economical in suitable soil profiles Noise levels are low and limited to the engine noise of the piling rigs There is no vibration associated with CFA piles OTHER CONSIDERATIONS The manufacture of the pile is largely in the hands of the operator and there needs to be adequate monitoring of concrete/grout flow and pressure as well as the extraction rate of the flight so as to be completely satisfied that the piles are made correctly In sandy soils below the water-table there is a reduction in soil strength in the immediate vicinity of the pile due to the drilling operation, as a result of this the load/deflection performance of a CFA pile is inferior to that of a driven pile The steel reinforcing cage is inserted into the wet concrete/grout after the pile is cast Soil falling off the flight can contaminate the concrete/grout of the pile shaft CFA piles have to be cast to ground level. This results in waste of concrete and excess trimming if there are deep cut-off levels There is very limited indication of soil strength during the drilling operation PILE DETAILS Pile Diameter (mm) Working Load (kN) Minimum Pile Spacing (mm) Typical Main Bar Reinforcing 300 300 - 425 750 6 x 10 mm 350 375 - 550 875 5 x 12 mm 400 500 - 750 1000 6 x 12 mm 450 650 - 950 1125 5 x 16 mm 500 800 - 1200 1250 6 x 16 mm 600 1150 - 1700 1500 6 x 20 mm 750 1750 - 2650 1875 6 x 25 mm The maximum depth of CFA piles is 25 metres at present and the recommended maximum rake 1:10. The above pile working loads are calculated using a shaft stress of 4 - 6 MPa. Where piles are founded on or socketed into rock, consideration can be given to using a shaft stress in excess of 6 MPa. The pile's tension capacity must be determined from a calculation of friction. 114 INSTALLATION TECHNIQUE The piling rig used to install CFA piles has the means to rotate, lift and lower a hollowstemmed continuous flight auger which is the main mechanical feature of the system. The hollow-stem of the auger is blocked off at the toe by means of a suitable plug prior to the flight being lowered onto the pile position. The leader of the piling rig is adjusted for position and verticality, or rake. The flight is rotated and at the same time allowed to penetrate into the soil. The penetration rate has to be controlled by the rig operator so as not to stall the auger head. On the other hand too slow a penetration rate will result in excess soil being removed with resulting decompression of the soil profile which is undesirable. During the drilling operation the only indication of the strength of the soil profile is gained from a measure of the torque. This is not a sensitive measurement and only gives an overall indication which is insufficient for pile depth determination. CFA piles are thus normally installed to a pre-determined depth. When this depth has been attained the flight is lifted slightly and the pumping of the concrete/grout commenced. The pressure of the concrete/grout blows the plug out the end of the flight and concrete/grout flows out to form the shaft. The flight is thereafter extracted at a controlled rate to match that of the flow of the concrete/ grout. If extraction is too fast, a necked pile shaft can be formed. If it is too slow, then the flight tends to become stuck in the ground. Rotation of the flight during the concreting/grouting operation is normally kept to a minimum. It may be necessary to rotate the flight during the initial extraction stages but thereafter rotation is ceased. CFA piles are always cast to ground level so as to ensure an adequate head for compaction and for maintaining the correct pile diameter. Once the concreting/grouting operation has been completed the rig moves away from the pile position. The head of the pile is cleaned up and any latence and soil are removed to expose sound concrete/grout. The reinforcing cage is then lifted up and placed into the concrete/grout. Vibrators attached to the cage are sometimes used to assist in penetrating the cage into concrete. FIGURE 7.7.1 shows the CFA piling sequence in diagrammatic form. In some instances it may be desirable to use a sand/cement grout instead of concrete and in fact some contractors use grout in preference to concrete as it is easier to pump. Provided the mix attains the desired strength both are equally acceptable. When grout is used the water in the grout slowly migrates into the surrounding soil if it is sandy and this results in a drop in the level of the grout at the head of the pile. Additional grout has to be added so as to maintain the cast level of the pile. This occurs to a much lesser degree with concrete. PLATE 7.7.1 shows CFA piling operations in progress. 115 FIGURE 7.7.1 116 B C D A. The hollow-stemmed Continuous Flight Auger is drilled into the ground by means of the drive-head. Penetration of the auger into the ground is slower than the combination of pitch and rotation, so a certain amount of soil is ejected at the surface. This decompression is necessary to allow penetration of the auger. B. The auger is drilled down to the founding level after which the concrete / grout pump is connected to the hollow-stemmed flight by means of high pressure hoses. C. The concrete/grout is pumped down the hollow-stemmed flight as the latter is gradually withdrawn. A high level of control is necessary to ensure that the rate of extraction matches the rate of flow of the concrete/grout. If the extraction rate is too fast, a necked pile shaft will be formed. If it is too slow, there will be a pressure build-up which forces grout up the sides of the auger, and can lead to the flight becoming stuck. D. The steel reinforcing cage is lowered into the wet concrete/grout in the pile shaft until it is at the correct level. This completes the installation sequence. A CONTINUOUS FLIGHT AUGER (CFA) PILE INSTALLATION SEQUENCE VARIATIONS IN INSTALLATION TECHNIQUE Intermediate Bentonite Stage After drilling is complete the auger is withdrawn as bentonite slurry is pumped into the excavation through the hollow-stemmed auger. The slurry prevents the hole from collapsing. A full length reinforcing cage can then be placed in position and the shaft concreted or grouted using a tremie. The advantage gained by using this variation is the fact that a heavy reinforcing cage can be placed right to the toe of the pile which may not be possible with the conventional technique. POTENTIAL PROBLEM AREAS Necked Pile Shafts Most of the problems experienced with this pile type are associated with the incorrect rate of extraction of the flight which results in necked pile shafts. It is important that adequate monitoring of the flow rate of the concrete / grout and the matching extraction rate of the flight is carried out to ensure a sound pile shaft. Modern piling rigs are equipped with an on-board recording device which provides a detailed installation record. These systems are mandatory in many parts of the world and should be adopted in our region. Reduction in Pile Capacity As mentioned above the drilling of the auger into the ground results in decompression of the soil. Overdrilling of the pile must be avoided, or kept to a minimum, as this causes further decompression. The power of the auger drive-head also has a bearing on the degree of decompression, with the more powerful heads giving better results, as they have the power to penetrate the auger into the ground with limited decompression. Because of the decompression the ultimate unit friction transfer for CFA piles is normally about 60 to 75 percent of that for driven displacement piles. The end-bearing component of CFA piles founded in sandy soils below the watertable is significantly affected by the drilling operation. For this reason the end-bearing is often ignored in these soil conditions and the pile is designed as a friction pile. Load tests have shown that in fact there is an end-bearing component, but it requires a pile toe movement of about five percent of the pile diameter for it to start to perform, so in effect the pile acts as if it is a friction pile in the range of acceptable pile-head deflections. CFA piles founded in or on materials which do not decompress, such as stiff cohesive soils and rock, have the full end-bearing component, and may be designed as such. Obstructions such as Boulders The auger should penetrate boulders which are less than one third of the pile diameter and are not tightly packed. Larger boulders present a problem and the only solution is to move the pile position. The CFA system is regarded as being sensitive to obstructions and this should be borne in mind when considering its use. 117 PLATE 7.7.1 CFA Rig Drilling 118 7.8 FULL DISPLACEMENT SCREWPILES This piling system has recently been introduced by Franki into the southern African region. Its well established track record in other parts of the world has shown it to be a fast and economical piling methodology with no vibrations and limited noise levels. It has overcome many of the limiting considerations of the CFA pile and exhibits equivalent pile performance to a driven pile. .. . .. . .. . . . POSITIVE FEATURES High productions levels are attainable in suitable soil conditions The system provides an economical solution in suitable soil profiles The soil displacement technology delivers good pile performance at significantly shallower depths to CFA and bored piles Noise levels are low and limited to the engine noise of the piling rigs The system is vibrationless The pile shaft can be reinforced over its full length using a variation to the normal installation methodology Pile load capacity can be ensured on each pile by monitoring the installation energy No soil removal is required with the soil displacement technology OTHER CONSIDERATIONS The penetration of boulder layers or the presence of obstructions can preclude the use of this pile type The piles must be concreted over the full shaft length resulting in excessive waste and pile shaft trimming if deep cut-off levels are required The excessive drilling torque required to penetrate substantially thick dense sand or stiff clay layers must be carefully assessed for heavily loaded foundations PILE DETAILS Pile Diameter (mm) Working Load (kN) Minimum Pile Spacing (mm) Typical Main Bar Reinforcing 300 400 - 550 750 6 x 10 mm 400 750 - 1000 1000 6 x 12 mm 500 1150 - 1550 1250 6 x 16 mm 600 1600 - 2200 1500 6 x 20 mm 700 2300 - 3000 1750 8 x 20 mm The maximum depth to which the displacement screwpile can be installed to is dependent on the size of the piling rig, rotary drive torque, and the soil profile. A maximum depth of 30 metres is achievable with Franki’s Bauer MBG25 piling rigs. The pile shaft working loads noted above are limited by a maximum pile shaft stress of 8 MPa with regionally available aggregates. The piles uplift capacity is governed by the pile shaft length, but will generally be 30% more than a bored pile, in comparable soil and similar pile geometry. 119 INSTALLATION TECHNIQUE The full displacement screwpile is a refinement of the well established Continuous Flight Auger (CFA) pile (soil replacement). The installation and concreting methodology is similar to the CFA with the refinement concentrated on the design of the auger flight, to overcome the negative effects of CFA pile installation. The piling rig used to install displacement screwpiles is similar to a CFA pile but has a high torque rotation head, up to 50 tonne metres, and a crowd and removal capability for the hollow-stem screwpile flight. The displacement auger tool is carefully designed to ensure full displacement during the downward pushing and rotation of the tool. The tool comprises a lower tapered augered portion of the flight, a central displacement section, and an upper auger section with reverse flighting. The displacement screwpile auger tool is shown on PLATE 7.8.1. The hollow-stem of the flight is blocked off at the toe by means of a suitable plug prior to the flight being lowered onto the pile position. The piling rig mast is adjusted for verticality or rake. The flight is rotated and at the same time pushed to penetrate the soil and the rate of penetration, torque and crowd are fully recorded on the rig’s data capturing system. The installation energy is calibrated against trial pile test data to ensure satisfactory pile load capacity during installation. When the required installation energy and penetration depth has been achieved, grout / concrete is pumped through the hollow-stem, and the flight removed during the pumping process. The rate of the flight withdrawal is carefully monitored against the volume of grout / concrete pumped thus ensuring satisfactory pile shaft integrity. Modern piling rigs provide this essential record which can be made available to all relevant parties. The flight is rotated during the concreting / grouting and flight extraction process and the full length of penetrated pile shaft is concreted / grouted. Once the concreting / grouting operation has been completed, the rig moves away from the pile position and the head of the pile is cleaned up. The reinforcement cage is then lowered into the fluid concrete / grout using vertical crowd or vibrators attached to the cage which ensures the required length of reinforced pile shaft is attained. FIGURE 7.8.1 shows the installation sequence of the full displacement screwpile graphically. VARIATIONS IN INSTALLATION TECHNIQUE Where the full pile shaft length must be reinforced and high durability in the form of well controlled cover is required, the pile can be installed using a disposable tip to the auger tool. The reinforcement cage can be placed into the hollow auger stem down to the pile-toe level before commencement of the pile concreting / grouting operations. A self-compacting high slump concrete, or fluid grout, is placed into the empty hollowstem and the auger flight is removed, as outlined, on completion of the concreting / grouting operation. PLATE 7.8.2 shows a full displacement screwpile being installed. 120 FIGURE 7.8.1 121 B C D A. The hollow-stemmed Screwpile Displacement Auger is drilled into the ground by means of a high torque drive-head. Penetration of the auger into the ground is maintained at a pre-defined rate to prevent soil removal and decompression. B. The auger is drilled down to the founding level after which the concrete pump is connected to the hollow-stemmed flight by means of high pressure hoses. C. The concrete is pumped down the hollow-stemmed flight as the latter is gradually withdrawn. A high level of control is necessary to ensure that the rate of extraction matches the rate of flow of the concrete. If the extraction rate is too fast, a necked pile shaft will be formed. If it is too slow, there will be a pressure build-up which forces concrete up the sides of the auger and can lead to the flight becoming stuck. D. The steel reinforcing cage is lowered into the wet concrete in the pile shaft until it is at the correct level. This completes the installation sequence. A FULL DISPLACEMENT SCREWPILE INSTALLATION SEQUENCE PLATE 7.8.1 Displacement Screwpile Auger Tool PLATE 7.8.2 Full Displacement Screwpile Installation 122 POTENTIAL PROBLEMS As this pile type is a refinement of the Continuous Flight Auger Pile (CFA) it has encountered similar installation problems. Problems associated with the difficulty in penetrating boulders or thick stiff / dense layers are similar to the CFA pile, but due to the soil displacement technology even greater difficulties can be encountered in achieving the required penetration depth. For this reason it is imperative that a robust rig with enhanced drilling torque is used. Problems associated with decompression are not encountered if the correct installation sequence and methodology are used. The use of modern piling rigs, which can monitor all the required functions during drilling and concreting, substantially overcome the risks associated with pile shaft integrity. 123 7.9 FORUM BORED PILES The main feature of this system is the relatively small and light piling equipment used to install the pile. This makes the system suited to low headroom conditions and sites with limited or difficult access. Site establishment costs are low so the system is also suited to small contracts. .. .. . .. .. .. POSITIVE FEATURES Site establishment costs are low The piling rig can operate in limited headroom and difficult access conditions The forum bored pile has an enlarged base similar to the Franki piling system The forum bored pile has good tension capacity A rock-socket can be formed as an alternative to the enlarged base OTHER CONSIDERATIONS High cost per metre of pile installed Low production rates System not suited to cohesionless soils below the water-table Only one size of pile available Depth of piles limited to about 12 metres depending on soil conditions Noise and vibration levels, though not high, can be a problem PILE DETAILS Pile Diameter (mm) Nominal Diameter of Pile Shaft (mm) 410 Typical Working Load (kN) 500 - 800 Maximum Depth (metres) * 12 Minimum Pile Spacing (mm) 1000 Maximum Rake 1:6 Typical Main Bar Reinforcing 5 x 16 mm Maximum Main Bar Reinforcing 8 x 25 mm Spiral Reinforcing at 150 mm Pitch 6 mm Nominal Cover to Reinforcing Steel (mm) 40 Maximum Tension Load (kN) 200 OD of Piling Tube (mm) 406 ID of Piling Tube (mm) 366 Typical Hammer Mass (kg) 660 * Depends on the soil conditions. Greater depths are possible. 124 INSTALLATION TECHNIQUE The plant used for the installation of the Forum bored pile consists of a tripod type mast with a winch and engine mounted on a frame fixed to one of the legs of the tripod. The system makes use of a temporary piling tube which is made out of one metre long screw-coupled sections. Various excavation tools are used for removing the spoil from the bore of the tube as well as for driving the tube. An internal drop hammer is provided for forming the enlarged base and a separate hydraulic jacking unit is used for extracting the tubes. Once the rig has been set up correctly the first section of casing is driven. The spoil is removed from this first section after which another section of tube is added by means of the screw-coupling. The tube is again driven and the spoil removed. The cycle is repeated until the founding level is reached. After cleaning out the hole a small charge of zero slump concrete is discharged into the tube. The tube is raised slightly by means of the jacking unit and the concrete is expelled from the tube by blows of the hammer. Further charges of concrete are expelled in a similar manner thus forming an enlarged base of the correct size. Having ensured that all the concrete has been driven out of the tube the steel reinforcing cage is lowered into the tube which is then filled with high slump concrete. Using the jacking unit the tube is extracted with individual sections being removed as the operation continues. The tubes are washed clean ready for use on the next pile. With the tube fully extracted the cut-off level is checked and adjusted. The installation sequence is shown diagrammatically in FIGURE 7.9.1. PLATE 7.9.1 shows a typical Forum Bored Pile operation in progress. PLATE 7.9.2 shows the rig operating in difficult access conditions. VARIATIONS IN INSTALLATION TECHNIQUE Rock-socket The system has the facility to form a rock-socket as an alternative to the enlarged base. Various chisels are used to penetrate the rock which should have an unconfined compression strength of not greater than 10 MPa. Conventional tools are used for removing the spoil and cleaning out the socket. The remaining concreting operation is carried out as detailed above. 125 FIGURE 7.9.1 126 B C D E F A. Driving the temporary casing which is made up out of one metre long screw-coupled sections. B. Coring out the soil from the bore of the casing using the coring tool. C. Forming the enlarged base by driving out semi-dry concrete using the hammer. Hydraulic jacks prevent the casing from penetrating further into the ground. D. The steel reinforcing cage is placed into the empty pile bore. E. The pile tube is filled with high slump concrete and the temporary casing is jacked out with each casing segment being removed. F. The completed pile. A FORUM BORED PILE INSTALLATION SEQUENCE Penetrating Obstructions The same chiseling technique used to form a rock-socket can be used to remove obstructions. Because of the small tube size and limited weight of the chisel there is naturally a limit to the size and hardness of obstructions that can be penetrated. As a guide, obstructions up to one tube diameter, and a hardness of up to medium hard rock, can be penetrated without undue difficulty. Permanently Sleeved Pile Shafts The installation technique can be altered to incorporate a permanent liner for the pile shaft. This is sometimes desirable in aggressive groundwater conditions and the liner can be made out of chemically resistant materials. The liner is placed in position after the formation of the enlarged base. It is sometimes fixed to the reinforcing steel cage and both are inserted as a unit. The liner is filled with high slump concrete after which the tube is then extracted. The annulus between the liner and the piling tube is filled with a sand cement grout. POTENTIAL PROBLEMS Boiling Sand and water are said to ‘boil’ into a pile excavation when they flow in past the toe of the piling tube due to the differential head between the water-table on the outside of the tube and the water-level inside the tube. From this it is clear that boiling only takes place when a water-table is present. Any suction effect caused by the excavating tools will encourage boiling to take place. If boiling takes place during excavation of the pile there is a risk of ground settlement around the pile position. Such settlement can undermine an existing foundation and needs to be avoided. To achieve this the temporary casing must be driven well ahead of the level of excavation inside the casing. Boiling can also seriously hamper the cleaning out of the pile prior to concreting. In sandy soil profiles below the water-table it is important that there is a cohesive layer immediately above the founding level into which the tube can be sealed. The pile can then be cleaned out successfully and the base formed. If such a clay layer does not exist then a driven tube pile may be the better solution. 127 PLATE 7.9.1 Boring in Progress PLATE 7.9.2 Forum Bored Piles in Difficult Access Conditions 128 7.10 OSCILLATOR PILES The main feature of the Oscillator pile is its ability to penetrate through rock and boulder layers and to socket into bedrock. These attributes make it an ideal pile for large river bridges and for marine construction. The pile diameters are relatively large and the range of pile sizes limited. The system is thus only economical when used under heavily loaded structures. A permanent metal liner can be incorporated into the pile and this feature is often specified on river bridges to protect the wet concrete of the pile shaft from flowing groundwater or artesian conditions. .. .. .. . .. . POSITIVE FEATURES Large pile sizes for large pile loads The ability to penetrate substantial boulder layers and other obstructions The ability to form rock-sockets The use of thin-walled permanent liners Noise levels are low and limited to the noise from the equipment There is no vibration associated with Oscillator piles Depths of up to 60 metres are possible OTHER CONSIDERATIONS It is an expensive pile type A large working platform area is required There is a limited range of pile sizes PILE DETAILS The following table applies to unlined Oscillator piles. Pile Diameter 1000 mm 1200 mm 1500 mm 1000 1200 1500 Typical Working Load (kN) 4750 - 6250 6750 - 9000 10000 - 14000 Maximum Depth (metres) 60 60 60 Minimum Pile Spacing (mm) 2500 3000 3750 Maximum Rake 1: 4 1: 4 1: 4 10 x 25 mm 14 x 25 mm 14 x 32 mm Typical Spiral Reinforcing at 150 mm Pitch 8 8 10 Nom. Cover to Reinforcing (mm) 75 75 75 OD of Piling Tube (mm) 1000 1200 1500 ID of Piling Tube (mm) 920 1120 1400 Nom. Shaft Diameter Unlined (mm) Typical Main Bar Reinforcing The above working loads are based on a shaft stress of 6.0 to 8.0 MPa where no permanent liner is installed. The pile's tension capacity depends on the calculated value of friction. The maximum depth is limited to 30 metres, due to problems relating to removal of the casings during concreting. 129 PILE DETAILS (continued) The following table applies to lined Oscillator piles. 900 mm Pile Diameter 1100 mm 1350 mm Nom. Shaft Diameter Lined (mm) 900 1100 1350 Typical Working Load (kN) 6500 9500 14250 Maximum Depth (metres) 60 60 60 Minimum Pile Spacing (mm) 2500 3000 3750 Maximum Rake 1: 4 1: 4 1: 4 10 x 25 mm 14 x 25 mm 14 x 32 mm Typical Spiral Reinforcing at 150 mm Pitch 8 8 10 Nom. Cover to Reinforcing (mm) 75 75 75 OD of Piling Tube (mm) 1000 1200 1500 ID of Piling Tube (mm) 920 1120 1400 Typical Main Bar Reinforcing The above working loads have been calculated using a shaft stress of 10 MPa. The pile's tension capacity depends on the calculated value of friction. Depths of up to 60 metres can be achieved with the use of the permanent liners. INSTALLATION TECHNIQUE The equipment consists of an oscillator which has the ability to clamp the piling casing, to move it rotationally through about 15 to 20 degrees and to lower and raise it, all movements being achieved by the use of hydraulic rams. A crane equipped with various excavation tools is used for removing the spoil from the bore of the piling tube and for forming the rock-socket. Once the oscillator machine has been set up on the pile position the initial casing which is fitted with a cutting edge is inserted into the oscillator and clamped. By oscillating the casing backwards and forwards and by raising the vertical rams the tube penetrates the soil under its own weight. Excavation of the spoil from the bore of the tube using a grab, proceeds concurrently with casing penetration. The casings, 2 to 6 metres long, are fitted with mechanical joints which allow them to be joined together. The boring process continues with additional casings added as and when required. Penetration of boulder layers is achieved by means of large chisels in combination with the cutting edge of the piling casing. Once the bedrock is reached the casing is allowed to penetrate into the rock to form a seal against the possible ingress of running sand. The socket is formed by chiselling the rock and removing the chips by means of a suction baler. This same unit is used to finally clean the bottom of the socket prior to concreting. Double wall casings are preferable since they are significantly more durable than single wall casings, and integrity at the casing joint is superior in difficult conditions. 130 The reinforcing cage fitted with roller spacers is lowered into position. If the pile has a permanent liner this is normally attached to the reinforcing cage and the two are lowered as a unit. Splicing of rebar cages and permanent liners is necessary on the deeper piles. If the pile excavation is dry then the concrete is deposited into the pile by means of a chute which prevents the concrete from striking either the reinforcing cage as it descends to the toe of the pile or the surface of the concrete. A minimum slump of 175 mm should be used and the mix should be designed to be self-compacting. The head of the pile is normally vibrated to assist compaction. On river bridges or marine work the pile excavation is invariably full of water. The concrete has thus to be placed under water using a tremie pipe which has a diameter of about 200 mm. The tremie is made up of different lengths joined together with a watertight screw-coupling. A high slump concrete is used for casting the pile shafts as it has to flow down the tremie and then self-compact. Some of the temporary pile casing can be removed during the concreting operation. Once the concreting of the pile shaft has been completed the remainder of the pile casing is withdrawn. Both the tremie pipe and the piling casings are washed thoroughly after each pile is cast. If the pile shaft does not have a permanent liner the depth that is concreted inside the temporary casing should be limited. The reason for this is that the concrete will tend to lose its workability with time, due to the pressure of the head of concrete and drainage paths at the toe of the casing and at the casing joints, which allow the water in the concrete to drain away. Concrete which has lost its workability will tend to arch in the pile casing during extraction which could cause defects in the pile shaft. Experience has shown that with a depth of about 30 metres the pile can be successfully concreted without these defects. Retarders and plasticizers have limited benefit in this situation, but their use is still recommended. Aspects of the installation process are shown in PLATES 7.10.1 to 7.10.3. VARIATIONS IN INSTALLATION TECHNIQUE Integrity Testing It has become common practice to carry out various integrity tests on these piles as they are heavily loaded and conventional load testing is very expensive. The contact between the base of the pile and the rock is normally checked with rotary core drilling. The integrity of the shaft itself is checked using either nuclear or sonic methods as described in SECTION 9.0: PILE LOAD AND INTEGRITY TESTING. To enable these tests to be carried out, three or four small diameter steel tubes are cast into the pile. These tubes are spaced equidistant around the perimeter of the pile and are normally fixed to the reinforcing cage and lowered with the cage. 131 POTENTIAL PROBLEM AREAS Stuck Casing There is a slight risk of this occurring on deep piles especially if the diameter is 1500 mm. To reduce the risk the casing should be kept moving so as to lubricate the outside and reduce friction. Stuck Chisels Chisels used in this type of work are large and heavy. They are dropped a couple of metres to break into the rock. They sometimes get wedged in the casing, or in the socket itself. It may take a few days to dislodge the chisel, and occasionally this may prove impossible. If this occurs the pile will have to be abandoned, the casing filled with sand and withdrawn. A revised pile layout will have to be designed and the pile installed on an alternate position. Tremie Concreting This is an operation which needs a large amount of experience on the part of the contractor for it to run smoothly and produce a sound pile shaft. Unfortunately even the best trained crews can run into problems, especially if the concrete itself is not one hundred percent. The most likely problem is a blockage in the tremie. This requires the tremie to be removed and cleaned out. If only a small amount of concrete has been discharged into the pile the best plan of action is to remove the steel reinforcing cage and permanent liner and to dig out the wet concrete using an excavating grab and chisel. Once cleaned out a new steel cage and liner (the old one will most likely be damaged) can be installed and the pile concreted. If a large amount of concrete has already been placed when the blockage occurs the tremie should be removed and cleaned out. A watertight end-cap should be placed on the toe of the tremie and the latter lowered until it penetrates the wet concrete by at least a metre. In this operation, as with all tremie operations, it is vital that the joints in the tremie pipe are absolutely watertight. This can be checked by shining reflected sunlight down the tremie pipe using a mirror. The tremie is then filled with concrete and raised slightly at which time the end-cap comes off due to the weight of the concrete. Thereafter the concreting operation can proceed as normal. Re-starting a tremie operation is a tricky process. For this reason the end result should be checked using one of the integrity testing methods. The alternative is to abandon the pile and install a replacement pile. 132 PLATE 7.10.1 Grab Excavating a Pile PLATE 7.10.2 Oscillator Installing Casing PLATE 7.10.3 Typical Oscillator Contract 133 7.11 ROTAPILES The main feature of this piling system is its ability to penetrate boulders and rock formations and socketing into hard rock is effected rapidly using the ‘Down The Hole Hammer’ (DTH) percussion drilling technique. This pile is particularly suited to pile installation in Karst conditions present in the Dolomitic areas of southern Africa. It is also suitable for river bridge foundations subject to scour, where penetration through boulder horizons and socketing into bedrock is required. The piles can easily be installed with a casing using propriety casing systems. The casings can be permanent and incorporated into the pile design to ensure pile shaft integrity in cavity formations or moving groundwater conditions. A range of pile sizes from 255 to 610 mm in diameter can be drilled with the equipment available in the southern African region. Diameters up to 1000 mm are possible using this technique, however, specialised equipment is required. .. .. .. . . . . POSITIVE FEATURES A good range of pile diameters are available for moderate applied loads The ability to penetrate hard rock and boulder horizons Rock-sockets can be formed relatively easily and economically Casing installation associated with the DTH drilling technique provides high levels of pile shaft integrity Noise levels are low and limited to the noise from the equipment There is limited vibration associated with the DTH operation Depths of up to 30 metres are possible depending on the diameter of the pile OTHER CONSIDERATIONS The pile type is relatively expensive compared to a driven pile with equivalent load capacity The introduction of large quantities of air by the DTH system into voided / sensitive subsoil conditions can cause problems with surrounding structures Grout must be used on smaller diameters PILE DETAILS The following nominal diameters apply to unlined rotapiles and corresponding permanently lined piles installed using a variety of propriety casing installation systems which are governed by available casing diameters. Since the piles can be founded in bedrock and heavily reinforced, high load capacities can be achieved especially with permanent thick-walled steel casings. For unlined piles the normal pile shaft working stress should be in the range 6 to 8 MPa, and working stresses in excess of 10 MPa can be adopted where piles are socketed into hard rock and incorporate a permanent liner. 134 PILE DETAILS Pile Diameter 355 mm 406 mm 255 mm 305 mm 457 mm 610 mm Shaft Diameter Lined (mm) 255 305 355 406 457 610 Nom. Shaft Diameter Unlined (mm) 250 300 350 400 450 600 Typical Working Load (kN) 300 - 450 450 - 600 600 - 900 Maximum Depth (metres) 20 20 30 30 30 30 Minimum Pile Spacing (mm) 750 900 1000 1200 1350 1800 800 - 1200 1000 - 1500 1500 - 2500 Maximum Rake 1: 4 1: 4 1: 8 1: 8 1: 8 1: 8 Typical Main Bar Reinforcing 4 x10 4 x12 6 x12 4 x16 6 x16 6 x 20 Typical Spiral Reinforcing at 150 Pitch 6 mm 6 mm 6 mm 6 mm 6 mm 8 mm Nom. Cover to Reinforcing (mm) 30 30 40 40 50 50 INSTALLATION TECHNIQUE Piling rigs currently used for the Rotary ‘Down The Hole Hammer’ (DTH) drilling systems are hydraulically powered units which are either track mounted to provide manoeuverability in difficult or sloping terrain, or truck mounted units where transportability is required and easy access and traversable site conditions apply. The ‘Down The Hole Hammer’ requires significant quantities of compressed air to operate the hammer and ensure adequate flushing of spoil from the hole is achieved. Drill bits activated by the hammer are specifically designed to penetrate hard formations and rapid penetration (up to 5 metres per hour) of hard rock is achievable. The drilling can be carried out unlined in hard soil or rock conditions where stability of the hole is ensured and no air loss is encountered during drilling. Where temporary linings are required, these are installed during drilling with specially designed couplings in lengths up to 6 metres. Where relatively shallow piles are required, a single length casing can be installed with suitable equipment. Piles of between 15 and 20 metres can be installed with a single casing using rigs with suitable masts or mast extensions. Where permanent casings are required, these are installed in lengths up to 6 metres generally, and casings are welded to provide the required penetration depth. Piles of unlimited depth can theoretically be installed, but maximum depths of 30 metres are generally applicable due to equipment and flushing limitations. Pile shaft reinforcement is placed in the drilled hole once the hole has been flushed clean. Piles are generally grouted on diameters of less than 450 mm and concrete is used for larger diameters in dry conditions. Where the drill hole cannot be kept in a dry condition, grout is generally used and is pumped using tremie techniques. It is inadvisable to use tremie concrete for piles less than 600mm diameter due to limitations in the size of the tremie pipe for concrete placement. The casing is withdrawn once the pile shaft is filled to the required cast level on temporarily cased piles. The Rotapile installation sequence is shown diagrammatically in FIGURE 7.11.1. 135 FIGURE 7.11.1 136 B C D E A. Drilling the pile with or without casing using a ‘Down The Hole Hammer’ (DTH). B. The drilling continues until the design founding depth or rock-socket is achieved and the hole is cleaned using air flush. C. The reinforcing cage is placed in position. D. The pile shaft is concreted or grouted using higher slump concrete / grout or the tremie technique if the pile shaft is filled with water. E. The temporary casing is removed once the concreting / grouting operations are completed for unlined piles. A ROTAPILE INSTALLATION SEQUENCE VARIATIONS IN INSTALLATION TECHNIQUE There are many DTH hammer designs and casing installation systems available which involve variations in flushing methods and the use of drilling foams to expedite spoil removal. The optimal drilling system is mainly governed by the anticipated soil conditions and extensive experience is required by the drilling contractor to find the optimal solution. POTENTIAL PROBLEMS The Rotapile is a pile type that is ideally suited to difficult ground conditions such as boulder profiles in riverbeds, or Karstic Dolomite profiles, and as such, is not generally subject to installation problems. One of the problems that may be encountered, particularly in open hole drilling, is the loss of air return and the inability to flush the drilling spoil. This pile type requires extensive skills in the choice of drilling methodology as well as a reasonably high level of operating skill if problems are to be avoided. The pile installation methodology is not suited to soft ground conditions and the use of other pile types (driven) would be significantly more economical. PLATE 7.11.1 shows Rotapile installation on a river bridge subject to extreme scour conditions for which this pile type is particularly suited. PLATE 7.11.1 Rotapile Installation on a River Bridge 137 7.12 MICROPILES The Micropile concept was initially developed in Italy by Fondedile in the 1950’s. It is recognized as an effective solution for underpinning existing foundations, as well as soil reinforcement in slope stabilisation, reinforcement of existing quay walls, and protection of buried structures. Micropiles are widely used in the support of light structures with high uplift load requirements, eg powerlines, and also as a foundation solution in difficult ground conditions, such as Karstic formations found in the Dolomite areas of southern Africa. A Micropile is a small diameter, typically less than 300 mm, drilled and grouted nondisplacement pile which is heavily reinforced and carries most of its loading on the high capacity steel reinforcement. The four methods of grouting micropiles are: .. .. .. . . . . . . . . . Grout placed under gravity conditions Pressure grouting through the casing Post grouting using single or multiple stages Grouting during drilling using high capacity steel threaded hollow-bar members POSITIVE FEATURES Micropiles can be installed in limited access and headroom conditions Micropiles can be installed with minimum disturbance to adjacent structures with the appropriate installation methodology Micropiles can be installed through existing foundations and are ideally suited to underpinning and as load enhancement of existing foundations Due to the high capacity steel reinforcing elements micropiles have high uplift load capacity and can be effectively used for tension structures Due to the wide range of installation methods available and the relevant ease of penetrating boulders or hard rock formations, micropiles can be economically installed in difficult ground conditions eg Karstic formations Micropiles can be utilised as soil reinforcing elements providing significant economies in suitable soil conditions where the applied load is shared between the base and the piles Micropiles can be installed as steeply raking piles providing significant horizontal load capacity for a pile group Micropiles generally provide a high degree of redundancy OTHER CONSIDERATIONS The relatively high cost of micropiles will preclude their use in normal access and soil conditions where conventional pile methodologies are more economical Careful assessment of groundwater conditions is essential to ensure corrosive durability of the main load-bearing steel reinforcing elements Micropiles are slender members and buckling effects need to be carefully assessed 138 PILE DETAILS .. .. There are a wide variety of installation methods available and the method of reinforcement can be in the form of: Steel casing Steel casing supplemented by internal reinforcement Heavy reinforcement without casing High capacity threaded hollow-bar members installed and grouted during drilling Micropile maximum axial load capacities of up to 2000 kN can be achieved. Franki have developed and implemented the drilled-in high capacity hollow-bar solution in co-operation with Ishebeck Titan, and typical details of this system are tabulated below. Bar Size Nominal Grouted Diameter (mm) Minimum Pile Working Load Working Load Spacing Compression Tension (mm) (kN) (kN) 30/11 100 300 225 180 40/16 150 450 450 350 52/26 200 600 650 500 73/45 250 750 1100 900 INSTALLATION TECHNIQUE There are many micropile installation techniques available which involve the drilling of the pile to the required depth, the placement of the reinforcement, and the grouting. A typical installation sequence is shown in FIGURE 7.12.1. 139 FIGURE 7.12.1 140 B C D E F A. Drill or install temporary casing. B. Complete drilling to design depth or into competent horizon. C. Remove drill bit and rods. D. Place reinforcement and grout by tremie technique. E. Remove temporary casing and inject further grout under pressure, as required. F. Complete pile (casing may be left in place through compressible stratum). A MICROPILE INSTALLATION SEQUENCE The drilling is usually carried out using hydraulic rotary power units which are track mounted to provide manoeuverability in difficult and sloping terrain. The drill mast can be adapted for low headroom conditions or fitted with mast extensions to facilitate the installations of long drill rod lengths. The holes can be drilled open with permanent or temporary casings installed using rotary wash boring, rotary percussive, hollow-stem augers (CFA), or sonic drilling methods. The drilling method adopted is chosen to suit the anticipated geotechnical conditions and the pile design requirements. The grouting of the micropile has a major impact on the design capacity since it provides the load transfer mechanisms of the high capacity steel reinforcement to the surrounding soil/rock. The grouts are designed to provide high strength and stability and are placed by pumping using tremie techniques. Grouts are usually made from neat cement water mixes with design compressive strengths in the range 25 to 35 mPa. Sand can also be used as a filler with suitable admixtures, such as plasticizers, to enhance the workability and reduce bleeding of the grout after placement. Grouting using neat cement water is used when installing self-drilling hollow-bar micropiles, with the pile being continuously grouted through the hollow-bar during the drilling of the pile. In many applications, the load-bearing element in the form of the thick-walled steel liners, or high capacity hollow-bars, is installed during the drilling phase. The additional reinforcement is placed with a centraliser after drilling is completed, but generally before the grouting commences. FIGURE 7.12.2 Typical Micropile Cross-section 141 VARIATIONS IN INSTALLATION TECHNIQUES As noted, there is a wide range of micropile uses and installation methodologies available. The methods commonly adopted are well documented in the reference manual Micropile Design and Construction, published by the USA National Highway Institute, Publication No. FHWA NH1-05-039, December 2005. POTENTIAL PROBLEM AREAS The most significant problem associated with this pile type is the use of the incorrect method of drilling for the actual geotechnical conditions encountered on the site. This can result in the inability to achieve the required design depth or pile shaft resistance. PLATE 7.12.1 Installation of a Franki Titan Hollow-threaded Bar Micropile Applicable Norms As the topic of Pile Installation Procedures and Techniques is broad, not all aspects can be adequately covered in one book. The reader can get further details on this important topic at the following links / references: BS EN 1536: Execution of Special Geotechnical Work – Bored Piles (2000) BS EN 12699: Execution of Special Geotechnical Work – Displacement Piles (2001) Publication No. FHWA NHI-05-039: Micropile Design and Construction (Dec 2005) Tomlinson, M., Woodward, J.: Pile Design and Construction Practice, 5th Edition (Taylor and Francis 2008) 142 8.0 UNDERPINNING The causes of cracking or other forms of distress in a building structure are usually due to either differential movement of the foundation, differential shrinkage, or temperature movements in the structure. Differential movement of the foundations can either be caused by uneven settlement of the foundation or by differential ground heave under the foundation. Before remedial measures are prescribed to stabilise foundation movement on a damaged structure it is essential that an adequate investigation is carried out to ascertain the nature of the foundation movement and in particular whether the movement is due to settlement or heave. If the differential movement is due to varying degrees of ground heave then the underpinning systems described in this section will not provide a complete solution to the problem. In fact it is debatable whether an underpinning solution is at all applicable in these circumstances. If the differential movement is due to uneven settlement of the foundation then an underpinning system is a feasible way in which to arrest further movement and stabilise the structure. Before deciding which underpinning system is best suited to solving the problem, the reasons for the settlement should be determined as well as details of the soil profile over the site. The history of the site coupled with knowledge of the local geological conditions can also assist in evaluating the cause of the settlement. When the investigation is complete, consideration must be given to which underpinning method should be used and which is likely to be the most economical. There are eight basic underpinning methods that could be considered as set out below: 8.1 Providing temporary support to the structure, removing the existing footing and replacing it with a new footing with greater bearing capacity, followed by the removal of the temporary support. 8.2 Excavation under sections of the existing foundation, constructing a new footing with better bearing capacity and ensuring load transfer to the new footing, with or without preloading. 8.3 Excavation under the existing foundation, jacking in piles and ensuring load transfer to the new piles, with or without preloading. 8.4 Installing piles alongside the foundation and casting a new section of footing which is keyed into the existing foundation. 8.5 Constructing a new foundation with or without piles and transferring load to this foundation by means of an additional column or similar structural member. 8.6 Installing piles through the existing foundation and keying the heads of the piles into the existing foundation. 8.7 Drilling a small diameter hole (150 - 250 mm diameter) through the existing foundation and installing a Micropile beneath the foundation, see SECTION 7.12. 8.8 Drilling a small diameter hole (250 mm diameter) through the existing foundation and installing a jet grouted column beneath the existing foundation, see SECTION 13.8. 143 FACTORS INFLUENCING THE CHOICE OF UNDERPINNING SYSTEM Which of the methods mentioned is best suited to the problem on hand depends on a number of factors. A list of the more important factors together with comments as to how these influence the choice of underpinning method are outlined below: Soil Profile It is necessary to know the soil profile to understand why the foundation movement is occurring in the first instance. Once this understanding has been obtained and a decision has been made that underpinning is the correct solution, the soils information is then vital in deciding at what level any remedial foundation should be founded so as to achieve the objective of preventing further settlement. If the use of piles is considered then detailed soils information will be required to assess the size and founding level of the piles, which in turn can influence the choice of pile type. Load on the Foundation For basic design calculations it is essential to have a good estimate of the load on the foundation. Structural Details of the Existing Foundation It is important to have this information so that the effect of any structural changes to the foundation can be checked. The use of piles along the perimeter of a footing could for example increase the bending moment in the footing, and thus the reinforcement needs to be checked for this increased moment. Structural Details of the Existing Superstructure Where additional columns and /or structural elements are envisaged it will be necessary to have these details. Level of the Water-table If excavation below the foundation is planned then it is important to determine the level of the water-table. A shallow water-table might well preclude excavation beneath the existing foundation due to risk of collapse, and the difficulty and high cost of shoring in these conditions. Access If any underpinning work is planned it is important to check that the workmen with their equipment can gain access to and around the foundation, so that the construction procedure can be adhered to. The presence of services should also be checked and taken into account in the planning. Available Headroom This is important, particularly if a piled solution is being considered. Inconvenience Inconvenience in the form of noise, vibration, security, site pollution and limited use of space may affect the public, landlord and /or tenant. Due consideration must be applied to these aspects when choosing a method and pile type for underpinning. 144 How Critical Further Settlement Would be There is a difference in performance between an underpinning method that employs preloading and one that does not. If preloading is not carried out then additional settlement can be expected during load transfer to the new foundation. It will be necessary to assess what additional settlement will take place and whether this additional movement will be detrimental to the structure. Whether it is Necessary to Raise the Footing that has Settled On some occasions it may be considered desirable to raise the level of the footing which is being underpinned. The normal requirement is to raise the foundation back to its original level, or even slightly above it. It is a high risk operation and one that should only be specified in exceptional circumstances. Careful planning and a high level of site supervision are needed for it to be successful. Cost of the Underpinning System As with all forms of construction the cost is very important. Detailed cost estimates need to be carried out to determine the most economical solution. Programme In some cases it may be that time is more important than cost. This can be especially true if an owner is losing rental because the tenants cannot occupy leased premises, or their trade is badly affected by the underpinning operations. 8.1 OLD FOUNDATION REMOVED AND NEW FOUNDATION PROVIDED To be able to do this it will be necessary to provide temporary support to the loadbearing member which is being underpinned. This can be achieved by fixing some brackets to the sides of the member and spreading the load through beams to stub columns supported on temporary footings or piles. This is illustrated in FIGURE 8.1.1. Once the load is temporarily supported the member can be cut through below the support point and the existing foundation removed. The new foundation is then constructed with a stub column connecting it to the existing column. The temporary support is then removed. This form of underpinning is not common. It is also expensive because of the necessity to provide temporary support and remove it afterwards, and is thus only used in exceptional circumstances. 145 8.2 NEW FOOTING LOCATED UNDER THE EXISTING ONE This is a more common form of underpinning and one which is considerably less expensive than that described in SECTION 8.1. It is used where the water-table is low enough to allow excavation under the existing foundation and where adequate bearing capacity can be obtained at a depth of one to two metres below the soffit of the existing foundation. This method involves an initial excavation of a narrow pit alongside the existing foundation. This pit must be excavated down to the anticipated new founding level and should be just large enough to allow reasonable working space. A narrow trench is then excavated from the pit extending under the foundation and down to the founding stratum. The base of this excavation is cleaned up and a reinforced concrete footing cast. A stub column of concrete, or load-bearing brickwork, is built on top of the footing, extending up to just under the existing foundation. The gap between the top of the column and the soffit of the existing foundation is filled with rammed sand/cement grout. The pit is finally backfilled. This method is illustrated in FIGURE 8.2.1. If preloading of the base is considered necessary, a gap large enough for an hydraulic jack is left between the top of the column and the soffit of the existing foundation. A pair of hydraulic jacks is positioned, one on either side of the column. The jacks are extended and a load of up to 50 percent in excess of the design load is applied to the column via the jacks. The level of the existing foundation is monitored to ensure that it is not lifting excessively and if such, the load is adjusted accordingly. Once the load has been held on for approximately 30 minutes, the load is reduced to between 100 and 125 percent of the working load and a fabricated steel stub column is placed in the gap between the jacks. Steel shims are used to finally ensure that the steel stub column is fixed firmly between the main column and the soffit of the existing foundation. The jacks are released and steel stub column is encased in concrete to protect it. The final operation is the backfilling of the pit. There are normally a number of underpinning points located at various positions under the structure. These are spaced at between two and four metre centres, the spacing depending mainly on the structural integrity of the wall and its footing to span between points and the design load of the new footing relative to the applied load from the structure. 146 - FIGURE 8.1.1 Old Foundation Removed and New Foundation Provided FIGURE 8.2.1 New Footing Located under Existing One 147 Before carrying out the underpinning it is necessary to plan the sequence of work. The more heavily loaded points should be installed first, and adjacent points should not be tackled simultaneously. Accurate bench-marks should be established at various points on the structure and these should be checked regularly during the underpinning operation. 8.3 JACK-PILES UNDER THE EXISTING FOUNDATION In cases where an adequate founding level is greater than about 1.5 metres below the soffit of the existing foundation consideration must be given to installing piles instead of constructing a new footing at great depth. A common way to install piles directly under the existing foundation is to jack them in short sections using an hydraulic jack and the load of the structure as reaction. A steel tube pile is commonly used for this purpose as the steel tube can be readily cut into short sections, or pile elements, and then welded together again as the pile is jacked into the ground. Precast elements have also been used as pile elements for underpinning but this is no longer common practice. The first step of this method involves the excavation of a narrow pit alongside the existing foundation. The size of the pit should be just large enough to allow working space. The depth of this pit should be about 1200 mm below the existing foundation. This allows for a 400 mm long jack and a 500 mm long pile element with the welded joint about 300 mm above the ground. The tube pile elements are normally jacked in open-ended, in which case the soil will tend to push up inside the tube during excavation. This soil is not normally removed as it is difficult to do so. In such instances the steel tube forms the structural body of the pile. This process is illustrated in FIGURE 8.3.1. In softer soils consideration should be given to welding a plate on the end of the first pile element. This will increase the pile's bearing capacity as well as leave the bore of the tube empty so that it can be filled with concrete or grout. The pile is normally jacked to refusal under a load in excess of the desired working load. This proof load can be up to 50 percent more than the required working load, but this must be adjusted during the underpinning operation, depending on the monitoring of the movements of the existing foundation. The proof load should be held on the pile for at least 30 minutes, or longer, if movement of the pile is detected. When the pile has been installed the tube is trimmed to the correct elevation and a steel plate is welded on to form a top bearing plate. A short section of steel RSJ is placed on top of the bearing plate and the jack in turn positioned on top of the RSJ. This is illustrated in FIGURE 8.3.2. The jack is extended and a load of 125 percent of the working load applied to the pile. 148 FIGURE 8.3.1 A Pile Being Jacked under an Existing Foundation FIGURE 8.3.2 Final Proof Loading of a Jacked Pile 149 Two short steel column sections are then placed, one either side of the jack, between the RSJ and the soffit of the existing foundation, shims are used to ensure a tight fit. The load on the jack is released and the load of the structure is transferred via the steel columns and RSJ to the pile. The steel columns and RSJ are normally encased in concrete to protect them. The final operation is the backfilling of the pit. There are often a number of underpinning points located at various positions under the structure. These are spaced at between two and four metre centres, the spacing dependent on the structural ability of the wall and its footing to span between points, and the capacity of the underpinning pile. Before carrying out the underpinning it is necessary to plan the sequence of work. The more heavily loaded points should be installed first and adjacent points should not be tackled simultaneously. Accurate bench-marks should be established at various points on the structure and checked regularly during the underpinning operation. 8.4 PILES ALONGSIDE THE EXISTING FOUNDATION This can be a more economical way in which to provide a piled underpinning solution than the Jack-pile system described in SECTION 8.3. It is also a quicker method than the jacking system, and piles of much greater capacity can be installed and to increased depths. Most of the pile types listed in SECTION 5.0 can be used with this method and the choice of pile type will be influenced by the factors enumerated in SECTION 4.0. In most cases it is the limited access, working area, and headroom constraints which will dictate which pile type can be installed. These restraints can be very severe and as a result special piling rigs and systems have been developed for the purpose. An example of this is shown on PLATE 8.4.1. This little rig, which is mounted on the chassis of a wheelbarrow, installs a driven micropile in the form of a steel tube 100 mm in diameter. It can manoeuvre through doorways in a house and into very narrow areas, eg between a toilet and bath. Another small rig used for this type of work is the conventional diamond drilling rig which can be used to install a drilled micropile of 100 mm diameter. There are many ways in which the load can be transferred onto the piles. One of the simplest methods is shown in FIGURE 8.4.1 (a) and involves a short cantilever tied and keyed into the existing footing. Another common method involves constructing a pile capping beam under the existing foundation as shown in FIGURE 8.4.1 (b). The construction of a completely new pile-cap over the top of the existing footing as shown in FIGURE 8.4.1 (c) is yet another method. The combination of the correct choice of size and type of pile together with one of the above methods for transferring the load to the piles can be used for a wide variety of underpinning problems. This method is suited to the underpinning of a house foundation at the lower end of the scale, and to underpinning a large bridge at the upper end. 150 PLATE 8.4.1 Miniature Piling Rig for Driving Micropiles 151 (a) (b) - (c) FIGURE 8.4.1 (a), (b) and (c) Piles Alongside the Existing Foundation FIGURE 8.6.1 Piles Through the Existing Foundation 152 8.5 NEW PILED FOUNDATION AND COLUMN There are occasions when the need to underpin is due to the fact that additional load is to be applied to the foundation, and not because settlement has occurred. A typical example is the addition of another floor or two to an existing building. Any of the underpinning techniques covered in this section could be used, but sometimes the structure itself also has to be strengthened. In these circumstances a completely new foundation, with or without piles, as well as a new column are constructed. This can be difficult work as low headroom and limited access restraints can place serious limitations on the size of piling rig that can be used, and thus also limits the choice of pile type. Here again unconventional solutions may have to be resorted to. An example is driving steel tube piles through the roof of a single storey building with a large crane. Another is the driving of 3 metre long sections of a mechanically jointed precast pile with a crane specially adapted to operate in a 6 metre headroom. Solutions to problems of this nature are best discussed with one of Franki's engineers and solutions developed for the particular application. 8.6 PILES THROUGH EXISTING FOUNDATION If a pile can be installed through an existing foundation and subsequently keyed into it, then there is the possibility that no additional structural concrete work will be required. The hole through the foundation can be cut with a diamond coring tool. The pile itself can be either driven or bored. Because of the task of drilling through the existing foundation and the fact that its structural integrity could be impaired, the size of hole and thus the size of the pile is normally limited. Piles with a diameter of up to 200 mm are used, although Micropiles with a diameter of about 100 mm are more common. The Micropile rig shown in PLATE 8.4.1 is often used for installing these piles. The structural transfer of load from the foundation to the pile is achieved by grouting up the hole in the foundation using an expanding grout once the pile has been installed and trimmed to the correct level. With shallow footings it is also possible to form an enlargement of the head of the pile so that it bears on the soffit of the footing. This is illustrated in FIGURE 8.6.1. This technique is fast and economical, but is generally suited to light structures only. PILE TYPES USED FOR UNDERPINNING Any of the eleven pile types listed in SECTION 5.0 could be used for underpinning. Factors such as those listed in SECTION 4.0 will assist in determining the choice of pile type. The steel tube pile is a frequent choice for underpinning as its features meet most of the requirements for underpinning. Special pile types developed for underpinning such as the drilled Micropile and the jacked Megapile can also be considered. Underpinning is an area suited to innovative solutions and a discussion with your local Franki office could be well worthwhile. 153 8.7 UNDERPINNING USING MICROPILES Micropiles are ideally suited to underpin existing foundations since they can be installed in limited access and headroom conditions, and can generally be installed through the existing foundation, thereby eliminating the need for an additional support structure to transfer the load from the existing footing to the underpinning pile. The Micropile solution generally results in the least disturbance to the existing foundation structure and minimises disturbance to occupants of the building. The Titan Micropile system is the most effective Micropile solution available since it requires small penetrations through the existing footing and provides good load transfer from the threaded hollow-bar into an existing reinforced concrete footing. The Titan system can be installed in very limited headroom conditions (2 metres) and in a wide range of soils and rock. High capacity Micropiles (up to 900 kN) can be installed in limited access conditions and can be very effective in supplementing existing foundation capacity where additions to existing buildings are required. Details of the available Micropile methodologies are outlined in detail in SECTION 7.12 and PLATE 8.7.1 shows underpinning with a Titan Micropile system. PLATE 8.7.1 Installing Micropiles Through an Existing Foundation 154 8.8 UNDERPINNING USING JET GROUTING Jet Grouting is used extensively in Europe for underpinning work. It has been used successfully in the African region by Franki Africa for the underpinning of the existing foundation structure on two multi-storey tower blocks for the Bank of Tanzania. The major advantage of the jet grout solution is that a large grout column, (up to 1.5 metre in diameter), can be installed with minimal penetration, (250 mm in diameter), through the existing foundation structures. A large diameter in-situ grout column can be formed beneath an existing foundation using the jet grout methodology outlined in SECTION 13.8: JET GROUTING. The jet grout methodology provides good load transfer from the existing foundation to the large grout body as the jetting process removes all compressible material within the grout column circumference, providing sound contact between the column grout and the existing foundation. Great care should be taken to ensure full spoil removal during the jetting process to minimise heave uplift of the existing structure during the formation of the jetting grout column. Constant monitoring of the structure should be carried out during jetting to ensure little or no uplift occurs. The large amounts of spoil generated by the jet grout operations must be considered since severe disruption to the occupation of the building could occur during the work on site. PLATE 8.8.1 shows Jet Grout underpinning of the Bank of Tanzania. PLATE 8.8.1 Installing Jet Grout Columns to Underpin Bank of Tanzania, Dar es Salaam, Tanzania 155 9.0 PILE LOAD AND INTEGRITY TESTING 9.1 PILE LOAD TESTING The load testing of piles is a well established practice and is often specified on medium and large piling contracts. The most common form of load test is the static compression test in which a load is gradually applied to the head of the pile while the deflection of the pile-head is monitored. Static test loading can also be carried out in tension, as well as laterally. Piles can also be tested using dynamic or semi-dynamic load testing procedures, but these are not commonly used in southern Africa. .. .. .. The main objectives in carrying out a load test are as follows: To verify that the pile's load/deflection performance meets the contract specification To verify the assumed pile design parameters To establish the pile's load/deflection characteristics to finalise the foundation design To assess the pile's ultimate capacity, if possible To study the pile/soil interaction To check the structural integrity of the pile shaft 9.1.1 STATIC LOAD TESTING Static load testing of piles can be carried out on working piles, or on trial piles specially installed for the purpose. Most pile load testing is carried out on working piles as the cost of installing additional piles for testing purposes is prohibitive on all but the very large contracts. Working Piles When load testing a working pile the load has to be limited so as not to damage the pile. For this reason the maximum test load is normally limited to one and a half times the design working load. In most cases this can be increased to twice the design working load without risk of damage to the pile, provided the load test is set up and executed in the correct manner. Tension and lateral load tests are normally limited to one and a half times the design working load. The principle reason for carrying out a load test on a working pile is to check the pile's performance compared to that specified in the contract documents. For this reason it is common practice for load tests on working piles to be part of the piling contract. The secondary benefits to be gained from such a test are the confirmation of the pile design parameters and a check on the pile's structural integrity. More economical means of testing pile integrity have been developed and are described in SECTION 9.2. The load test can also highlight whether installation problems, such as pile heave, have had any effect on the pile's performance. 156 Trial Piles Trial piles are specially constructed for the purpose of carrying out load tests and thus they can be loaded to failure. Testing a pile to its ultimate load capacity provides more accurate and meaningful design data which can result in achieving further economies in the foundation design. For structures sensitive to settlement the results from an ultimate test can provide the basis for a more accurate specification of working load settlement acceptance criteria. On large contracts the substantial cost of trial piles and an extensive load testing programme can often be recovered many times over through achieving economies in design. A trial pile programme also offers the opportunity to install more than one pile type and to compare their performance. This will assist in deciding which pile type will provide the most economical solution. Trial piles can also be installed to different depths so that the optimum founding level can be confirmed. Special instrumentation in the form of strain gauges cast into the pile shaft at various levels can also increase the amount of data that can be obtained form these tests. STATIC LOAD TESTING PROCEDURES Two compressive load test procedures are outlined in detail in SABS 1200F. These are termed the British and Danish procedures and both describe a series of test load cycles in which the pile is loaded gradually in increments and then unloaded in a similar way. Most specifications call for the use of the British method, or a variation thereof, as the Danish procedure is very time consuming. A common procedure, which is satisfactory for most soil profiles, is a variation of the British method. This involves a first cycle in which the load is increased in 25 percent increments up to the design working load, held for 12 hours and then unloaded in 25 percent increments back to zero. The intermediate load increments are maintained until two successive readings 30 minutes apart show that the head deflection has not changed by more than 0.1 mm. The load is kept at zero for a period of one hour after which the deflection is checked and the second load cycle is begun. The second cycle is similar to the first but the maximum load is 1.5 times the design working load. After unloading from the second cycle, the residual deflection is monitored over a 12 hour period. A typical pile performance specification will state that the pile-head should not deflect more than 8 mm under the design working load and not more than 15 mm under 1.5 times the design working load. The residual deflection after the second cycle should not exceed 6 mm. If the piles are very long and/or slender these figures may have to be adjusted to allow for the elastic shortening of the pile shaft. Allowance should also be made in the calculation of the allowable residual deflection for friction preventing the recovery of the pile. A typical result from a three cycle load test is shown in FIGURE 9.1.1. Procedures for tension and lateral load tests are not given in SABS 1200F. A two cycle load test procedure similar to that described above could be programmed for these tests. 157 STATIC LOAD TEST CONFIGURATION For piles loaded in compression the load should be applied concentrically to the pilehead using hydraulic jacks. A loading beam is used to transfer the load from the jacks to the source of reaction which can be Kentledge, anchors or anchor piles. The installation of anchors and anchor piles can take a considerable time, whereas Kentledge can often be provided at short notice. If a number of tests are required then Kentledge will be the most economical form of reaction. Load tests of up to 500 tonnes are possible with Kentledge. For greater loads, anchors or anchor piles must be used. The movement of the pile-head is monitored using a minimum of two deflectometers reading to 1/100th of a millimetre resting on glass plates cemented to the pile-head. The deflectometers are supported on a reference beam which in turn is mounted on posts driven into the ground at least two metres from the test pile. As the reference beam itself may move due to temperature effects, its deflection is sometimes monitored with dial gauges as well. It is advisable to have a backup system for monitoring pile-head movement, and this is normally provided by a precise dumpy level and scale rules fixed to the pile-head. A suitable reference bench-mark is located well away from the test area. In most instances it is sufficiently accurate to obtain the load on the pile from the product of the hydraulic pressure in the jack(s) and the area of the bore of the jack(s). It is common practice to have the jacks calibrated and the calibration certificate submitted with the test load results. A calibrated load cell should be used if a more accurate measure of the load is required. A typical test arrangement using anchor piles is shown in FIGURE 9.1.1. - - FIGURE 9.1.1 Typical Compression Test Arrangement 158 One of the load test configurations that can be used for a tension load test is shown in FIGURE 9.1.2. A central threaded, high capacity rod which is cast into the pile or fixed to the reinforcement, passes through a centre-hole jack, with a nut providing the seat for the jack-ram. The jack rests on a beam which transfers the load onto two spread footings, one either side of the pile. The measurement of load and deflection is achieved in a similar manner to that described for the compression test load. - FIGURE 9.1.2. Tension Load Test Configuration A lateral load test is most easily carried out by jacking two test piles apart. Measurement of lateral movement should be recorded by at least two deflectometers mounted on a reference beam with supports are well clear of the zone of influence. It is important for the accuracy of the test that the lateral load is applied through the centre line of the piles, thus eliminating any torsional effects. Such an arrangement is shown in FIGURE 9.1.3. FIGURE 9.1.3 Lateral Load Test Configuration 159 9.1.2 INTERNAL JACK METHOD OF PILE TESTING A recent development in the testing of large diameter bored piles is the use of an internally placed flat-jack which is installed above the end-bearing segment of the pile. The flat-jack isolates the end-bearing and shaft friction zones of the pile shaft, and is used to apply equal load to the end-bearing and shaft friction segment, providing separate assessments of the two components of pile load capacity. The movements and stresses in the base and pile shaft are monitored separately and provide full pile performance data to optimise pile design. The method is particularly suited to large diameter bored piles founded on, or socketed into, bedrock. The method requires no external heavy reaction assembly, and is therefore particularly suited to high capacity large diameter piles and piles in a marine environment. There are specialists in this field of pile testing who provide the necessary equipment and advice for a particular project. 9.1.3 LOAD TEST RESULTS A typical result from a compression load test is shown in FIGURE 9.1.4. The pile was loaded in three cycles to 1.0, 1.5 and 2.0 times the design working load with pile-head deflections of about 6, 10 and 15 mm and residual deflections of 2.5, 3.7 and 6.5 mm respectively. These results indicate a successful load test. Similar curves can be plotted for tension and lateral load tests. 9.1.4 INTERPRETATION OF LOAD TEST RESULTS There are a number of methods outlined in the literature that can be used for analysing a load test result. These methods are generally aimed at predicting the ultimate pile capacity and splitting the piles capacity into friction and end-bearing components. Two such methods are described by Chin and Vail (1973) and van Weele (1957). More recently methods for modelling the performance of piles on a computer have been developed and one of these methods is described by Everett (1991). With a computer model, the various parameters can be altered until the predicted load / deflection curve matches the actual. Once this has been achieved the computer simulation is an accurate model of the pile and every aspect of the piles performance such as ultimate load, load distribution between shaft friction and end-bearing, as well as end-bearing performance, can be obtained. The model can then be used to predict the performance of similar piles of varying sizes and depths in the same soil profile. FIGURE 9.1.4 Typical Compression Load Test Results 160 9.2 INTEGRITY TESTING OF PILES Due to the limitations, both from a logistic and economic point of view of testing a representative sample of working piles using the static load test methods outlined, non-destructive integrity test methods have been developed to aid the detection of defects in the structural integrity of pile shafts. It must be realised, that the results of these tests are not definitive and are subject to interpretation, but they do provide another tool with which to make an initial assessment. Any defect indicated by integrity testing should be investigated further using more positive testing methods. Integrity testing is at present carried out using either sonic or nuclear technology. The sonic methods involve either the collection of data from reflections of a sonic wave generated by tapping the head of the pile, or by the collection of data by a collector of sonic waves generated by an emitter, both the emitter and collector being lowered down the pile in separate small diameter tubes cast into the pile shaft for this purpose. These two methods are referred to as Sonic Impact and Sonic Logging. Nuclear testing is very similar to Sonic Logging except that a nuclear isotope is used to generate the signal. 9.2.1 INTEGRITY TESTING USING SONIC IMPACT With this test a sonic wave is propagated down the longitudinal axis of the pile while a transducer is held against the surface of the head of the pile. The sonic wave will be reflected off the toe of the pile, or off any intermediate defects, and these reflected waves are picked up by the transducer. A computerised signal processing unit records the reflected waves and prints out this record, which is known as a reflectogram. In this way a major defect in the shaft of a cast-in-situ pile, or a crack in the shaft of a precast pile, can be detected. The time taken for the wave to travel down the pile shaft and for the reflected wave to travel back up is recorded by the equipment and from this the length of the sound pile shaft can be ascertained. The system can thus be used to check the depths of existing piles, as well as the integrity of the shafts. The reflectogram shown in FIGURE 9.2.1(a) was taken from a sound pile. The pile exhibits even response over the full pile shaft length recorded as 12 metres in the pile records. The reflectogram in FIGURE 9.2.1(b) was taken from a defective pile and the difference in the two curves is very noticeable. The defect in this pile is indicated to be at a depth of about 2.7 metres. Interpretation of these reflectograms has to be carried out by an experienced technician with a considerable amount of experience. This method requires exposure of the pile-head which will have to be brushed clean or even trimmed so as to expose sound concrete. The tests can be carried out by one person, and forty to fifty piles can be tested in a day, provided the heads have been prepared. The test is thus a very quick and economical one. All pile types can be tested, but there are depth limitations. If the depth to diameter ratio exceeds about 40 to 50, then the reflected wave may be damped out by the friction on the pile shaft. The test is thus not successful on long slender piles where the ratio might be as high as 100. Joints in precast piles can also present a problem, as the wave tends to reflect off the joint. Not all pile joints have this problem. 161 There are a few variations in the type of equipment for carrying out this type of test, which the various suppliers claim have certain advantages. In essence however, the ability to detect a defect suffers from the same interpretation limitations. FIGURE 9.2.1(a) Reflectogram of a Sound Pile FIGURE 9.2.1(b) Reflectogram of a Defective Pile 162 9.2.2 INTEGRITY TESTING USING SONIC CORING The sonic coring method involves the lowering of electronic equipment down tubes cast in the shaft of the pile. These tubes are about 50 mm in diameter and can be made of steel or plastic. Normally three or four such tubes are cast in the pile at even spacing around the perimeter. The tubes have to filled with water before a test. The electronic equipment consists of a sonic wave transmitter, a sonic wave receiver, and a computer for storing the data and plotting out the results. To carry out a test the transmitter is lowered down one of the tubes while the receiver is lowered down one of the other tubes. Both the transmitter and the receiver are lowered at the same time by the same winch which has a means of automatically recording the depth. The sonic waves travel through the concrete between the transmitter and the receiver. If there is a defect, the pattern of the trace diagram is deformed as is shown in FIGURE 9.2.2(a). The test only covers a narrow band between the two tubes. If there are four tubes cast into the pile then a total of six different records can be taken. If there is a major defect this should be detected by at least one of these. On smaller piles three tubes are used, and on larger piles six or more tubes can be installed. A typical arrangement of tubes and the resultant coverage is shown in FIGURE 9.2.2(b). If a defect is detected with the equipment, further investigation in the form of visual inspection, rotary core drilling, or a load test should be carried out to determine the exact nature of the defect, and whether the performance of the pile is affected. A single sonic coring test result should not be used as the final arbiter of the integrity of a pile. The disadvantages of this form of test are the cost of the tubes and the fact that the piles to be tested have to be selected prior to casting. For large diameter piles which are normally heavily loaded, these disadvantages are often outweighed by the comfort of knowing that the integrity of the piles has been verified. 9.2.3 INTEGRITY TESTING USING A NUCLEAR ISOTOPE The nuclear method is similar to the sonic coring in that the device is lowered down a tube cast in the pile. The equipment available in southern Africa consists of a dual gamma ray emitter and detector. This obtains a measure of the density of the concrete by recording the amount of radiation reflected back. The device is lowered down the tube by means of a winch which automatically records the depth. The signal is recorded and processed in a similar manner to that in the Sonic Coring method. This type of test suffers from the same disadvantages as that of the Sonic Coring method. It however has one additional disadvantage, in that a nuclear isotope has to be transported from the laboratory to the site, with the strict controls that are placed on the transporting of such materials. 163 FIGURE 9.2.2(a) Typical Arrangement of the Sonic Coring Method FIGURE 9.2.2(b) Typical Arrangement of Monitoring Tubes 164 10.0 FACTORS INFLUENCING THE SELECTION OF A SOIL IMPROVEMENT SYSTEM Before choosing a soil improvement system the engineer must have detailed information from a site investigation, highlighting the existing conditions on the site. Based on this, along with a knowledge of the proposed structure and nature of the foundations and applied loads, the effect of the soil horizons below and adjacent to the structure can be assessed. This will highlight the need for improvement of either soil stiffness, shear strength, or both. . .. .. The following information is required to assess a suitable soil improvement system: Detailed geotechnical information including soil types, grading, Atterberg limits, depth of water-table, presence of cavities, rock level, fill etc All loads, loading conditions and foundation size (depth of influence) Allowable total and differential settlements Knowledge of the site and its environs Details of the various soil improvement systems available Once the information is available, consideration has to be given to the following points so that the most suitable system can be selected for the project: .. .. .. .. .. .. .. . .. .. .. .. STRUCTURAL Allowable bearing pressures on the improved soil The layout and size of the foundation structure The number and spacing of treatment points The effect of the process on the surrounding buildings, if any SOIL PROFILE Sections of the soil profile across the site and extent of any fill must be established Grading of the soils in the profile The presence and level of the water-table The depth of treatment to meet bearing and settlement requirements The degree of improvement necessary to meet bearing and settlement requirements The presence of very soft layers which cannot be compacted mechanically (presence of clay/silts and their degree of saturation) The ease or difficulty in achieving the required depth of improvement The presence of obstructions such as boulders, rubble, bio-degradable material etc ENVIRONMENTAL The sensitivity of the environs to vibration (shallow rock facilitates vibrations) The sensitivity of the environs to noise Problems associated with large volumes of water used in some treatment processes CONTRACTUAL Access to and from the site for equipment Headroom clearance on site for equipment The cost of soil improvement The cost of the footings The installation risks associated with each system The remoteness of the site The availability of skills and plant to carry out the work The availability of suitable material as replacement fill, if applicable 165 Most of these points and others are covered in SECTION 12.0: SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS and in SECTION 13: TECHNICAL DETAILS OF SOIL IMPROVEMENT SYSTEMS. An initial selection can be made from SECTION 12.0, but this should be checked by reading the more detailed information given in SECTION 13.0. There are a number of factors to consider and the assessment of some of these will be difficult for someone not experienced with soil improvement techniques. Should there be any doubt you are welcome to contact your local Franki office for advice. An incorrect choice can be an expensive and risky mistake, so it is advisable to make sure beforehand. 166 11.0 CLASSIFICATION OF SOIL IMPROVEMENT SYSTEMS The use of soil improvement techniques to solve geotechnical problems is on the increase, and numerous methods have been developed for this purpose. The following is a classification and listing of the soil improvement systems that are on Franki's product list. The section number for detailed information is quoted on the right (in brackets). SOIL COMPACTION Vibratory Compaction Dynamic Compaction Compaction Grouting (13.1) (13.2) (13.3) SOIL REPLACEMENT Vibratory Replacement Dynamic Replacement Driven Stone Columns (13.4) (13.5) (13.6) CONSOLIDATION Accelerated Consolidation (13.7) IN-SITU SOIL MIXING Jet Grouting Deep Soil Mixing Cutter Soil Mixing (13.8) (13.9) (13.10) There are other methods of improving the soil such as chemical grouting and soil reinforcement. The latter involves techniques such as soil nailing, reticulated micropiles and other forms of soil reinforcement. Soil nailing and reticulated micropiles are also lateral support systems and are covered under SECTIONS 17.7 AND 17.8. The other methods are not covered in this text. A summary of the details of these various methods is given in a readily referenced format in SECTION 12.0: SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS. For a full description of each of the systems and its method of application refer to SECTION 13.0: TECHNICAL DETAILS OF SOIL IMPROVEMENT SYSTEMS. 167 168 Soil Improvement System Dynamic Compaction Compaction Grouting 13.2 13.3 Dynamic Replacement Driven Stone Columns 13.5 13.6 Accelerated Consolidation Jet Grouting Deep Soil Mixing Cutter Soil Mixing 13.8 13.9 13.10 IN-SITU SOIL MIXING 13.7 CONSOLIDATION Vibratory Replacement 13.4 SOIL REPLACEMENT Vibratory Compaction 13.1 600 - 1200 Wide Wall 300 - 750 500 - 3000 N/A 400 to 600 1500 to 2500 500 to 1200 150 to 200 N/A N/A Approx. Column Diameter (mm) Continuous Wall 1000 - 2500 Varies with diameter/ loading Maximum 1200 - 2000 2500 - 5000 1500 - 2500 1000 - 2000 3000 - 7000 1500 - 2500 Spacing of Compaction Points (mm) UCS of 1 to 10 MPa UCS of 1 to 10 MPa UCS of 1 to 15 MPa N/A 300 to 750 kPa on column area 300 to 600 kPa on column area 300 to 500 kPa per point 100 to 200 kPa 100 to 250 kPa 100 to 250 kPa Typical Bearing Capacity 30 24 20 25 18 8 15 10 12 20 Normal Maximum Depth (metres) Medium Medium High Low High Medium High High Low Low Unit Cost per m2 per m SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS SOIL COMPACTION Sec. No. 12.0 Low Low Low Low Medium Medium Low Low Medium Low Noise Pollution Level Low Low Low Low Medium High Medium Low High Medium Vibration Level Medium Medium Medium Medium Medium Medium Medium Small Medium Medium Site Area Required 15 - 25 25 - 30 15 - 25 25 - 30 20 25 - 30 20 - 25 6 - 12 25 - 30 25 - 30 Normal Headroom Requirements (metres) SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS TECHNIQUE BENEFIT Higher Bearing Capacity Vibration Vibrocompaction Vibroreplacement Dynamic Compaction Vibratory Probing Compaction Piles Blasting Adding Load Pre-Compression Vertical Drains Inundation Vacuum Preloading De-watering Fine Soils ... ... ... ... ... ... ... ... ... . Less or Faster GroundReduced Increased More Settlement Water Liquefaction Erosion Even Time Control Potential Resistance Settlement ... ... ... ... ... ... ... ... ... ... . ... ... ... ... .. . .. . ... ... . . .. ... ... ... ... . ... . . ... ... . ... ... ... Embankment Piles . . ... . . ... . ... ... . ... ... . . ... Structural Reinforcement Reinforced Soil Soil Nailing Root and Micropiles Slope Dowels Embankment Piles Structural Fill Remove and Replace Displacement Reduced Load Admixtures Lime/Cement Columns Mix-in-Place Lime Stabilisation of Slopes Stabilisation of Sub-grades Grouting Permeation Hydro-Fracture Jet Grouting Compaction Grouting Cavity Filling ... .. ... ... . Other Methods Freezing Heating . ... . ... ... ... ... . .. .. ... . ... ... ... ... ... . . . ... Vegetation KEY: Improved Face/Slope Stability Main benefit or purpose Associated benefit or possible purpose 169 .. ... .. 13.0 TECHNICAL DETAILS OF SOIL IMPROVEMENT SYSTEMS 13.1 VIBRATORY COMPACTION Vibratory compaction is achieved using a vibratory immersion probe of one form or another. Compaction to considerable depths is possible. The degree of soil improvement is largely dependent on the grading of the soil, the natural resonant frequency of the soil, and the level of energy available. Where the grading of the soil is suitable, this system achieves very effective and economical compaction. .. .. . .. POSITIVE FEATURES Fast and economical system The degree of compaction can easily be checked Noise and vibration levels are low Compaction to depths of 20 metres is possible Suited to soil profiles with a high water-table OTHER CONSIDERATIONS The degree of compaction is sensitive to grading Not suited to materials with high silt and/or clay content SUITABLE SOIL PROFILES Not all soils are suited to compaction by deep vibrators. It is important to carry out a grading analysis and to study the grading of the soil to assess its suitability before deciding whether to use this technique. There are two methods that can be used for this purpose. One of these was proposed by Brown (1977) and involves the calculation of a suitability number β. The formula for calculating this number is as follows: β = 1.7 3 1 1 + + 2 2 (d50) (d20) (d10)2 (13.1a) Where d50 , d20 and d10 are the particle sizes in millimetres at 50, 20 and 10 percent passing by mass, a number less than 10 indicates a soil that is highly suitable for compaction by vibratory means, whereas a number in excess of 30 indicates an unsuitable soil. The second method was proposed by Mitchell and Katti (1981) and is illustrated in FIGURE 13.1.1. This consists of a grading envelope which shows the limits for the most desirable grading. The degree of compaction is very sensitive to the amount of silt and clay in the soil and should not exceed 15 percent. The clay fraction should not exceed 3 percent. 170 FIGURE 13.1.1 Soil Grading Suitable for Vibratory Compaction, Mitchell and Katti (1981) When the fines exceed the percentages stated, the increase in pore water pressure resulting from re-arrangement of the grains cannot dissipate quickly, and there is a tendency for liquefaction. Under these conditions, it is not possible to achieve any significant compaction with what is a relatively quick compaction method. Unfortunately the soils in southern Africa generally have a significant silt and clay content so compaction of this nature is not widely used. The more common soil improvement process for these conditions is Vibroreplacement, which is covered in SECTION 13.4: VIBRATORY REPLACEMENT. COMPACTION DETAILS Compaction points are normally spaced at between 1.5 and 2.5 metres centre to centre. The actual spacing is best decided upon by carrying out test compaction patterns and monitoring the results by carrying out ‘pre’ and ‘post’ compaction soil strength measurements using an in-situ test (CPT, SPT and DPSH), as well as carrying out plate load tests on the compacted soil. If adequate compaction is achieved then a shallow foundation can be placed on the improved soil, and can be designed using a bearing pressure of up to 250 kPa. The zone of soil compacted should extend beyond the edges of the foundation by about 10 percent of the depth treated. Normally the upper one metre will not be compacted using this technique and must either be removed, or compacted in-situ, using an impact roller or dynamic compaction. 171 INSTALLATION TECHNIQUE There are two basic types of vibrator used for deep compaction, either with horizontal amplitude or vertical amplitude. The former type has a built-in motor with eccentric weights which rotate about a vertical axis thus providing horizontal amplitude. The latter type consists of a separate vibrator unit with eccentric weights which rotate about the horizontal axis thus imparting a vertical amplitude. The vibrator clamps to a long slender metal section which is called a probe. Horizontal Amplitude The vibrator used with this system is a large immersion type with the motor and eccentric weights located at the lower end of the unit. The eccentric weights rotate about a vertical axis so the amplitude of the vibration is in the horizontal plane. The vibrating section is coupled to a follower section by means of a flexible coupling. A suspension point for handling the vibrator is located at the upper end of the follower section. The whole vibrator unit and part of the follower section becomes immersed in the ground during the compaction process. The vibrator section has two or more water-jets. Two high volume jets are located at the tip of the vibrator and are used for aiding penetration of the vibrator into the ground. On some vibrators there are another two low volume jets located above the vibrator and these are used to feed water into the cavity to keep the side-walls stable. The hydraulic fluid for the motor and the water for the jets are fed down from the head of the vibrator through pipes located in the hollow core of the unit. Large quantities of water are used in this process and can present a site control problem. A crane is normally used to lower and raise the vibrator. With the vibrator set up on the compaction position the motor is started and the high volume water-jets are activated. The crane lowers the vibrator into the ground. When the vibrator has penetrated to the full depth, the waterflow through the high volume jets is shut off, leaving the upper low volume jets to feed water into the cavity. The vibrator is raised slowly and then lowered again into the soil, which flows into the cavity under the tip of the vibrator. The raising and lowering of the vibrator continues in a repetitive cycle as it is gradually withdrawn. The waterflow must be controlled and at some stage it must be shut off completely. The effectiveness of the compaction can be monitored at all times using an oil pressure gauge mounted in the cab of the crane. The compaction process must continue up to ground level or a minimum of one metre above the footing soffit level. Vertical Amplitude The probe is a long steel section to which the vibrator is clamped at the head. The crosssectional shape of the probe can vary considerably and there are some patented types on the market. Not all probes have a constant cross-section and some are fitted with wings which protrude from a central column. 172 The vibrator, which is either electrically or hydraulically powered, is clamped onto the head of the probe using a hydraulically activated mechanism. The whole unit is suspended from a crane which lowers and raises the unit as required. With this system only the probe enters the ground. The probe with the vibrator attached is set up over a compaction position. The vibrator is activated and the probe is slowly lowered into the ground. There are no water-jets or any other means for assisting penetration, so the vibrator has to be powerful enough to drive the probe to the full depth. Once full penetration is achieved, the probe is lifted about a metre and then lowered again. This lifting and lowering is continued in a cyclical manner as the probe is gradually withdrawn. An indication of the degree of compaction can be obtained by monitoring the electric current in the case of electrically powered vibrators, and the hydraulic pressure in the case of hydraulically powered vibrators. The probe is not capable of compacting the upper one metre of soil due to limited containment, so this must be compacted using an impact roller or dynamic compaction, or alternatively, the level of the footing soffit must be below this depth. PLATE 13.1.1 shows a typical horizontal amplitude vibrator and PLATE 13.1.2 a typical vertical amplitude unit with a y-probe. VARIATIONS IN INSTALLATION TECHNIQUE Variable Frequency Vibration All soil profiles have a natural resonant frequency which varies depending on the profile. It has been found that the best compaction results are achieved when the vibrator is operating at this resonant frequency. In Europe, variable frequency vibrators have been developed and some contracts have been completed. This is the latest in deep compaction technology, but it has not been introduced to southern Africa as yet. POTENTIAL PROBLEM AREAS Variable Soil Profile Soft layers of silt, peat and clay cannot be compacted using any form of mechanical compaction. Even silty and clayey sands can prove difficult to compact. Soil profiles in southern Africa often have strata of these materials present, with the result that only a certain percentage of the profile can be compacted. This situation is not normally acceptable, and the vibratory compaction solution has to be rejected for this reason. When selecting a soil improvement system for a site, it is important to determine whether any of these soft layers are present. A common solution to the variable soil profile problem is the use of Vibroreplacement, which is covered in SECTION 13.4. 173 PLATE 13.1.1 A Typical Horizontal Amplitude Vibrator PLATE 13.1.2 A Typical Vertical Amplitude Vibrator 174 13.2 DYNAMIC COMPACTION Dynamic compaction is achieved by dropping a large weight known as a pounder from a considerable height onto the soil to be compacted. It is a compaction system that is very effective in the right conditions. It is fast, economical, and suited to lighter loaded structures, such as shopping centres, industrial buildings and low-rise residential buildings, as well as earthfills for roads and dams, where it will increase the bearing capacity of the in-situ soil and reduce settlement potential. Another major use is to reduce liquefaction potential. A wide range of soils can be compacted including fills which contain rubble. It is often used to solve collapsible soil problems under large loaded areas. .. .. .. POSITIVE FEATURES It is a fast and very economical compaction system Most soil profiles can be compacted Compaction can be achieved both above and below the water-table Fills contaminated with rubble, boulders and rocks can be compacted OTHER CONSIDERATIONS The impact shock wave can cause damage to surrounding buildings The depth of improvement with locally available equipment is limited SUITABLE SOIL TYPES All soil types with the exception of soft silts, peat and clays can be compacted using this system. Materials both above and below the water-table can be compacted. The Dynamic Replacement system is often used where there are soft silts, peat and clays, and is covered in SECTION 13.5: DYNAMIC REPLACEMENT. DETAILS OF COMPACTION POINTS The depth of compaction is a function of the weight of the pounder and the height of the drop. For the normal energy levels this approximates to the following relationship: D = k Wh Where: D k W h = = = = (13.2a) the depth of compaction in metres an influence factor which varies between 0.375 and 0.7 the weight of the pounder in tonnes the drop height in metres Experience has shown that the depth of influence and the degree of improvement are also influenced by the shape of the pounder. A wide range of pounders has been developed for varying site conditions, with different pounders often being used on different phases of the same contract. A useful rule of thumb given by Berry et al (2004) indicates that the depth of improvement is typically 3 to 5 pounder diameters and a typical soil improvement profile is given in FIGURE 13.2.1. 175 FIGURE 13.2.1 Typical Improvement Profile after Dynamic Compaction, Berry et al, (2004) 176 The compaction is usually carried out in three different phases known as the primary, secondary and ironing phases, in this order. Compaction of the deepest layer is achieved with the primary phase. The secondary phase achieves compaction mainly in the intermediate layers. The ironing phase ensures overlapping of the initial phases by compacting the shallow layers between the initial prints. FIGURE 13.2.2 illustrates how the various phases compact the different levels in a soil profile. The initial choice of spacing for the primary compaction points is based on experience, but one expects these to be between six and ten metres apart. Various field tests are carried out during the early stages of the contract to check on the level of compaction being achieved and the spacing of the primary points, while the energy input is also determined at this stage according to the results of the tests. Once the primary phase is complete, work proceeds on the secondary phase points. These are positioned midway between the primary phase points. Here again the energy input of the secondary phase points is determined by the results of field tests. The final ironing phase is aimed at compacting the upper two to four metres. In this phase the drop height of the pounder is limited (4 to 8 metres), and the whole area is compacted. COMPACTION TECHNIQUE The equipment consists of a heavy weight, which is referred to as a pounder, and a means of lifting and dropping this weight, which is usually a crane or a special frame fitted with a linear winch. Earth-moving equipment is used to backfill the craters formed by the dynamic compaction and to re-establish site levels. The compaction process involves the repeated lifting and dropping of the pounder on a compaction point. The number of times the pounder is dropped on one point is determined through tests on site. The sequence of points follows the primary, secondary and ironing phases as set out. PLATE 13.2.1 shows a typical pounder, PLATE 13.2.2 shows a crater formed by the impact of the pounder, and PLATE 13.2.3 shows a dynamic compaction contract in progress. VARIATIONS IN COMPACTION TECHNIQUE Collapsible Sands A typical collapsible sand has a relatively open grain structure with individual grains connected by a clay bridge. When the soil moisture content rises, the clay bridges soften, with a resultant loss in shear strength. The overall shear strength of the soil is thus high when it is dry, and low when it is wet. This change in the shear strength due to wetting can lead to the collapse of the grain structure of a soil under load, resulting in foundation failure. 177 FIGURE 13.2.2 Compaction Patterns for Primary, Secondary and Ironing Phases 178 If one attempts to compact a collapsible sand when it is dry, a high level of energy is required to do so. By wetting up (not saturating) the area to be compacted with water, before the compaction is to be carried out, the energy levels required for compaction are reduced dramatically, and densification using the dynamic compaction method can be readily achieved. POTENTIAL PROBLEM AREAS Excess Pore Water Condition Under saturated or near saturated conditions, the pore water pressure will increase with each blow of the pounder. If it becomes excessive the pounder will have little compacting effect, as the blow is being cushioned by the pore water. In these circumstances further compaction in these areas may have to be delayed until the pore water pressures have dissipated. In coarse grained materials the dissipation of excess pore water pressures takes place virtually immediately. With saturated clays on the other hand dissipation could take weeks, and such delays often make the dynamic compaction option impractical. Damage to Surrounding Buildings Due to its nature the generation of vibration by dynamic compaction is inevitable. While in open areas this is of little significance, problems can be experienced in developed areas unless correct precautions are taken. The magnitude of the vibration and the transmission thereof is greatly dependent on the nature of the materials being compacted, the depth of the compaction, the nature of the underlying materials, the presence of the water-table and the energy input procedure. As a general rule a saturated layer underlain by a hard rock will give the highest energy/vibration transmission. Techniques that have been developed to control and isolate vibration include the excavation of isolation trenches, compacting from reduced levels, reducing the energy input per blow, and the development of low vibration pounders. As a result of this, dynamic compaction has been carried out successfully immediately adjacent to existing structures. Vibration Limitations The sensitivity of various structures to vibrations varies and detailed analysis is complex. To simplify matters Franki has found that working to the following target and maximum levels of Peak Particle Velocity (PPV) generally yields safe results: .. Target < 25 mm /sec PPV Maximum < 50 mm /sec PPV 179 PLATE 13.2.1 A Typical Pounder PLATE 13.2.2 Crater Formed by a Pounder 180 PLATE 13.2.3 A Dynamic Compaction Contract in Progress 181 13.3 COMPACTION GROUTING Compaction Grouting is a soil compaction technique in which the density of the soil is improved by introducing a thick grout, under pressure, into the soil. The thick grout forms an enlarged bulb or series of bulbs in the soil and in so doing, it displaces the soil immediately surrounding the bulb, thereby increasing its density. It is a relatively expensive technique but one ideally suited to remedial work associated with soils of low density, such as poorly compacted fills. .. . .. . POSITIVE FEATURES Small rigs can get into difficult access and low headroom conditions No vibration Noise levels limited to the engine noise only OTHER CONSIDERATIONS Relatively expensive technique Low production rate, more suited to small contracts Risk of ground heave SUITABLE SOIL PROFILES The ideal soils for compaction grouting are loose sandy soils and gravels, above or below the water-table. Silty and clayey sands as well as partially saturated clays and silts can also be treated using compaction grouting, provided the soil mass has good drainage characteristics. The process cannot compact saturated clays. SOIL IMPROVEMENT DETAILS The diameter of the grout pipe is normally in the 50 to 100 mm range. The centres at which the points are arranged are in the 1.0 to 4.0 metre range but 1.5 to 2.0 is more common. If compaction near the surface is required, the points have to be positioned at the closer spacing. The full depth of a stratum can be compacted as a series of enlarged grout bulbs can be formed to cover the full depth. The overall maximum depth of treatment is normally limited to about ten metres with conventional equipment. INSTALLATION TECHNIQUE The grout pipe can be installed using either driving or drilling techniques. The sequence of grouting is generally planned as a series of primary and secondary compaction points. All the primary points are drilled and grouted first followed by the secondary points some days later. The secondary points are positioned midway between the primary points. The presence of the primary compaction points act as a containment when grouting the secondary points. A tertiary stage could be used as well if found necessary. 182 The grouting of each compaction point can be carried out from the bottom up, which is referred to as upstage grouting, or from the top down, which is known as downstage grouting. It is possible, and sometimes desirable, to use a combination of the two. With upstage grouting, the expanded bulbs are formed as the grouting tube is gradually withdrawn. With downstage grouting the uppermost bulb is formed first, and after the initial set has taken place, the grout pipe is drilled through the bulb to a lower depth where the next bulb is expanded. It is sometimes beneficial to form the top bulbs first in a downstage operation, followed by upstage grouting of the balance of the stratum. This technique is advantageous when there is limited overburden, as the upper enlarged bulbs act as a containing mechanism. The pressure at which the grout is injected has to be carefully monitored as excessive pressure will cause fracturing of the soil and result in ground heave. The pumps generally have a pressure capability of 40 bar but a limiting pressure of 20 bar at the head of the grout pipe is a typical figure for deep compaction. One of the objectives when compaction grouting a soil mass, is to attempt to even out the volume of grout injected over the whole area. The percentage replacement should be decided on and the volumes of grout controlled according to this figure. Ground heave must be monitored and the volumes reduced if heave is taking place. A sand/cement grout with a slump of between 25 and 75 mm is used for compaction grouting. It does not have to meet any strength requirements as the objective is not to form a structural element in the ground but to compact the ground itself. Cement contents can vary from zero to 500 kg per cubic metre, but 300 kg per cubic metre is more typical. Flyash is often used as a substitute for up to 50 percent of the cement as flyash extends the working life of the grout and improves workability. Retarders are also used for the same purpose. The grading of the sand is important to ensure the workability of the mix, even under high pressure. Often two or more sands are blended to produce the ideal grading. If a well-graded sand is not available, a bentonite slurry can be blended with the sand, and the cement partially substituted by flyash to aid the workability. Pumping rates should also be carefully monitored and controlled. The pumping rate should be in the range of between 15 and 100 litres per minute. The rate should be lower in soils with poor drainage characteristics and when the compaction process is carried out close to the ground surface. Higher rates can be used in free draining soils with significant cover. VARIATIONS IN INSTALLATION TECHNIQUE Raising Footings That Have Settled The fact that compaction grouting causes ground heave can be used to an advantage in that footings that have settled can be raised again using the compaction grouting technique. While the level of the footing is closely monitored further pumping of grout is undertaken until the footing has risen to the required level. 183 13.4 VIBRATORY REPLACEMENT This method is commonly referred to as Vibrocompaction. It is a replacement method of soil improvement, and crushed stone is the most common form of replacement material. The system has been used in the coastal areas of southern Africa for the past forty years and has proven to be a reliable product. It is ideally suited to structures with a large area of uniform distributed loading such as tank bases, the ground floors of warehouses and industrial buildings, as well as road embankments. The solution has also been used on bridges, multi-storey buildings and silos. .. .. . . POSITIVE FEATURES Well-proven soil improvement system Low noise levels limited to engine noise only Low vibration levels, except close to vibrator Can provide an economical form of foundation Fast installation rate OTHER CONSIDERATION Uses large quantities of water which needs good site management SUITABLE SOIL PROFILES As described in more detail later, the Vibrocompaction method forms a compacted stone column in the ground which behaves under load in a similar manner to that of a pile. In forming the stone column the vibrator will also compact the surrounding in-situ soil providing it has a suitable grading. See SECTION 13.1 for methods to assess this suitability. The lack of improvement in individual soil layers is however not a problem, because the stone column can transfer the load through these layers. If the layers are very soft or are thicker than one third of the column diameter, the stability of the stone column should be checked. A method for carrying out this check is given in Section 21.0: DESIGN AIDS - SOIL IMPROVEMENT. This process is commonly used in sand and silty sand soil profiles in the coastal areas of southern Africa. SOIL IMPROVEMENT DETAILS The vibrocompaction columns are generally about 1000 to 1100 mm in diameter. They are designed to carry loads of between 300 and 500 kN. The spacing varies between 1500 and 2500 mm with 1500 and 1750 mm being the more common spacings. The spacing is often the result of field tests carried out on site to determine the effectiveness of the soil improvement. 184 INSTALLATION TECHNIQUE The vibrator is of the immersion type with the motor and eccentric weights built into the unit. The vibrator section is coupled to the follower section by means of a flexible coupling. There are water-jets located at the tip of the vibrator so as to assist initial penetration and to keep the annulus around the vibrator clear during the compaction stage. Fins welded to the vibrator assist in preventing rotation of the vibrator in the ground, which is a natural reaction to the spin of the motor. The vibrator is set up over the compaction point and the motor and water-jets are activated. The vibrator is then lowered slowly so that it penetrates the ground. A slight tension is maintained in the crane cable so as to keep the vibrator in the vertical plane. When the vibrator has reached the treatment depth, it is surged up and down a few times so as to enlarge the annular gap around it. The waterflow to the jets is then reduced, the vibrator is lifted about 1.5 metres and a quantity of crushed stone is fed into the crater formed around the vibrator. The vibrator is lowered so that it penetrates into the stone which is now lying at the bottom of the hole. The vibrator compacts the stone to a high degree, as monitored during the process by means of the current drawn by electrical vibrators, or alternatively, the hydraulic pressure on hydraulic vibrators. The cycle of lifting the vibrator, feeding in the stone, and lowering the vibrator again, is repeated until the complete stone column has been formed. The vibrator is shown in PLATE 13.1.1 in SECTION 13.1. A Vibratory Replacement contract in progress is shown in PLATE 13.4.1. POTENTIAL PROBLEM AREAS Very Soft Soils The presence of very soft layers is generally one of the reasons for choosing the vibratory replacement method in the first place, but these same layers can still present a problem if they are excessively thick. The stone column manufactured by the process needs the lateral support of the ground to be able to carry axial load. Soft layers do not provide strong lateral support and thus the vertical carrying capacity of the stone column can be limited. If the soft layers are less than one third of the column diameter, then there is limited concern provided there are only one or two such layers. If the soft layers are much thicker, then the stability of the stone column should be checked. See SECTION 21.0. Silty or Clayey Soil Profile For the compaction of saturated soils to take place the pore water has to dissipate. If the soil profile is saturated and has a large silt or clay content, then excessive pore water pressures will develop and the whole soil mass will become liquefied. Under these conditions the formation of stone columns will not be possible. 185 PLATE 13.4.1 A Vibratory Replacement Contract in Progress 186 13.5 DYNAMIC REPLACEMENT This technique is used in very soft cohesive soil profiles, usually fine grained with a high moisture content, where the compaction of the in-situ material is not possible. The soft soil is replaced by columns of stone, rubble or other suitable materials which are driven using a special pounder designed for this purpose. These large stone columns can form a more rigid foundation for many types of structure, including low-rise buildings, earth dam walls, road embankments etc. The stiffer materials used attract load and are often covered with a soil mattress to assist with arching over the soft compressible, less frictional materials. .. . .. . POSITIVE FEATURES Dynamic Replacement is an economical solution in difficult soil conditions Noise levels limited to the engine noise of the plant Large diameter stone column with significant load carrying capacity OTHER CONSIDERATIONS Limited depth of installation using conventional equipment (6 to 8 metres) Shock wave from pounder impact Ground heave must be monitored TYPICAL SOIL PROFILE The ideal soil profile has an upper sandy stratum which is one to two metres thick, underlain by two to six metres of soft compressible material, which in turn is underlain by competent soil or rock. As this is a replacement process the grading of the in-situ sandy soils is not of significance although the process will compact the in-situ materials provided they have a suitable grading. The stability of the stone column may need to be checked if the soft stratum is very soft and/or its depth is greater than half a column diameter. See SECTION 21.0. DETAILS OF DYNAMIC REPLACEMENT POINTS The diameter of the stone columns is usually between 1.5 and 2.5 metres. Larger diameters are possible with purpose made equipment. The minimum spacing between the edges of the columns should be one metre. The load capacity of each individual column can be calculated using a column stress of between 300 kPa for very soft profiles to as high as 600 kPa for more competent profiles. The cut-off level for the capping footing should be a minimum of one metre below the natural ground surface. INSTALLATION TECHNIQUE The equipment used for carrying out dynamic replacement consists of a pounder and device for lifting and dropping the pounder. The latter is normally a crawler crane but special lifting frames fitted with linear winches are sometimes employed for the larger contracts and heavier pounders. 187 The crane is set up in position and the pounder is dropped from a reduced height so as to form an initial crater. This crater is then filled with rock or rubble after which the pounder is dropped the full height. The energy drives the stone into the ground displacing the in-situ material. Additional charges of stone are added to the crater each time it re-forms and thereby the stone column is driven deeper and deeper. Records are kept of the quantity of stone used and checks are made regarding the depth of the stone column. The column has to be driven so that it penetrates through the soft stratum and into the denser founding stratum. Once this has been achieved the crane moves on to the next position. If the stone columns are closely spaced excess displaced material will be forced to the surface and allowance should be made for the removal of this material from site. PLATE 13.5.1 and PLATE 13.5.2 show Dynamic Replacement in progress. POTENTIAL PROBLEM AREAS Inadequate Penetration In a dynamic replacement solution it is essential that the stone be driven down through the full depth of the soft layer. Should the problem of lack of penetration occur, the solution is either to increase the mass of the pounder, or to reduce the density of the soil by pre-drilling prior to commencing the dynamic replacement operation. PLATE 13.5.1 A Dynamic Replacement Crater 188 PLATE 13.5.2 A Dynamic Replacement Contract in Progress 189 13.6 DRIVEN STONE COLUMNS This method involves the bottom driving of a steel piling tube as used in installing Franki piles, see SECTION 7.1. Once the tube has been driven, the stone column is formed by expelling measured quantities of stone out the tube, using an internal drop hammer. The displacement caused during the driving of the tube and the forming of the stone column results in compaction of the surrounding soil. The stone column acts as a structural member in much the same manner as a pile shaft. Stone columns have been used for the foundations of light to medium structures including buildings and bridges. .. . .. .. POSITIVE FEATURES Depths of up to 18 metres can be treated (similar depth limitations as the Franki pile) It is a clean system Noise levels are low OTHER CONSIDERATIONS Production rates are low The relative cost of the method is high A medium level of vibration is associated with this system Ground heave is a potential problem SUITABLE SOIL PROFILES The method can be used in any soil profile that will not heave during the driving of the tube. It should be avoided in saturated cohesive soil profiles as these soils are the most problematic when it comes to ground heave caused by displacement. DETAILS OF STONE COLUMNS The stone columns can be made with either a 410, 520 or 610 mm diameter piling tube. The minimum spacing should not be less than 2.5 times the tube diameter. The allowable working loads can be based on a shaft stress of between 300 and 750 kPa depending on the strength of the surrounding soil. INSTALLATION TECHNIQUE The equipment consists of a piling rig, piling tube and internal drop hammer. The piling tube is located in the mast of the piling rig. The piling rig is set up on position and the tube is lowered onto the ground. A plug is formed at the toe of the tube by placing a charge of crushed stone into the tube and compacting this with a few blows of the hammer. The drop height of the hammer is then increased and the tube is driven into the ground. 190 On reaching the required depth the tube is held by the piling rig while the hammer is used to drive the plug out of the tube. A charge of crushed stone aggregate is then placed in the tube, the tube is extracted a little using the extraction winch, and the stone is expelled using blows of the hammer. This cycle is repeated as the tube is gradually withdrawn forming a continuous stone column over the full depth. VARIATIONS IN INSTALLATION TECHNIQUE Compacting a Pile Founding Stratum A situation can arise where there is a competent founding stratum for a piled foundation, however considerable benefit can be derived from the additional compaction of that stratum prior to the installation of the piles. In order to achieve this compaction, stone columns are formed in the stratum material for the full depth of the stratum. The degree of compaction can be controlled by varying the spacing of the stone columns as well as the quantity of stone expelled to form the column. POTENTIAL PROBLEM AREAS Ground Heave The driving of the piling tube causes displacement. In saturated silty and clayey soils this displacement can often cause ground heave where the soil being displaced moves outwards and up. This upward movement of the soil imparts a tension into the previously installed stone columns. As they are not capable of resisting tensile forces, the columns part and a gap is formed. This is obviously detrimental to the loadbearing performance of the columns and the affected columns have to be rejected. In most cases the only solution to the problem of displacement heave is to change to some other soil improvement or piling system. 191 13.7 ACCELERATED CONSOLIDATION When a preload is applied to a saturated soil there is an instantaneous increase in the pore water pressure. The rate of dissipation of the pore water and the consolidation associated with it depends on the permeability of the soil and the length of the drainage path. When the soil is relatively impermeable and the drainage path is long, it will take considerable time for the pore water pressures to normalise and for full consolidation to take place. Accelerated consolidation is a technique which involves the introduction of drains into the soil to reduce the length of the drainage path and thus decrease the time taken for consolidation to take place. It is used in situations where a large preloaded area, such as a road embankment or material stockpile, is underlain by a considerable depth of very soft silt or clay. .. .. .. POSITIVE FEATURES A well established method to improve considerable depths of soft cohesive soils Minimal vibration associated with installation of the soil drains Noise levels are low and limited to engine noise during installation of the drains A relatively low cost method OTHER CONSIDERATIONS A time consuming method The cost of importing and removing any surcharge material DETAILS OF SOIL DRAINS FOR ACCELERATED CONSOLIDATION There are three main types of drain used for accelerating consolidation: The sand drain, the sandwick drain and the band drain. The band drain has become the most popular, as it is the most economical, due to its fast rate of installation. The soil profile suitable for band drain installation needs to be fairly soft or loose as the mandrel installing the drain is driven into the ground. Drilling methods are mainly used for sand and sandwick drain installation and therefore can be installed in stiffer and denser soil profiles. The Sand Drain This was one of the first systems used for consolidation drains and is simply a column of highly permeable sand formed in the ground. A hole is initially formed by either drilling or driving and this is filled with sand of a suitable grading. If the drilling technique is used then the drains can be installed in a fairly stiff or dense soil profile. A temporary casing is often used to keep the hole open. The casing is extracted during the placing of the sand to reduce the risk of arching in the tube. The diameter of these drains is normally in the 150 to 250 mm range but larger sizes can be used if necessary, although the cost will be considerably greater. 192 Various problems have been experienced with this form of drain which have seriously reduced the efficiency of the drains. In some cases disturbance of the soil during the installation procedure has reduced the permeability of the soil immediately around the drain thus reducing the effectiveness of the system. In other cases the column of sand has become necked due to incorrect installation procedure, or due to high lateral soil pressures, and this has also reduced the effectiveness of the drains. These negative case histories have tended to reduce the use of sand drains to situations where there is confidence that these problems will not occur. The Sandwick Drain The sandwick drain is very similar to the sand drain with the exception that the sand is contained within a sock made out of a geofabric. The sand-filled sock is referred to as the sandwick. The hole is formed in the ground in a similar manner to that used for the sand drain and the sandwick is lowered into the hole. The diameter of the wick is between 50 and 75 mm so the installation can be readily achieved by conventional drilling techniques. The presence of the sock assists the contractor with the installation of the drain and reduces the risk of discontinuities and necking. Problems can still be experienced with the installation procedure affecting the permeability of the soil immediately around the drain. The Band Drain The band drain has taken over from sand type drains to a large degree. The main reasons for this are the fact that band drains are very economical, they are strong and able to resist necking, squeezing and buckling, and the small mandrel used to install them causes a minimum of disturbance to the surrounding soil. In ideal conditions the speed of installation is very fast, taking only two to three minutes to install each drain. The soil profile does however need to be soft or loose, as the mandrel used to install the drain is driven into the ground by means of a vibrator. The band drain itself is about 100 mm wide and between 2 and 7 mm thick. It consists of a strip of flexible cardboard or plastic which has longitudinal drainage channels formed in it. In some cases this strip is fitted with a surrounding filter sleeve. The band drain is supplied in large rolls. The equipment for installing band drains consists of a crane with a leader, a vibrator, and a hollow mandrel. The mandrel is rectangular in shape with a length exceeding the depth to which the drains need to be installed. Connected to it, at the head and at various intervals over its length, are guides located in the leader. The vibrator is clamped to the head of the mandrel. The band drain is fed in at the top of the mandrel and emerges at the bottom, allowing a plastic anchoring shoe to be attached to it. The band drain is supplied in a coil which is mounted on a spool on the leader. 193 The vibrator drives the mandrel with the anchor shoe and band drain into the ground. The coil of the band drain unwinds as the mandrel penetrates. When the desired depth has been reached, the mandrel is extracted leaving the band drain behind in the ground. Once the mandrel is clear of the ground a cut is made through the band drain just above ground level. Another anchor shoe is attached to the piece protruding from the tip of the mandrel and the crane moves to the next position. The installation of the mandrel does cause some disturbance to the surrounding soil, but experience has shown that this is limited to 2d, where d is the diameter with the same circumference as that of the band drain. Other Forms of Drain Whilst the three types of drain mentioned are the most common used, there are other forms which can be considered. Stone columns have a low permeability and can function as drains as well as act as structural columns, see SECTION 13.6. Drainage Blanket The water which flows up through the drains under a preloaded area has to be led off to the sides of such an area. This is achieved by placing a layer of sand with a high permeability over the tops of the drains prior to placing the remainder of the fill or surcharge material. This layer, which is normally about 500 mm thick, is referred to as a drainage blanket. In certain instances there may be a natural drainage layer at the surface, in which case there is no need for any additional measures to be taken. Design The design of drainage systems is covered in SECTION 21.0. 194 13.8 JET GROUTING Jet Grouting involves the mixing and partial replacement of the in-situ soil with cement slurry as opposed to conventional grouting which involves the injection of cement slurry into the voids in the soil. In its simplest form the process involves the ejection of cement slurry from a rotating grout tube, fitted with a nozzle, under very high pressure. The jet cuts a path outwards from the grout tube in a radial direction, the cement mixing with the coarse particles in the soil while replacing the fines. The combination of rotation and gradual step-wise withdrawal enables a large diameter in-situ soil grout column to be formed in the ground. Typical grout column geometry is graphically illustrated in PLATE 13.8.1. .. .. . .. . POSITIVE FEATURES A wide range of soil types can be treated Jet grouting is a vibrationless system Noise levels are low and limited to engine noise only It has features which provide unique solutions to difficult geotechnical problems Large diameter columns of up to 3 metres can be formed OTHER CONSIDERATIONS It is a relatively expensive technique Ground heave can occur, although with good control it can be avoided Spoil is generated when pre-cutting clays and must be disposed of SUITABLE SOIL PROFILES The ideal soil type for jet grouting is a clean, loose, medium to coarse sand. The sand particles are readily eroded away by the grout jet and therefore the jet is able to penetrate up to half a metre radius with a single jet, and up to over a metre radius with air and water assisted grouting. Gravels are also amenable to treatment using jet grouting, especially the finer gravels. Larger particles such as cobbles will tend to shield the jet and limit the size of the grouted column, making the column cross-section very irregular. Cohesion in the soil tends to reduce the ease with which the particles are eroded. The diameter of the grout column will reduce as the silt and/or clay content increases. In silty sands the reduction in diameter can be 15 to 30 percent. With purely cohesive soils such as silts and clays the diameter is even further reduced to roughly half that in clean sand. The stiffness of the cohesive soil is also important and only soils with a very soft consistency (SPT value of up to 6) should be regarded as being suitable. Clays with a higher stiffness require pre-cutting using the high pressure water-jet. The clay cuttings are then forced to the surface, when jet grouting on the second pass, from the base of the column. Good curing is achieved when jet grouting below the water-table, however the water/ cement ratio may need some adjusting. INSTALLATION TECHNIQUE The equipment consists of a crawler mounted drill rig, grout mixing plant, a high pressure grout pump and a grouting tube fitted with high pressure nozzles. The grouting tube, usually 50 to 75 mm in diameter, is either drilled in by the machine itself, or placed in a pre-drilled hole to speed up the process. A temporary casing, or a bentonite slurry, is used to keep the holes open. The holes can be vertical or inclined at a rake. 195 The jetting operation involves the pumping of a cement slurry under high pressure so that it emerges from the nozzle, at the base of the grout tube, at a high velocity. If a cylindrical column is required the jet tube is rotated and gradually raised in small steps at a constant rate. Other shapes, such as flat panels, can be formed by omitting rotation of the grout tube. The simplest form of jet grouting involves a single jet of cement slurry which is pumped in at pressures of up to 600 bar. A double jet grouting system has been developed to increase the column diameter with the introduction of a jet of air which acts as a shroud around the cement slurry jet. An air pressure of between 2 and 15 bar is used. This system has been modified to achieve even greater column diameter, with the addition of a high pressure water-jet for eroding away the soil, and is known as the triple jet system, which can achieve twice the radial penetration of the single jet system. The water pressure is high, up to 500 bar, but the grout pressure is reduced to between 5 and 30 bar. Any excess soil in suspension flows up the annular gap between the grout tube and the soil to the surface from where it is channeled into settling ponds for subsequent removal from site. The SG of the soil can be measured to obtain an indication of the column diameter formed. The SG of the initial grout mix is typically around 1.5. The addition of an air-jet assists this excess material to rise to the surface and keep the annulus clear. Blockage of the annulus can cause a build-up of pressure which can result in soil fracturing and resultant ground heave and should be avoided. The diameter of the grout column is a function of the speed of withdrawal, the soil type, the system of jets and the pressures used. Different pressures and / or rates of withdrawal have to be used in different soil strata to produce a grout column of reasonably constant diameter. Field tests are necessary to determine what these parameters should be for the various soil strata. PLATE 13.8.3 shows jet grouting being used to underpin an existing raft foundation for a large tower block. APPLICATIONS OF JET GROUTING .. .. .. . Jet grouting has a wide range of applications which include: Forming of grout columns to support structural loads Forming a contiguous wall for a caisson or cofferdam Forming a contiguous wall for lateral support Forming cut-off walls for groundwater control Forming a base seal to an excavation below the water-table Underpinning of foundations and quay walls Sealing between piles on a contiguous pile wall construction VARIATIONS IN INSTALLATION TECHNIQUE Horizontal Jet Grouting It is possible to install jet grouted columns in the horizontal plane. This technique has been used in various parts of the world to form a protective portal arch over a soft ground tunnel excavation. 196 PLATE 13.8.1 Typical Grout Column Geometry PLATE 13.8.2 Jets in Operation 197 PLATE 13.8.3 Jet Grout Columns Being Installed in Tanzania 198 13.9 DEEP SOIL MIXING Soil mixed columns are formed by in-situ mixing of a binder (cement or lime) with the soil. This mixing is achieved in-situ by means of a specially designed mixing tool, fitted on the end of a hollow kelly, driven by an auger machine. The system is mainly suited to the improvement of softer loose soil profiles. Soil columns can be used in these soil types to improve bearing capacity and reduce settlement. They have also been used in a secant pile form for lateral support, embankment stabilisation and reduction in the risk of liquefaction. The addition of lime also increases the permeability of the clay and lime columns can be used as drains for the acceleration of consolidation. Soil mix columns can be formed to depths of up to 25 metres. .. .. .. POSITIVE FEATURES Versatile and flexible system Fast and economical in certain soil profiles It is a vibrationless system Noise levels are limited to engine noise only OTHER CONSIDERATIONS Bulk supplies of the binder must be economically available Equipment not readily available in southern Africa at present THE STABILISATION EFFECT OF THE BINDER The shear strength of the soil decreases during the mixing operation. Thereafter a shortterm increase in the shear strength of stabilised clays and silts is caused by flocculation of the clay and by a reduction in the water content. This is followed by a long-term increase in shear strength resulting from various pozzolanic reactions when the binder reacts with silicates and aluminates in the clay. Observations have shown that approximately one third of the final shear strength is achieved after one month and seventy five percent after three months. The undrained shear strength of lime stabilised clay will increase from 10 to 50 times the initial shear strength. In favourable conditions undrained shear strengths of up to 10.0 MPa have been achieved with the use of a cement binder. INSTALLATION TECHNIQUE Deep mixed columns normally have a diameter of 400 to 600 mm. The spacing of the columns depends on the application. For lateral support they can be installed in a contiguous layout. Under foundations a spacing of between 1200 and 2000 mm is common. Larger spacings are sometimes used for other applications. The equipment normally used, is either a powerful top drive drill rig, or an auger rig. The machine is fitted with a hollow kelly which allows the dry binder to be blown down into the mixer-head at the bottom of the drill stem. The mixer-head is the same diameter as the required column and consists of a pair of blades on either side of the kelly, located on the diameter. The blades are shaped so as to obtain good mixing of the binder with the soil. Wet binders can also be introduced, but lower column strengths are associated with additional water added, and therefore dry mixing is the preferred method. 199 The installation commences with the drilling of the mixer-blade into the ground down to the full depth of treatment. With the blade rotating the binder is blown in as the kelly is gradually withdrawn. The withdrawal rate has to be controlled so that the binder content is relatively constant throughout the length of the column. The quantity of binder is normally in the range 6 to 10 percent relative to the dry weight of the soil. Other materials such as gypsum, cement and flyash can be blended with the lime and have given improved results in organic clays with high water content. SOIL MIX COLUMN APPLICATIONS Soil mix columns have been used extensively in countries such as Scandanavia and Japan where there are considerable depths of soft clay, or loose liquefiable sands. The main applications are: Enhancement of Bearing Capacity Soil mix columns can be effectively used to improve the in-situ shear strength and stiffness of soft soils supporting embankments and other forms of distributed surface loading. The columns require a granular soil mattress to transfer surface load to the soil improvement columns. In the case of embankments, the embankment itself will provide adequate load transfer. Foundations for Structures Soil mix columns with a diameter of 400 to 600 mm are installed in groups under the foundation footings in a similar layout to piles. The columns are not designed to carry the full foundation load but rather to act in combination with the surrounding soil to provide the required bearing. The solution is only suited to lightly loaded structures as the strength of the columns is not high. The soil mix columns act more as settlement reducers and to limit differential settlement. Lateral Support Soil mix columns can be installed to form a secant pile wall for lateral support. A typical application is in the form of a circular cofferdam which acts as the pit for a pipe jacking operation. The pipe can be jacked through the soil mix columns, which is not the case with concrete piles. Soil mix columns can also be installed behind a sheet pile or diaphragm wall to improve the stability of the wall, as well as in embankments. Drains Lime columns installed in a deep soft clay profile under a road embankment will increase the settlement rate and also act to control the settlement. Columns placed outside the loaded area will contribute to increasing the overall shear resistance on any potential failure surface. Limitation of Liquefaction Potential Deep soil mix columns have been successfully used in Japan and other seismic regions as containment elements within liquefiable loose sand layers. This containment eliminates the risk of bearing capacity failure during a seismic event. 200 13.10 CUTTER SOIL MIXING (CSM) Bauer Maschinen GmbH have adapted their well established trench cutter technology to construct in-situ soil mixed walls which can be used for: .. . Soil Improvement Cut-off Walls Soil Retaining Structures The trench cutting technology comprises a pair of rotary cutters which excavate a rectangular trench in a wide range of soils and soft rock to depths of up to 25 metres. The width of the rotary cutters varies from 500 to 1200 mm and the length of the excavated rectangular panel varies from 2400 to 2800 mm. The cutters have been adapted for soil mixing to facilitate full disintegration of the soil structure with the addition of a binder to form a homogeneous mass of binder and soil. Conventional binders are normally used. Two methods of mixing are adopted. The first method comprises the addition of a binder, usually a cement / bentonite slurry during the downward driving of the mixing tool. The in-situ mixing process is continued to the required depth of penetration and the mixing tool is removed while rotating the mixer-head further, enhancing the homogenization of the soil mix panel. The second and more common method of installation is the use of a two phase mixing process. The first excavation phase comprises the mixing of bentonite during the downward driving of the mixing tool and then adding the binder (cement / bentonite) on the upward stroke. The soil mixed wall is constructed in panels up to a maximum depth of ± 25 metres, using a ‘Fresh-in-Fresh’ sequence, where a panel is cut into the previously formed panel, ensuring full continuity through the panel joint. A ‘Hard-in-Hard’ sequence can also be adopted, where secondary panels are excavated into the adjoining semihardened primary panels. Reinforcement can be installed into the completed soil mix wall in panels similar to diaphragm wall construction and this solution can provide an economical temporary retaining structure. Limited penetration of the mixing tool into rock is feasible, since adequate mixing cannot be achieved if extensive rock penetration is required. Applicable Norms As the topic of Soil Improvement is relatively new and broad, not all aspects can be adequately covered in one book. The reader can get further details on this important topic at the following links / references: BS EN 12715: Execution of Special Geotechnical Work – Grouting (2000) BS EN 12716: Execution of Special Geotechnical Work – Jet Grouting (2001) CIRIA C 514: Grouting for Ground Engineering (2000) CIRIA C 573: A Guide to Ground Treatment (2002) 201 14.0 FACTORS INFLUENCING THE SELECTION OF A LATERAL SUPPORT SYSTEM In this context a lateral support system is considered to be a system which provides stability to any surface excavation or constructed slope, eg basement excavations, cut slopes, fill slopes etc. With such a broad definition it is apparent that there is a wide range of lateral support systems that could be used for slope stabilisation. Certain authors, Tomlinson (1970), consider the selection and design of lateral support systems to be an art rather than a science. Whether one agrees with this, or not, it must be accepted that the selection, design, and installation of a lateral support system requires the application of considerable skill, experience, and sound engineering judgement. .. .. . The following is the information required to select a suitable lateral support system: A detailed layout of the proposed surface excavation or constructed slope Detailed geotechnical information, including shear strength parameters Allowable ground movements Knowledge of the site and its environs Durability requirements (is it a temporary or permanent solution?) Using this information the designer will need to consider the following aspects in evaluating the most suitable lateral support system: .. . .. . .. .. . .. .. . SITE CONDITIONS Topography before and after construction Possible variations in site geology, or soil profile close to the site Groundwater conditions PROPOSED DEVELOPMENT Geometry and depth of the excavation or slope Proximity of the excavation to site boundaries Possibility of incorporating the lateral support system into the permanent works ADJACENT DEVELOPMENTS Presence of buried services in close proximity to the site Surcharge loading (dead, live or transient?) Feasibility of installing support systems into adjacent property The extent of ground movement acceptable during and after construction The sensitivity of adjacent developments to ground movements CONTRACTUAL Restrictions on anchoring / nailing into adjacent properties Access to and from site for plant Availability of skills and plant Availability of materials The remoteness of the site 202 Considering all the factors mentioned, an initial evaluation of a suitable lateral support system can be made from the details given in SECTION 16.0: SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS, this should however be checked using the more detailed information given in SECTION 17.0: TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS. There are numerous factors which need to be taken into consideration before deciding on a suitable lateral support system, and the assessment of these factors will be difficult for someone not experienced in lateral support. An incorrect choice can lead to delays with resulting extra costs to a project. If there is any doubt contact the nearest Franki Africa office for expert advice and assistance. 203 15.0 CLASSIFICATION OF LATERAL SUPPORT SYSTEMS The different lateral support systems which can currently be installed in the African region have been classified as either embedded walls or reinforced soils. 15.1 EMBEDDED WALLS An embedded wall uses an external structural wall against which stabilising forces are mobilised. Passive soil pressures are an integral part of the stability of embedded walls. The structural wall forms an integral part of the overall system and is either designed as a cantilever to provide the primary method of support, or to act as a support system against which braced or tie-back forces are mobilised. Typical examples of embedded walls are shown in FIGURE 15.1. A cantilever wall is probably the simplest method of providing support. The wall relies on the passive resistance generated by penetration below the excavation depth to provide the required support. It must be noted that large displacements are required to fully mobilise the passive resistance and hence there is risk of large movements with such a system if not carefully designed. The wall system should be installed before excavation commences and should remain in contact with the soil at all times. The disadvantages are that the cantilever system is generally only suitable for limited depths of excavation, 3 to 5 metres, and besides the wall stiffness itself there is no other positive method of controlling adjacent ground movements due to excavation. Space restrictions can also limit the maximum width of the wall, which in turn limits the height and stiffness of the cantilever. For excavations deeper than 3 to 5 metres additional stability can be provided by braced or tie-back support systems. Bracing is usually in the form of horizontal or raking props. In recent years post-stressed anchors have become the most popular tie-back support system. This type of system allows a positive force to be mobilised onto the wall element, which in turn transfers the force onto the excavation face. A major advantage is that the post-stressing procedure can be used to limit adjacent ground movements. Anchored walls are suitable for stabilising a wide variety of soil and rock types. Walls with multiple anchors have been used to support very deep excavations, 25 metres or greater, and excavations with deep-seated failure surfaces. In most multiple anchor wall applications the anchors generally act as the primary support system to provide overall stability, with the wall system acting as a structural element against which the anchor forces are mobilised. The various elements associated with embedded wall systems have been classified as follows: TYPES OF EMBEDDED WALLS Steel Sheet Piles Precast Concrete Sheet Piles Steel Soldiers Concrete Soldier Piles Contiguous and Secant Pile Walls Diaphragm Walls (17.1.1) (17.1.2) (17.1.3) (17.1.4) (17.1.5) (17.1.6) 204 EMBEDDED WALL SUPPORT SYSTEMS Prop Supports Post-stressed Anchors Threaded Hollow-bar Anchors (17.2.1) (17.2.2) (17.2.3) The number given on the right (in brackets), refers to the specific sub-sections in SECTION 17.0: TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS, in which detailed descriptions are given for the individual support systems. The most important factors relating to each of the lateral support systems are summarised in tabular form in SECTION 16.0: SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS. This allows for a quick and easy comparison between the various systems when evaluating a suitable system for a specific project. CANTILEVER BRACED TIED-BACK FIGURE 15.1 Examples of Embedded Walls 205 15.2 REINFORCED SOILS Walls that are not embedded for stability are internally stabilised by the installation of reinforcing elements which extend beyond any potential failure surface, and do not rely significantly on passive soil pressures for their stability. This system relies on shear transfer along potential failure surfaces to mobilise stabilising forces, the objective being to make the soil behave as a rigid block. Some movement is required to mobilise stabilising forces within the reinforcing elements. FIGURE 15.2 shows typical applications for the following three internally stabilised lateral support systems that can be provided by Franki Africa. GeoNails Reticulated Micropiles Soil Dowelling (17.3.1) (17.3.2) (17.3.3) Anchors are sometimes also used as a means of reinforcing soils. The number given on the right (in brackets), refers to the specific sub-sections in SECTION 17.0: TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS, in which detailed descriptions are given for the individual support systems. The most important factors relating to each of the lateral support systems are summarised in tabular form in SECTION 16.0: SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS. This allows for a quick and easy comparison between the various systems when evaluating a suitable system for a specific project. Although there are some fundamental differences in the mechanical action of the three internally stabilised systems, the illustrations in FIGURE 15.2 show that there are circumstances where more than one technique may be used as lateral support. Research work, Jewell (1980 and 1991), has shown that internally stabilised systems work most efficiently when the reinforcement is angled across a potential failure surface so that the reinforcement is loaded mainly in tension. The following general conclusions may therefore be arrived at in this regard: . . . Where a steep slope is to be excavated, Detail (a) in FIGURE 15.2, it is more efficient to use GeoNails installed close to the horizontal. To stabilise such a slope using reticulated micropiles will require a much higher density of reinforcement; In marginally stable, relatively flat slopes, Detail (b) in FIGURE 15.2, where overall stability must be improved, then either GeoNails or reticulated micropiles could be used. Other factors, eg access for plant or geological conditions, may be more significant in deciding on a final reinforcement system; In flat slopes of soft clay soils for example, where stability is governed by a well defined failure surface, Detail (c) in FIGURE 15.2, then large diameter soil dowels would be the most appropriate. 206 (a) Excavated Steep Slopes (b) Marginally Stable Flat Slopes (c) Flat Slopes with Well-defined Failure Surface FIGURE 15.2 Examples and Applications of Soil Reinforcement 207 16.0 SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS Ref Type Nominal Size (mm) Nominal Spacing (metres) Additional Secondary Support Normal Max Depth that can be Supported (metres) Vertical LoadBearing Capacity 17.1 EMBEDDED WALLS 17.1.1 Steel Sheet Piles Per supplier's details Continuous None Cantilever: 3 Braced: 10 Anchored: 15+ Poor 17.1.2 Concrete Sheet Piles Per designer’s Continuous details None Cantilever: 3 Braced: 10 Anchored: 15+ Fair 17.1.3 Steel Soldiers Standard H section, RSJ or channel profiles 1 - 2.5 Timber lagging or gunite Cantilever: 3 Braced: 10 Anchored: 20+ Fair 17.1.4 Concrete Soldier Piles 300 - 1200 diameter 1 - 2.5 Gunite Cantilever: 4 - 5 Braced: 10 Anchored: 25+ Good 17.1.5 Contiguous and Secant Pile Walls 300 - 1200 diameter 0.5 - 1.0 x diameter None Cantilever: 4 - 5 Braced: 10 Anchored: 25+ Good 17.1.6 Diaphragm Walls None Cantilever: 4 - 6 Braced: 10 Anchored: 25+ Good 17.2 Width 400, 500, 600, Continuous 800, 1000, 1200, 1500 EMBEDDED WALL SUPPORT SYSTEMS 17.2.1 Prop supports 300 - 1200 diameter 3-8 Anchors possible 5 - 20 — 17.2.2 PostStressed Anchors 165 1.5 - 5 Piles < 30 — 17.2.3 Anchor Piles 500 - 1500 Varies Varies — Good 17.3 REINFORCED SOILS 17.3.1 GeoNails 80 - 125 diameter 1.0 - 2.0 vertical and horizontal Gunite 12 Poor 17.3.2 Reticulated Micropiles 80 - 250 diameter 0.15 - 1.0 None or gunite 8 Good 17.3.3 Soil Dowelling 450 - 1500 diameter 1.0 - 3.0 None or gunite 8 Good NOTE: For specific details on bracing and anchoring refer to SECTIONS 17.2.1 to 17.2.3. 208 SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS continued Ref 17.1 Type Establishment Costs Cost per m 2 Noise Site Area Pollution Required Flexibility in the Event of Obstructions Water and Collapse EMBEDDED WALLS 17.1.1 Steel Sheet Piles Medium High High Medium Poor Good 17.1.2 Concrete Sheet Piles Medium High High Medium Fair Good 17.1.3 Steel Soldiers Medium Medium High if driven, else low Medium Fair Poor 17.1.4 Concrete Soldier Piles Medium Medium Low Medium Fair to Good Poor 17.1.5 Contiguous and Secant Pile Walls Medium High Low Medium Fair to Good Fair to Good 17.1.6 Diaphragm Walls High High Low Large Fair to Good Good 17.2 EMBEDDED WALL SUPPORT SYSTEMS 17.2.1 Prop Supports Medium Medium Low Large Fair Good 17.2.2 PostStressed Anchors Medium Medium Low 5-6m bench Good Use hollowbar in poor soils 17.2.3 Anchor Piles Medium Medium to High Low Medium Poor Good (driven) Poor (bored) 17.3 REINFORCED SOILS 17.3.1 GeoNails Low to Medium Low to Medium Low Small Good Poor 17.3.2 Reticulated Micropiles Medium Medium to High Low Small Good Fair 17.3.3 Medium Medium to High Low Medium Poor Good (driven) Poor (bored) Soil Dowelling NOTE: For specific details on bracing and anchoring refer to SECTIONS 17.2.1 to 17.2.3. 209 17.0 TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS 17.1 EMBEDDED WALLS 17.1.1 STEEL SHEET PILES One of the easiest and quickest ways in which to form a watertight retaining wall, in soft or loose saturated soil profiles, is to use steel sheet piles. These are steel sections which have the facility to interlock, one with another, and which can be driven into the ground to form a watertight wall. The sheet pile sections can be extracted once they have performed the function for which they were installed, reducing costs considerably. The steel sheet piles can either be used as cantilever walls or as braced / tie-back walls. .. . .. .. .. .. .. POSITIVE FEATURES Fast method for forming a wall or cofferdam in soft or loose saturated soil profiles Sheet piles can be extracted and used many times, thus reducing costs Wall flexibility can be used in the design to reduce wall pressures, leading to economical designs OTHER CONSIDERATIONS Sheet piles are imported and are expensive Delays in procurement can delay the start of a contract Installation needs to be carried out by persons skilled in the operation High noise levels associated with installation if driven TYPICAL USES / NEED FOR SHEET PILE INSTALLATION Temporary cofferdam for the construction of the pile-cap for a pier in a river Temporary cofferdam for the construction of a pump-house below ground level Temporary cofferdam for the construction of a basement to a building Permanent wall as part of the construction of a harbour quay Permanent cut-off wall to restrict the flow of groundwater Temporary or permanent retaining walls As the cost of steel sheet piles is high, their use is generally only justified economically for temporary support of excavations below the water-table in soft saturated soil profiles. Despite the high costs, sheet piles are often used for certain permanent works or in situations where the speed of installation is a distinct advantage. They are commonly used in marine construction. STEEL SHEET PILE SECTIONS Steel sheet piles are hot-rolled steel sections with a shape typical of that shown in FIGURE 17.1.1. Each sheet pile has a pair of clutches formed in a way that allows the male clutch of one sheet to interlock with the female clutch of another. When a number of steel sheet piles are interlocked they form a wall commonly referred to as a steel sheet pile wall. 210 Each manufacturer has a range of sizes to meet various requirements. FIGURE 17.1.1 shows U and Z type profiles commonly manufactured in Europe. The sheet pile sections can be ordered in whatever length is required. There is however a maximum recommended length which is dictated mainly by the handling and driving. These lengths are given for each section profile in TABLE 17.1.1. TABLE 17.1.1 Maximum Recommended Lengths of Steel Sheet Piles Moment of Inertia of Section cm4 per Metre Width Maximum Recommended Length in Metres 4110 13513 23885 39831 49262 92298 6 14 18 23 24 26 Ninety degree corners, forty-five degree corners and junction sheet piles are also manufactured to enable the installation of sheet pile walls to a desired plan configuration, such as, a rectangular cofferdam with cross-walls. Other shapes can be fabricated locally from standard sheet pile sections. Steel sheet pile sections are not manufactured on the African continent and are normally imported from Europe. They can be manufactured from various grades of steel in accordance with EN10249-1-1995, as set out in TABLE 17.1.2 for cold-formed U and Z sections. The length of a sheet pile can be successfully extended by welding on an additional section. TABLE 17.1.2 Details of Grades of Steel and Working Stresses Grade of Steel Minimum Yield Strength ReH (MPa) Tensile Strength Rm (MPa) Former Reference in the UK S 235 JRC S 275 JRC S 355 JRC 235 275 355 340 - 470 410 - 560 490 - 630 40B 43B 50C When shipped from the manufacturer the sheet pile sections are plain and uncoated. Should there be a requirement to coat the sheet piles with a protective coating this is normally carried out locally, or on site. The sheet piles are sandblasted prior to applying the coating. An epoxy tar is one of the most effective coatings. 211 Section Width Height Thickness b mm h mm t mm s mm cm2/m kg/m of single pile kg/m2 of wall cm4/m cm3/m AU14 AU16 AU17 750 750 750 408 411 412 10.0 11.5 12.0 8.3 9.3 9.7 132 147 151 77.9 86.3 89.0 104 115 119 28710 32850 34270 1410 1600 1665 AU18 AU20 AU21 750 750 750 441 444 445 10.5 12.0 12.5 9.1 10.0 10.3 150 165 169 88.5 96.9 99.7 118 129 133 39300 44440 46180 1780 2000 2075 AU23 AU25 AU26 750 750 750 447 450 451 13.0 14.5 15.0 9.5 10.2 10.5 173 188 192 102.1 110.4 113.2 136 147 151 50700 56240 58140 2270 2500 2580 Section Width Height b mm h mm t mm s mm cm2/m kg/m of single pile kg/m2 of wall cm4/m cm3/m AZ12 AZ13 AZ14 670 670 670 302 303 304 8.5 9.5 10.5 8.5 9.5 10.5 126 137 149 66.1 72.0 78.3 99 107 117 18140 19700 21300 1200 1300 1400 AZ17 AZ18 AZ19 630 630 630 379 380 381 8.5 9.5 10.5 8.5 9.5 10.5 138 150 164 68.4 74.4 81.0 109 118 129 31580 34200 36980 1665 1800 1940 AZ25 AZ26 AZ28 630 630 630 426 427 428 12.0 13.0 14.0 11.2 12.2 13.2 185 198 211 91.5 97.8 104.4 145 155 166 52250 55510 58940 2455 2600 2755 AZ34 AZ36 AZ38 630 630 630 459 460 461 17.0 18.0 19.0 13.0 14.0 15.0 234 247 261 115.5 122.2 129.1 183 194 205 78700 82800 87080 3430 3600 3780 Thickness Sectional Area Sectional Area Mass Moment Elastic of Section Inertia Modulus Mass Moment Elastic of Section Inertia Modulus FIGURE 17.1.1 Steel Sheet Pile Sections from Arbed 212 INSTALLATION TECHNIQUE Steel sheet piles are normally installed by driving them into the ground. The driving can be achieved by using a vibratory type hammer, a drop hammer, or any diesel, hydraulic or air hammer. Prior to driving, the sheets have to be lifted up one at a time and threaded one end into the other. A crane is used for this purpose and a scaffold platform is often constructed to provide support for the assembled sheets until such time as they are driven. For smaller cofferdams it is advisable to assemble the whole cofferdam prior to driving. Driving has to be carried out in a controlled manner as the sheet piles have a tendency to fan out, resulting in non-verticality in the direction of the line of the wall. A slight non-verticality is not critical on a straight wall, but it becomes a problem when the wall has to make a 90 degree corner. These problems can be overcome with planned pre-assembly and controlled driving where all the sheets in a section are driven down one or two metres at a time. External lateral force can be provided to the head of the sheet pile by means of a steel cable and a jack, as an additional measure to control verticality. If verticality is not maintained in driving a section of sheet pile, then this can be corrected by driving a special tapered pile. This should be considered when approaching a corner or when closing a cofferdam. Tapered piles can be fabricated from standard sheet pile sections. When using a conventional hammer a helmet is provided to protect the head of a sheet pile during driving. The helmet should be close fitting and rigid if it is to distribute the blow of the hammer evenly over the area of the section. The buckling of the heads of steel sheet piles during driving can occur in very hard driving. With vibratory hammers, the hammer clamps the steel sheet pile hydraulically, and thus there is no need for a helmet. Vibratory hammers are limited in their driving ability in certain soil profiles. Conventional hammers should thus be used in hard driving conditions or where the vibratory hammer fails to drive the sheet pile to the required depth. The toe of a sheet pile can be damaged when driving through a cobble or boulder layer, or when driving onto an uneven hard bedrock surface. Measures to limit the risk of damage include the use of sheet piles made from high tensile steel, reinforcement of the toe of the sheet pile by welding on additional strengthening plates, and reducing the energy per blow of the hammer. Most steel sheet pile sections have a small amount of play in the clutches, so a steel sheet pile wall can be installed on a radius. Special straight web sheet piles can be used to form a circular cofferdam. The interlocks between the sheet piles, if installed correctly, are watertight. Leakage on a clutch can occur if the correct driving procedure is not followed. A split clutch, in which the male clutch disengages from the female during driving, can also occur with poor driving technique. Steel sheet piles can be extracted from the ground once they have performed the function for which they were installed. Various types of equipment are available for extracting steel sheet piles, the most common being a vibratory extractor. Impact extractors powered by air or diesel are also available and are more suited when the extraction proves to be very difficult. A large crane in combination with an extractor, is used to provide the extracting force. PLATE 17.1.1 shows a steel sheet piling operation in progress. 213 PLATE 17.1.1 Vibrating Steel Sheet Piles 214 LIFE OF STEEL SHEET PILES The life of a permanent steel sheet pile wall depends very much on the corrosiveness of the surrounding environment and what protective measures are used. Unfortunately the coast of southern Africa is a highly corrosive environment and thus steel sheet piles are not often used for permanent marine structures with a long life expectancy. Where they are used for marine work, the section of sheet pile above the seabed level should be coated with a good protective layer. In the zone above the low water mark, the protection coating should be of the highest quality. In fresh unpolluted water the life of unprotected mild steel sheet piles is reported to be in excess of 80 years. Above the minimum water-level the protection given to the sheet pile should be matched to the corrosiveness of the environment. An examination of steel sheet piles extracted a considerable period after installation has shown that very little, if any, corrosion takes place below ground level in a land or even a marine environment. VARIATIONS IN INSTALLATION TECHNIQUE Composite Sheet Pile Walls The use of conventional sheet pile sections is limited by the size and moment of inertia of the sections. By reinforcing the sheet pile wall with additional steel sections which are welded to the sheet piles, the depth of soil that can be retained can be increased. Circular steel pipe piles in combination with sheet pile sections can be used to achieve greater depths and stiffness. FIGURE 17.1.2 shows how these two alternative variations look in plan. They are often used in marine construction. POTENTIAL PROBLEM AREAS Split Clutches A split clutch, in which the male clutch disengages from the female during driving, can occur with poor driving technique. This can be a serious problem which may necessitate the injection of chemical grouts or jet grouting behind the sheet pile wall to cut off the flow of water through the opening formed by the split clutch. Sealing on Uneven Bedrock Surface The sealing of a steel sheet pile wall on an uneven bedrock surface, where loose sand overlies the bedrock, is another potential problem area. A positive solution is to chemically grout or jet grout the sand overlying the rock, and behind the wall, before excavation takes place. Drawdown of Surrounding Soil The driving of steel sheet piles tends to compact the loose soil either side of the wall causing settlement which is referred to as drawdown. When driving steel sheet piles immediately alongside a structure with shallow foundations, this drawdown can cause foundation settlement resulting in cracking of the structure. The drawdown settlement can also result in overturning of adjacent boundary walls. Vibratory hammers are likely to cause more drawdown that conventional hammers. 215 Noise Pollution Driving steel sheet piles with a conventional hammer is an extremely noisy operation and should not be attempted in environmentally sensitive areas. Driving with a vibratory hammer is less noisy than with a conventional hammer. Encasing the leader of the piling rig with soundproofing materials has been successful in reducing the noise from the driving operation to an acceptable level. FIGURE 17.1.2 Composite Sheet Pile Wall Sections 17.1.2 PRECAST CONCRETE SHEET PILES An alternative to steel sheet piles is precast concrete sheets which interlock using the tongue and groove methodology. The one side of the toe is usually chamfered to force the concrete sheets to stay together during driving. Similar positive features given for steel sheet piles are applicable to precast concrete sheet piles and, in addition, they are produced locally and are significantly more cost effective. The other major considerations with concrete sheet piles are the reduced penetrability of the concrete section and the limited depth to which they can be installed. A typical cross-section of a concrete sheet pile is illustrated in FIGURE 17.1.3 and PLATE 17.1.2 shows a pit excavation supported by anchored concrete sheet piles. 216 GROUT SOCK GROUT SOCK 150 mm diameter ANCHOR SLEEVE FIGURE 17.1.3 A Typical Concrete Sheet Pile Cross-section PLATE 17.1.2 Pit Excavation Supported by Anchored Concrete Sheet Piles 217 17.1.3 STEEL SOLDIERS Over the years steel soldiers comprising single or twin H-sections, joists or channels, have been commonly used for cantilever walls, braced or tie-back walls, or as a structural element to transfer anchor forces onto the face of an excavation. .. .. .. POSITIVE FEATURES .. . Driven steel soldiers can be used in soft unstable soil profiles ‘Hand-over-hand’ method can be used where driving or pre-drilling is not possible Driven or pre-drilled soldiers can be incorporated into the permanent works Driven or pre-drilled soldiers can be designed to carry vertical loads Use of twin sections facilitates installation of anchor systems The profile of steel soldiers facilitates the installation of secondary support between soldiers OTHER CONSIDERATIONS Steel sections are relatively expensive Noise levels are high if sections are driven Limited stiffness for cantilever walls INSTALLATION TECHNIQUE Steel soldiers which are used either as a cantilever support system, or as an anchored structural element, are generally installed prior to the commencement of excavation. The steel sections can either be driven into the ground, or they can be placed in an auger drilled hole. With the latter technique the steel section is embedded in concrete below the excavation level, and a weak grout above the excavation level. The weak grout facilitates subsequent trimming during the main bulk excavation. The type of installation procedure that is adopted will generally depend on the nature of the ground conditions. In soft or loose unstable soil profiles driving is the preferred procedure. When deciding on which steel section to use, due consideration must be given to the possibility of the section buckling during driving if it is too slender. Other negative factors are noise problems associated with driving, the limited bearing area for passive resistance below excavation level, and possible problems with minimum depth requirements. The latter two factors are particularly significant where the steel soldiers are to be used as cantilever support systems, or where the soldiers are required to carry vertical loads. Installation by pre-drilling will be mainly suited to soil profiles that remain stable during the pre-drilling operation. For more specific details regarding installation techniques reference should be made to SECTION 7.4: STEEL H-PILES. 218 Steel soldiers are generally installed at 1.0 to 2.5 metre spacing. This implies that the soil being supported has some inherent stability and will arch between the soldiers. For certain soil profiles, stiff clays or dense residual soils, arching between the soldiers may provide sufficient secondary support for short and medium term stability. Soft clays and even some non-cohesive soils may remain stable for sufficient length of time to enable secondary support to be provided between the soldiers. Timber lagging or a lightly reinforced 50 to 100 mm thick gunite skin is generally used for secondary support. Where the steel soldiers are to be used with anchors it is common to use twin sections welded together, with a suitable gap between the sections, to allow for anchor installation. The anchor head is then stressed directly onto the twin soldier section and in this way the anchor forces are transferred to the excavation face without the use of expensive waling beam systems. Under certain circumstances it may not be practical to drive the soldiers, or to pre-drill a hole. Problems in this regard are usually associated with access difficulties, or ground formations, where driving or pre-drilling is not possible. Under these circumstances prestressed anchors are usually used as a tie-back system to provide overall stability, with steel soldiers being used as a structural element to transfer the anchor forces onto the excavated face. Soldiers and anchors are then installed as the excavation proceeds, adopting the ‘hand-over-hand’ method. This requires that the excavation be formed in shallow stable benches with the anchors and soldiers for each bench being installed as soon as possible after excavation. After stressing of the anchors the next bench is excavated and the procedure repeated with the soldiers being welded together to form a continuous vertical structural element. It may be necessary to provide temporary stability by forming berms at a suitable slope angle with the soldiers and anchors being installed in slots formed through the berms. Once all the anchors are stressed, the berm can then be completely removed and the procedure repeated at the next level. This procedure is only suitable for anchored walls and cannot be used for cantilever walls. Experience with soldier installation procedures for anchored walls has shown that in instances where adjacent ground movements are to be kept to a minimum, driving or pre-drilling of the soldiers is preferable to using the ‘hand-over-hand’ construction technique. A further advantage for deep excavations supported with multiple anchor systems is that installation by driving, or pre-drilling, can also speed up overall excavation and lateral support construction procedures. In particular, the cantilever support, provided by the driven or pre-drilled soldiers, allows excavation to proceed rapidly to a depth of 2.0 to 3.0 metres before installation of the first row of anchors. The stiffness provided by steel soldiers also allows unrestricted excavation to be carried out between rows of anchors, thus avoiding time delays associated with berm and slot type installation procedures. PLATE 17.1.3 shows a typical steel soldier pile installation with anchor tie-backs for the support of an excavation. 219 STRUCTURAL STEEL DESIGN CONSIDERATIONS In certain instances the steel soldiers may be incorporated as part of the final structure. This will generally require some form of corrosion protection for the soldiers. Secondary support between the soldiers is then usually provided by a suitably designed structural gunite skin spanning between the soldiers. Where the soldiers are driven or pre-drilled they can also be designed to carry vertical loads. This will however require suitable penetration below the excavation level. In the design of the steel soldiers the maximum axial, bending and shear stresses during all phases of construction should be considered and the design should then be carried out in accordance with the appropriate Code of Practice. Plastic design methods are used under certain temporary conditions. Where the steel soldiers are to be incorporated as part of the final structure, the final structural loading and support conditions may be the controlling factors in the design. PLATE 17.1.3 Steel Soldiers with Anchor Tie-backs and Timber Lagging 220 17.1.4 CONCRETE SOLDIER PILES As for driven or pre-drilled steel soldiers, concrete soldier piles can be installed to provide support either as cantilever walls, braced or tie-back walls, or to act as structural elements against which anchor forces are mobilised. .. .. . . . POSITIVE FEATURES Good flexibility in relation to available pile types and diameters Large diameter piles can provide increased stiffness for certain applications Easily incorporated into permanent works Can be designed to carry vertical loads No corrosion protection required OTHER CONSIDERATIONS A waling beam system may be required for anchor installation where anchors are not installed through the piles Large diameter piles will encroach into available space INSTALLATION TECHNIQUE Concrete perimeter piles are usually installed prior to commencement of the excavation at between 1.0 and 2.5 metre spacing. Usually a 0.3 to 2.0 metre gap is left between piles. This implies that this system is most suited to relatively stable soil profiles. The following pile types are usually the most suitable for use as concrete soldier piles: .. .. Auger Piles CFA Piles Forum Bored Piles Franki Piles Reference should be made to SECTION 7.0: TECHNICAL DETAILS OF PILING SYSTEMS to obtain specific details for these pile types, the soil conditions for which they are most suited, and installation methods. Where the concrete soldier piles are to be used as a cantilever support system, or the piles are required to carry vertical loads, it is necessary to ensure that sufficient embedment of the pile shaft is obtained below the bottom of the excavation. Whether this can be achieved needs to be carefully evaluated in relation to the soil profile and the type of pile being used. For anchored walls the anchors can be installed between piles, with loads being transferred to the piles with a suitable waling beam system. Alternatively, installation of anchors can be done by drilling through the piles. This may be a less expensive option since waling beams are then not required. The benefit of using walers is that the amount of drilling is halved, but the soil must be competent as the anchor capacity is doubled, otherwise a longer fixed length may be required to generate the design anchor force. 221 A further alternative is to use twin channels, joists or H-beam sections, as pile reinforcement. These sections are welded together with a suitable gap, and after excavation to a selected level, the front concrete of the pile is broken away to expose the steel sections. The anchors are then installed through the sections. This procedure is generally only suitable where auger piles are used in a stable soil profile, although special techniques can also be adopted to use this procedure with Franki, Forum Bored and CFA piles. As for driven or pre-drilled steel soldier piles experience has shown that for deep excavations with multiple anchor systems, the use of concrete soldier piles can significantly reduce adjacent ground movements and speed up overall excavation and lateral support construction procedures. The greater stiffness associated with concrete soldier piles is more effective than steel soldier piles in this regard. As indicated the use of concrete soldier piles requires that the soil is stable enough to arch between the piles for at least a short time period. For certain soil profiles, stiff clays or dense residual soils, arching between the soldiers may provide sufficient secondary support for short and medium term stability. In less stable soil profiles, soft clays or loose soils etc, secondary support in the form of a lightly reinforced 50 to 100 mm gunite skin is provided between the piles. PLATE 17.1.4 shows typical concrete soldier piles with anchor tie-backs. STRUCTURAL DESIGN CONSIDERATIONS In most cases the concrete soldier piles are incorporated as part of the final structure. Secondary support between the piles is then usually provided by a suitably designed structural gunite skin spanning between the piles. Where vertical loads are to be carried, it is necessary to ensure that the piles are installed to a sufficient depth below final excavation level. In the design of the piles the maximum axial, bending and shear stresses during all phases of construction should be considered and the design should then be carried out in accordance with the appropriate Code of Practice. Where the concrete piles are to be incorporated as part of the final structure, the final structural loading and support conditions may be the controlling factors in the design. 222 PLATE 17.1.4 Concrete Soldier Piles with Anchor Tie-backs 223 17.1.5 CONTIGUOUS AND SECANT PILE WALLS A contiguous pile wall is simply a row of concrete soldier piles installed so that each pile is in contact, or near contact, with piles on either side of it. The positive features and other considerations given in SECTIONS 17.3 for concrete soldier piles are therefore also applicable to this system. There is however a relatively large increase in cost for this system in comparison with concrete soldier piles. The technique is therefore generally only used in unstable soil profiles (soft saturated clays or sands) which do not have an ability to arch between adjacent piles. In certain instances the increased overall stiffness of a contiguous pile wall as compared to steel soldiers, or concrete soldier piles, may be a significant factor in reducing the risk of adjacent ground movements. Under these circumstances a contiguous pile wall may be preferred even when relatively stable soil conditions occur. This wall is not watertight, unless specific steps are taken to achieve this, so leaching and/or piping of non-cohesive soils through gaps between the piles can be a problem below the watertable. A special case of a contiguous pile wall is called a secant wall. Here the piling is carried out in a sequence in which subsequent piles are cut into the previously installed piles thereby affecting a seal between the units. INSTALLATION TECHNIQUE A contiguous pile wall is generally selected in preference to steel soldiers, or concrete soldier piles, due to the unstable nature of the soil profile or the presence of a high water-table. The pile type selected for this system should therefore be capable of dealing with these conditions. In general terms, bored pile systems such as augured underslurry, CFA, Forum bored or temporary cased auger piles are preferred. Franki piles can be used where there is a particular reason for requiring a driven pile type. Reference should be made to SECTION 7.0: TECHNICAL DETAILS OF PILING SYSTEM to obtain specific details for these pile types. Whichever pile type is used the installation sequence is to first install odd numbered piles and when these have set, the even numbered piles. The spacing between the odd numbered piles has to be carefully chosen so as to allow unobstructed installation of the even numbered piles, while at the same time limiting the gap between the piles. Special measures must be taken to improve the plan and verticality tolerances of pile installation. A plan tolerance of + 25 mm and a verticality tolerance of 1:100 should be possible in soil profiles free of obstructions. The following techniques can be used to improve water tightness of a contiguous wall: .. . The gap between the piles can be drilled and grouted in-situ; Odd numbered piles can be installed slightly back of the centre line and cast in a cement bentonite mix. Even numbered piles are then drilled so that they cut into the bentonite cement columns thus forming a seal. This solution can only be used with an auger/bored pile system (auger underslurry or CFA piles); The contiguous piles can be formed in a secant wall manner where even numbered piles cut a secant into odd numbered piles thus forming a seal. This solution can only be used with an auger/bored pile system (auger underslurry or CFA piles) or an oscillator pile system. 224 Typical details used for contiguous and secant pile walls are shown schematically in FIGURE 17.1.4. A contiguous auger pile with anchor tie-backs is shown in PLATE 17.1.5. A typical guide frame for the construction of a secant wall on the Vaal Dam, Gauteng, South Africa is shown in PLATE 17.1.6. STRUCTURAL DESIGN CONSIDERATIONS Due to the expense associated with the installation of a concrete contiguous pile wall the system invariably forms part of the permanent works. In the design of the piles the maximum axial, shear and other stresses during all phases of construction, as well as those due to final structural loading, should be taken into consideration. Design should then be carried out in accordance with the appropriate Code of Practice. CONTIGUOUS PILE WALL SECANT PILE WALL FIGURE 17.1.4 Typical Contiguous and Secant Pile Wall Details 225 PLATE 17.1.5 Contiguous CFA Piles with Anchor Tie-backs PLATE 17.1.6 Guide Frame for Installation of Secant Pile Wall 226 17.1.6 DIAPHRAGM WALLS A diaphragm wall is a reinforced concrete wall constructed in the ground using underslurry techniques. Walls with widths of between 300 and 1500 mm can be formed in this way to depths in excess of 60 metres. .. .. .. . .. .. POSITIVE FEATURES Walls can be installed to considerable depths Walls with substantial thickness can be formed The system is flexible in plan layout The wall can easily be incorporated into the permanent works The wall, or certain sections, can be designed to carry vertical load Basement construction time can be reduced Economical, positive solution for large, deep basements in saturated and unstable soil profiles Noise levels limited to engine noise only No vibration during installation OTHER CONSIDERATIONS Not normally economical for small, shallow basements The system needs a relatively large site area Under certain conditions diaphragm walls may be used as cantilever, braced or tie-back walls in preference to the systems discussed in the previous sections. Diaphragm walls are necessary: . . . In very unstable soil profiles below the water-table where continuous support and watertight conditions are required to prevent mud flows, piping and erosion of the soils; Where construction time is important and the use of a diaphragm wall can shorten the programme; In conditions where deeper than normal cantilever support may be required. These conditions could occur where the wall is to act only as a cantilever, or where a very deep initial excavation is required before the first braced or tie-back supports can be installed. INSTALLATION TECHNIQUE A diaphragm wall is constructed in a series of separate adjoining panels with each panel keyed into the adjacent panels. The reason for constructing the wall in panels is the limitation on the length of excavation that will remain stable under a head of bentonite. The minimum panel length is dictated by the size of the excavation grab. The stability of the trench side-walls, the plan shape of the panel, and problems associated with the flow of concrete during concreting, control the maximum length of panels. For anchored walls, steel sleeves are attached to the reinforcing cage to allow for the anchors to be installed through the wall panels. Typical details for diaphragm wall construction are given in FIGURE 17.1.5. The excavation of a diaphragm wall panel is shown in PLATE 17.1.7 and the insertion of the reinforcement cage is illustrated in PLATE 17.1.8. 227 FIGURE 17.1.5 Typical Details for Diaphragm Wall Construction 228 The correct layout of panels is an essential part of a diaphragm wall design. The layout must allow effective excavation and concreting of the panels and must take into consideration factors such as the size of the grab, the plan shape of the wall, the dimensions of steel cages and site logistics. Conventional diaphragm wall construction commences with the construction of a pair of guide walls, one on either side of the main wall. These guide walls are usually formed in concrete and are about 1.2 metres deep. They provide guidance to the excavating grab, support to the side-walls at the surface as well as a convenient platform for controlling the concreting operation. The internal guide wall is removed during the main excavation. The construction of the diaphragm wall may be started in more than one place with the initial panels known as starter panels. A panel constructed adjacent to a another panel is termed an intermediate panel, and one constructed between two existing panels, a closure panel. When concreting a panel, the ends of the panel have to be formed so that the one keys into the other. To achieve this a shaped steel form is placed in position at one or both ends of the panel prior to concreting. These forms are known as ‘stop-ends’. Once the concrete has taken its initial set, the stop-ends are gradually withdrawn leaving the end face of the concrete panel with the desired key. Starter panels need to have two stop-ends, intermediate panels only one and closure panels have no stopends. Excavation is carried out under a head of bentonite slurry by means of a grab suspended from a crane in a similar manner to that described for underslurry barettes under SECTION 7.7 UNDERSLURRY PILES. The breadth of the grab is generally between 2.0 and 3.0 metres and the widths between 300 and 1500 mm. Panels are normally 2 to 5 metres in length and thus require more than one pass of the grab. The trench is kept topped up with bentonite slurry during both the excavating and concreting operations. Once excavation of the panel is complete, the bentonite in the trench is processed to reduce the density and adjust the pH. The required stop-ends are placed in position followed by the steel reinforcing cage and the tremie pipe. A pump is located in the trench for pumping the slurry back to storage tanks during concreting. The concrete operation is carried out using normal tremie concrete techniques and a concrete mix with a 200 mm slump. The level of the concrete should be cast at least 750 mm above the required cut-off level. Once the concrete has taken its initial set any stop-ends are gradually extracted leaving the end-face of the concrete panel with the desired key. The panels are excavated and concreted according to the planned sequence until the full diaphragm wall is complete. Where the diaphragm wall is to be used as a cantilever support system or where the wall is required to carry vertical loads, it is necessary to ensure that there is sufficient embedment below the bottom of the excavation. In many instances the wall is also designed to act as a groundwater barrier and a minimum embedment depth is then usually required for this purpose. A deep basement excavation using a diaphragm wall with tie-back anchors, in Tanzania, is shown in PLATE 17.1.9. 229 PLATE 17.1.7 Diaphragm Wall Panel Being Excavated PLATE 17.1.8 Insertion of Diaphragm Wall Panel Reinforcement 230 PLATE 17.1.9 Diaphragm Wall with Anchor Tie-backs 231 17.2 EMBEDDED WALL SUPPORT SYSTEMS 17.2.1 PROP SUPPORTS A bracing system comprising prop supports can be used to support the various wall systems described in SECTION 17.1.1 to 17.1.5. A system of horizontal struts to provide cross-bracing is common in trench excavations and other excavations of limited width. Inclined raking struts are used to support walls where the distance is too great for horizontal struts. These systems are illustrated in FIGURE 17.2.1. The struts can be made of steel, concrete or timber depending on the loads to be carried. .. . .. POSITIVE FEATURES No encroachment into adjacent property No specialist expertise is required for installation of the system Simple and quick construction procedure for smaller excavations OTHER CONSIDERATIONS Restricts access and construction working space Specialist procedures such as pre-stressing of struts may be required to limit adjacent ground movement on larger excavations INSTALLATION TECHNIQUE The first phase of construction comprises the installation of one of the wall systems described in SECTION 17.1.1 to 17.1.5. With horizontal struts, excavation usually proceeds until the first level of support is reached. If necessary, a horizontal waler is attached to the wall and the strut is tightly attached to the waler, or directly onto the wall. The excavation then proceeds to the next level. Horizontal and vertical spacing of the struts is a function of the support forces required and the type of wall system that has been used. To limit deflections, attention must be paid to construction and design details regarding stiffness of the struts and the connection between the struts and the walers or wall. Struts can be prestressed in instances where it is necessary to specifically restrict adjacent ground movement. Pre-stressing is usually carried out to about fifty percent of the anticipated working load. Inclined raking struts are used to support walls where the distance is too great for horizontal strut support. The usual installation procedure is to phase the excavation so that a berm is left behind to support the wall element. The foundation system for the raking strut is then installed and if necessary a horizontal waler is attached to the wall. The wall is then braced by installing the raking strut with suitable connections to the foundation and the wall. This is illustrated in FIGURE 17.2.1. If required, the berm is lowered and the next level of raking struts installed. This will be dependent on the type and height of wall to be supported. Horizontal spacing of the struts will be a function of the support forces required and the type of wall being supported. To control deflections, attention must be paid to the construction and design details regarding stiffness of the struts and the connection between the struts and the wall. Particular attention must also be paid to the design of the foundation system for the raking strut. Under certain circumstances it may be necessary to transfer large horizontal forces into the foundation system. In poor soil conditions a piled foundation system may be required to support the raking struts. 232 VARIATIONS IN INSTALLATION TECHNIQUE Raking piles can be installed to act as props as illustrated in FIGURE 17.2.1. This procedure can speed up basement excavation since it is not necessary to leave berms and wait for the raking struts to be installed before removal of the berms. HORIZONTAL STRUTS RAKING STRUTS FIGURE 17.2.1 Typical Prop Support Systems 233 17.2.2 POST-STRESSED ANCHORS Post-stressed anchors are frequently used as tie-backs for lateral support of deep excavations. Anchors can be used as a tie-back for the wall systems described in SECTIONS 17.1.1 to 17.1.5. An anchor system can also provide primary support, with the wall providing secondary support as well as acting as a structural element which transfers the anchor forces onto the excavated face. .. .. . .. .. . POSITIVE FEATURES Post-stressing will assist in limiting adjacent ground movements Suitable for most soil and rock types Usually the only procedure that can stabilise deep-seated failures High stabilising forces can be exerted with multiple anchor systems Provides an unobstructed area for basement construction OTHER CONSIDERATIONS Anchors may encroach into neighbouring property Considerable expertise required for drilled anchor installation A structural element is required to transfer anchor forces onto the excavated face The corrosion protection of permanent anchors is expensive De-stressing of temporary anchors must be allowed for ANCHOR TYPES The various components used for post-stressed anchors are shown in FIGURE 17.2.2. The following components and construction procedures are used by Franki Africa for post-stressed anchors: .. Tendons comprising threaded bars or strand; .. The free anchorage is normally formed by suitably sheathing of the tendon; The procedures used to form the fixed anchorages are shown in FIGURE 17.2.3. Deadman fixed anchorages, Type A, are used when the anchorage position is close to the ground surface. A straight shafted gravity grouted anchor, Type B, is used in very stiff/very dense soils or in rock. A re-injectable anchor using a high pressure grouting procedure, Type C, is the most common system adopted since this procedure is suitable for most soil and rock types. The fixed anchorage is formed by hydro fracturing and / or compaction of the surrounding soil or rock. The high pressure grouting is generally carried out using a tube-a-manchette system; The anchor head consists of a device which can post-stress the tendon against a suitable bearing plate and lock the tendon at the required load. Various anchor wedge systems are available for strand anchors. A nut is used for threaded bars. Anchors can be used for temporary or permanent applications. In general terms a temporary anchor will require little or no corrosion protection, whereas specific attention will have to be paid to corrosion protection for permanent anchors. The terms temporary and permanent are dependent on a number of factors (service life, corrosion environment, consequences of failure etc) and it is therefore not possible to provide an exact definition. The classification system given in the SAICE Geotechnical Division Code of Practice (1989) Lateral Support in Surface Excavations is recommended as a classification system in this regard. This classification system is reproduced as TABLE 17.2.1. 234 MANUFACTURE AND TRANSPORT Where possible anchor capacity, fixed lengths and free lengths for a project should be pre-determined prior to commencing site operations. This enables the anchors to be pre-assembled off site with obvious quality control benefits, especially in the case of permanent anchors where attention to detail is essential to ensure reliable corrosion protection measures. Transport and on site storage are equally important to ensure the anchor integrity prior to installation. FIGURE 17.2.2 Components for Post-stressed Anchors 235 FIGURE 17.2.3 Procedures to Form Fixed Anchorages 236 TABLE 17.2.1 Corrosion Protection Guide. After SAICE Geotechnical Division Code of Practice (1989) Lateral Support in Surface Excavations CONDITIONS PERTAINING TO SERVICE LIFE RESULT OF FAILURE No. Temporary Up to six months Six months to two years Permanent Over CORROSION ENVIRONMENT No. MATERIAL AROUND FREE LENGTH No. LATER ACCESS No. 0 Not serious 0 Non-corrosive 0 Rock A 1 Serious 1 Corrosive 1 Sand B 2 Catastrophic 2 Very corrosive 2 Clay C No. Needed / Re-stressable anchor X Not needed / Non-stressable anchor Y two years TOTAL THE SCORES IN THE COLUMNS AND REFER TO THE PROTECTION ALTERNATIVE TYPE CODE BELOW PROTECTION ALTERNATIVES FIXED ANCHOR LENGTH FREE LENGTH HEAD AND UPPER TENDON Type Code Treatment Type Code Treatment Type Code Treatment 0 - 1, ABC, XY 0-6, AY Tendon grouted bare 0, AXY Bare tendon empty hole 0 - 1, ABC, XY Bare head 2 - 3, ABC, XY Tendon epoxy coated 1 - 6, AY Bare tendon grouted in hole 2, ABC, XY Painted head 4 - 5, ABC, XY Tendon grouted into corrugated sheath while grouting anchor 0, BC, XY 1, A, XY Sheathed tendon in empty hole 3 - 4, ABC, X Epoxy coated head 1, BC, XY Sheathed tendon in grouted hole 2 - 6, ABC, X Head covered with cap filled with grease 2, ABC, XY Greased and sheathed tendon in empty hole 5 - 6, ABC, Y Head concreted or grouted into box-out 3 - 6, ABC, XY Greased and sheathed tendon in grouted hole 5 - 6, ABC, XY Greased and sheathed tendon in common sheath in grouted hole 5 - 6, ABC, XY Tendon pre-grouted in corrugated sheath FOR EXAMPLE: Temporary anchors up to four months in very corrosive conditions in sand with serious consequences of failure and tested at one month intervals. SCORE: 0 + 1 + 2 = 3BX FIXED ANCHOR LENGTH: Tendon epoxy coated before grouting into holes FREE LENGTH: Tendon greased and sheathed before grouting into hole HEAD AND UPPER TENDON: Epoxy coated 237 INSTALLATION TECHNIQUE Pre-assembled anchors are installed into pre-drilled holes drilled by rotary or percussion techniques. It is important that holes are thoroughly cleaned to ensure efficient bond between the anchor grout and in-situ rock or soil. For this reason it is normal to drill holes 0.3 to 0.5 metres deeper than the actual total anchor lengths to provide a sump for drilling debris. Temporary casings can be used during the drilling operation in instances where stability of the drilled hole is of concern. PLATE 17.2.1 shows a crawler rig installing temporary anchors. Grouting procedures are adapted to suit site conditions and the procedures used must ensure full grout cover between the anchor and the surrounding soil or rock, without the risk of grout contamination. Where necessary temporary or permanent casings are used to ensure that this is achieved. Normal practice is to home the anchor system into a fully grouted hole. High pressure grouting of the fixed anchorage using a tube-a-manchette system has become a common procedure for most anchor installations in soils and soft or fractured rocks. The tube-a-manchette forms part of the overall anchor assembly. The stressing and testing of anchors is one of the most important phases of any anchor installation. During this phase each anchor is tested to a specified percentage above its design load and the performance of the anchor recorded on a stress / strain graph which is compared to predicted performance. In addition, creep relaxation is recorded which allows a prediction of the long-term behaviour of each individual anchor. The correct stressing and testing of anchors is an extremely important aspect in the overall installation procedure. By adopting the correct procedure, each anchor on a project is tested to a specified percentage above the required working loads. This effectively confirms the suitability of the entire anchor assembly with the exception of the corrosion protection aspects. The standard stressing procedures for different anchor types and applications are well documented in the SAICE Geotechnical Division Code of Practice (1989) Lateral Support in Surface Excavations and reference should be made to this document in this regard. ANCHOR CAPACITY The procedures used for anchor installations are such that anchor capacity is usually controlled by the permissible tendon stress which should be determined in accordance with the recommendations given in the SAICE Geotechnical Division Code of Practice (1989) for Lateral Support of Surface Excavations. This requires that the stress of the tendon when locked off should not exceed 70% of the characteristic strength and that during stressing and testing the stress should not exceed 80% of the characteristic strength. This latter criteria is usually the controlling factor for both temporary and permanent anchors which are usually stressed to 125% and 150% of working load respectively. Typical anchor working loads using this criteria are presented in TABLE 17.2.2. 238 TABLE 17.2.2 Typical Anchor Working Loads Tendon System Typical Working Load (kN) E.A (kN) Temporary Permanent Axial stiffness 20 mm 105 85 64717 25 mm 165 135 101120 1 strand (141 mm2) 168 140 30900 2 strands (282 mm2) 336 280 61800 3 strands (423 mm2) 4 strands (564 mm2) 504 672 420 560 92700 123600 5 strands (605 mm2) 840 700 154500 Threaded Bars Characteristic Strength 525 MPa Strand 15.2 mm diameter Characteristic Breaking Load 265 kN PLATE 17.2.1 Crawler Rig Installing Temporary Anchors 239 17.2.3 THREADED HOLLOW-BAR ANCHORS Franki Africa are able to provide innovative and versatile soil anchor solutions in soft or very difficult ground conditions, where hole collapse renders conventional methods impractical. This is done by installing threaded hollow-bar anchors which are pressuregrouted as the bars are installed. Small holes at the tip of the disposable bit allow grout to be injected at pressure, compacting and improving the soil to the extent that ultimate bond stresses of between 100 kPa and 200kPa, depending on soil type, can be guaranteed. The angle of inclination of the bars can be changed so that the bars can also be used in compression or as a reticulated system. Typical ultimate loads of between 220 kN and 1950 kN can be accommodated and grouting undertaken with pressures up to 6 MPa (60 bars) for high capacity hollowthreaded bars with diameters of between 30 and 130 mm. Details of the Franki Titan hollow-bar system are given in TABLE 17.2.3. TABLE 17.2.3 Titan Anchor Details ANCHOR TYPE Unit Titan Titan Titan Titan Titan Titan Titan Titan Titan Titan Titan Titan 30/16 30/14 30/11 40/20 40/16 52/26 73/56 73/53 73/45 103/78 103/51 130/60 * * * 40 40 52 73 73 73 103 103 130 11 20 16 26 56 53 45 78 51 60 260 320 539 660 929 1194 1160 1630 2282 3460 7940 220 260 430 525 730 785 970 1180 1800 2750 5250 140 165 205 345 425 595 765 745 1045 1460 1760 5080 N/mm2 470 610 580 590 590 550 550 590 510 570 500 550 Crosssection (A) mm2 382 395 446 726 879 1337 1414 1631 2260 3146 5501 9540 Weight kg/m 2.7 2.9 3.3 5.6 7.0 10.0 11.1 12.3 17.8 24.9 43.4 75.0 left left left left left left or right right right right right right right 3/4 3/4 2/3/4 3 2/3 3 3 3 3 3 Nominal Outside Diam mm 30 30 30 Nominal Inside Diam mm 16 14 Ultimate Load kN 220 Yield Point kN 180 Working Load (Temp) kN Yield Stress T0.2 Thread Left/ Right Hand Lengths m 3 6.25 * Most commonly available bar sizes NOTE: The working load for permanent anchorages requires a 16% reduction of the temporary working load 240 17.3 REINFORCED SOILS 17.3.1 GEONAILS GeoNail is the registered trade mark for Franki Africa's in-situ soil nailing system. The GeoNail system is a method whereby slopes or excavations are stabilised by reinforcing the soil in-situ with closely spaced tensile inclusions. In the majority of cases these inclusions are fully bonded high yield steel bars which are introduced into the soil mass to produce a zone of reinforced ground. In a conceptual sense this zone can be considered to act as a homogenous and resistant unit to support the unreinforced soil behind, in a manner similar to a gravity retaining wall. This system is most suited to relatively steep slopes or excavations. The GeoNails are installed sub-horizontally so that the shearing resistance along a potential failure plane is improved by the reinforcing elements acting in tension. Some form of secondary support, usually in the form of a lightly reinforced gunite skin, is required. The main function of the gunite skin is to prevent local ravelling and deterioration, rather than to provide primary structural support. .. .. . .. .. POSITIVE FEATURES Extremely cost effective system under suitable conditions Can be installed faster than most other comparable lateral support systems No load transfer onto face of excavation being supported Easily adaptable to changes in site conditions Relatively simple and inexpensive procedures can be used to provide corrosion protection of permanent GeoNail installations OTHER CONSIDERATIONS Passive system which requires some movement to mobilise stabilising forces Generally not suitable for deep-seated failure surfaces Not suitable in non-cohesive strata below the water-table Usually not economical in soft clays or boulder formations The GeoNail system can be used for both temporary and permanent support. As indicated previously, it is often difficult to define ‘temporary’ and ‘permanent’ support requirements and guidelines in this regard have been taken from the SAICE Geotechnical Division Code of Practice (1989) Lateral Support in Surface Excavations, TABLE 17.2.1. With GeoNails the first defence against corrosion is to ensure that the high yield steel bars have adequate grout cover. This is generally sufficient for most temporary applications. In permanent applications consideration must be given to the fact that the reinforcements act in tension and the influence of cracks within the grout need to be evaluated. For permanent applications other corrosion protection measures such as galvanising the reinforcement and / or the use of a protection sheath can also be incorporated into the GeoNail system. For permanent applications specific attention also needs to be paid to the reinforced gunite skin between GeoNails. In general terms the gunite skin is made thicker than for temporary applications (this ensures sufficient cover to the reinforcement) and under specific highly corrosive conditions the reinforcement can also be galvanised. Where the GeoNail system is to form a permanent support, the design engineer should discuss the overall problem with Franki Africa to decide on suitable corrosion protection measures. 241 As for all internally stabilised techniques the GeoNail system can be considered to provide stability in a ‘passive’ mode since some movement is required to mobilise the stabilising forces. Although this can be problematic in certain applications where it is necessary to limit deformations, there are also certain advantages to the use of a passive system. The main advantage is that no large load transfer elements are required on the face of the excavation or slope being supported. This makes the system ideal for stabilisation of gravity retaining walls made either of masonry, brick or poor quality concrete. The use of a passive reinforcement system will ensure overall stability of the soil mass behind the wall and reduce the earth forces applied to the wall. This allows the wall to be retained and repaired. INSTALLATION TECHNIQUE .. . . The basic installation procedure is illustrated in FIGURE 17.3.1 and is as follows: . Excavation is carried out in benches of limited height, 1.0 to 2.0 metres The GeoNails are installed as soon as possible after excavation. Installation comprises the drilling of a sub-horizontal hole, 75 to 120 mm in diameter using rotary percussion techniques, fully grouting the hole and then homing the reinforcing element A vertical GeoDrain system is usually installed at regular intervals between the GeoNails. This system comprises a geofabric wrapped geonet A reinforced gunite skin is applied. For temporary applications the overall gunite skin is usually 75 to 100 mm thick and nominally reinforced with a suitable mesh. For permanent application the gunite is 100 mm or thicker and reinforced to suit the specific application. Usually a flash coat of gunite, 25 to 50 mm thick, is applied directly onto the excavated face. The reinforcement is then installed before completing the guniting operation. As part of this phase of the operation a face-plate and nut are installed on the reinforcement and the GeoNail is nominally tensioned using a torque wrench. The next bench is excavated and the process described above repeated PLATES 17.3.1 and 17.3.2 illustrate some of the procedures associated with the installation of GeoNails. . . . VARIATIONS IN INSTALLATION TECHNIQUE Where short-term ravelling of the excavated benches can occur, the gunite skin is applied before the installation of the GeoNails to provide additional stability Excavation and installation can be carried out in slots where there is concern about overall stability of the excavated faces An application that is becoming more common in basement excavations is to use a GeoNail system for temporary support and the gunite wall as the permanent basement wall. In this case the gunite wall is usually 150 to 250 mm thick and designed to span between suspended floors. Techniques are adopted during the guniting operation to leave box-outs in the wall as a connection for the suspended floors, or the floor slab can be designed to simply butt up against the wall as a permanent prop. The gunite wall can also be designed to carry vertical loads. This however requires that a suitable foundation system be constructed below the wall after completion of the basement excavation. 242 . . . Rotary percussion drilling of the soil nail can be carried out with a temporary casing in instances where the stability of the drilled hole is of concern. Alternatively, hollow-bar anchors can be employed in unstable ground conditions The reinforcement elements can be driven and hammered into the ground rather than drilling and grouting Post grouting techniques using a tube-a-manchette system can be used to improve bond values between the grout and soil / rock FIGURE 17.3.1 Installation Procedures for GeoNails 243 . . . . . POTENTIAL PROBLEM AREAS The passive nature of the GeoNail system may be problematic where deformations are to be limited and/or high surcharge loads occur; The use of a GeoNail system is often not suitable in soil profiles with non-cohesive strata below the water-table. With this type of profile excavated benches are generally unstable. Specialised techniques are usually also required to ensure stability of closely spaced drilled holes for GeoNail installation. The additional costs associated with these aspects often makes other lateral support systems more attractive; In soft clays the very close nail spacing required may make GeoNails uneconomical compared to other systems; The overall support mechanism associated with a reinforced soil system may make GeoNails uneconomical where potential failure could occur along deep-seated shear planes; Drilling in boulder formations is a problem for any lateral support system. The additional length of drilling that is often required for a closely spaced GeoNail system may be a major disadvantage in boulder formations. GEONAIL CAPACITIES The available capacity of a GeoNail is controlled by the tensile capacity of the high yield steel bars and the GeoNail pull-out resistance which is controlled by the bond achieved between the grout and surrounding soil / rock. Threaded bars are usually used in GeoNail applications. These threaded bars have a characteristic stress of 525 MPa and are available in 20 and 25 mm diameters. In the calculation of working loads a permissible stress of 70% of the characteristic strength is used which gives working load values of 120 kN and 180 kN for the 20 and 25 mm diameter bars respectively. The available capacity is usually controlled by pull-out resistance. Assumptions are usually made in this regard at design stage and it is important to check these assumptions during installation. Working GeoNails are fully grouted and as a result tests to check pull-out resistance of working nails are not practical. For this reason test GeoNails are installed using reduced length to check pull-out resistance. These tests are usually designed to allow tests to be carried out to failure, or to a value of at least twice the assumed design value. 244 PLATE 17.3.1 GeoNail Installation PLATE 17.3.2 GeoNail Wall Nearing Completion 245 17.3.2 RETICULATED MICROPILES Reticulated micropiles are closely spaced small diameter (75 to 300 mm) piles that are installed into the ground vertically or at a steep rake. The piles are suitably reinforced to resist tensile and shearing forces. The overall objective is to form a stable block of reinforced soil which supports the adjacent unreinforced soil by acting as a gravity retaining structure. Some applications for the system are illustrated in FIGURE 17.3.2. Reticulated micropiles have not been widely used in Africa. It is however a system that has been widely used in certain European countries, by Lizzi (1982), and to a lesser extent in North and South America, Murray (1980) and Dash and Jovino (1980). From the applications illustrated in FIGURE 17.3.2 it is apparent that it is a unique solution that can be used to solve a specific problem. .. .. .. POSITIVE FEATURES No encroachment onto adjacent property Suitable for use in most soil and rock types Compact equipment allowing access to restricted areas and remote sites Unique solution that can be used for a specific problem OTHER CONSIDERATIONS Relatively expensive system due to high density of reinforcement that is required Passive systems require some movement to mobilise stabilising forces A typical reticulated micropile system comprises three rows of piles. The front row is normally installed vertically and in the remaining two rows alternate piles are installed vertically and at a rake, or all piles in these two rows are installed at a rake. After installation of the piles, a reinforced concrete capping beam is constructed to provide a rigid connection between the rows of piles. The reticulated micropile system can be used for both temporary and permanent support. As indicated previously it is often difficult to define ‘temporary’ and ‘permanent’ support requirements and guidelines in this regard have been taken from the SAICE Geotechnical Division Code of Practice (1989) Lateral Support in Surface Excavations, TABLE 17.2.1. With reticulated micropiles the first defence against corrosion is to ensure sufficient cover to the pile reinforcement. This is generally sufficient for most temporary applications. In permanent applications consideration must be given to the fact that the micropiles can act in tension and the possible influence of cracks within the grout need to be taken into consideration. For permanent applications the reinforcement can be galvanised. Where reticulated micropiles are to be used for permanent applications it is recommended that the design engineer discuss the overall problem with Franki Africa to decide on suitable corrosion protection measures for a specific project. INSTALLATION TECHNIQUE The micropiles are installed by forming a hole to the required diameter depth and rake using rotary percussion techniques. After drilling, the hole is blown clean using compressed air. The hole is then fully grouted and the pile reinforcement is installed into the grouted hole. The pile reinforcement usually comprises a single high tensile bar. 246 If required a reinforcing cage made up of 10 or 12 mm high tensile bars or a mild steel tube of 3 to 5 mm wall thickness can be used. The reinforced concrete capping beam is cast after completion of the micropile installation. . . . . VARIATIONS IN INSTALLATION TECHNIQUE Rotary percussion drilling can be carried out with temporary casing in instances where the stability of the drilled hole is of concern In soft or loose unstable soil conditions the micropiles can be installed by driving a thin-walled permanent casing Post grouting techniques using a tube-a-manchette system can be used to improve bond values between the grout and soil / rock Hollow-bar micropiles can be installed in unstable conditions FIGURE 17.3.2 Typical Applications for Reticulated Micropiles, Lizzi (1982) 247 17.3.3 SOIL DOWELLING Soil dowelling, Gudehaus (1983), is a technique whereby the shearing resistance on a well defined failure plane in weak soils is mechanically stabilised by the installation of large diameter piles which combine a large shear surface area with a high bending stiffness. A typical application for soil dowelling is shown in FIGURE 15.2(c). . . . . POSITIVE FEATURES Range of pile types and pile diameters available for various applications. No encroachment into adjacent property Reinforced concrete piles are particularly suited to permanent applications due to inherent ability to resist corrosion OTHER CONSIDERATIONS Costly heavily reinforced large diameter piles are required to provide sufficient shear and bending stiffness Passive systems that require some movement to mobilise stabilising forces INSTALLATION TECHNIQUE Soil dowelling techniques are generally used to stabilise marginally stable slopes which are often in a state of limiting equilibrium. It is therefore important to ensure that pile installation procedures do not have a negative influence on the stability of the slope being stabilised. Pile types with low vibrations during installation should therefore be used. A further requirement is that the piles used must be able to withstand large shear and bending forces. Heavily reinforced large diameter concrete piles are therefore most suitable. Actual pile diameters and spacings have to be decided upon to suit a specific project. Taking the above factors into consideration the following pile types are most suited to soil dowelling applications: .. .. . Auger Piles Auger Underslurry Piles CFA Piles Forum Bored Piles Oscillator Piles Jet grout columns can also be used to form soil dowels with adjacent columns intersecting for added stiffness. Reference should be made to SECTION 7.0: TECHNICAL DETAILS OF PILING SYSTEMS to obtain specific details for these pile types, the soil conditions for which they are most suited, and installation methods. 248 17.3.4 JET GROUT WALLS AND PLUGS A useful way of reinforcing the in-situ material is the use of jet grout columns intersecting each other to form a solid wall or an underground raft or plug. Where used to enhance the passive resistance between two retaining walls, an inverted arch is sometimes formed to resist heave and uplift pressures from the soil and water retained below. These plugs need careful design and they should not be highly stressed since the jet grout process provides low and variable strength in various soil types. Applicable Norms As the topic of Lateral Support and Slope Stabilisation is broad, not all aspects can be adequately covered in one book. The reader can get further details on this important topic at the following links / references: BS EN 12063: Execution of Special Geotechnical Work – Sheet Pile Walls (1999) BS EN 1537: Execution of Special Geotechnical Work – Ground Anchors (2000) BS EN 1538: Execution of Special Geotechnical Work – Diaphragm Walls (2000) CIRIA C 637: Soil Nailing – Best Practice Guidance and Fills (2005) SANS 207: The Design and Construction of Reinforced Soils and Fills (2006) 249 18.0 .. .. .. PROBLEM SOILS AND THEIR FOUNDATION SOLUTIONS The following six problem soils have been identified in the southern African region: Expansive Soils Collapsible Soils Soft Clays Dolomites Dispersive Soils Liquefiable Soils Most of these problems were identified in the early and mid 1950's with pioneering work carried out by Jennings, Knight and others. Research into the problems was carried out over a 30 year period and culminated in the State of the Art Conference on Problem Soils in South Africa, in 1985. The State of the Art papers published in the Civil Engineer in South Africa, in July 1985, provide a detailed reference for these problem soil types. The problem of dispersive soils is generally limited to embankments, dams and slopes, and has no particular relevance to foundation problems. Liquefiable soils are particularly relevant to areas subject to seismicity. Each of the problem soils has unique characteristics determined by several factors, such as the nature of the parent bedrock, the origin of the soil, the climate, vegetation and topography. Expansive soils are the most common and widely distributed of the problem soils, while collapsible soils are the next most frequently encountered. Dolomites are limited to areas underlain by rocks of the Campbell, Witwatersrand and Chuniespoort Groups while soft clays are generally limited to flood plains and estuaries of the eastern seaboard. Like most geotechnical engineering problems, problem soils require detailed and competent site investigation and laboratory testing procedures for their detection and evaluation so that the behaviour of the proposed structural foundation can be predicted. A basic geotechnical investigation will alert the engineer to the presence of a problem soil, and depending on the importance and type of development, further investigation and testing to more accurately define the problem may be required. General techniques of site investigations and testing outlined in SECTION 2.0 are applicable to sites underlain by problem soils. Specific investigation techniques and laboratory testing that may be required to evaluate and predict their behaviour are given in each problem soil section. Most of the recent work carried out on problem soils has concentrated on solutions and construction methods to treat the problem soil, or alternatively, to engineer solutions to overcome its effects. Many of these solutions have been developed to suit particular physical and economic constraints and are based on well established principles. The discussion of the engineering solutions for each problem soil will address both the treatment of the soil, as well as, engineering around the problem. 250 18.1 EXPANSIVE SOILS Nature of the Problem In southern Africa foundation problems relating to expansive soils are entirely due to the presence of secondary minerals derived from the decomposition of the parent bedrock. The geological origin of the parent material is the most important factor in the composition of these clay minerals. The formation of 2 :1 clay lattice minerals with water molecules occurring between successive sheets in the crystalline structure, characterises expansive clays. It is the variations in this moisture content which results in volumetric change of the soil skeleton. In drier areas particularly, increase in moisture content due to a change in the boundary conditions results in swelling of the soil skeleton with consequent upward movement of the soil and the foundations placed on it. The problem of expansive soils was first noted by Jennings and Henkel as early as 1947, and in southern Africa these soils constitute the most extensive problem for foundation engineers today. The most severely affected type of development, damaged by the swelling of clays underlying foundations, are single storey dwellings and light structures. Foundation problems due to heave are not limited to this type of structure however, and several case histories related to heavy industrial structures, such as power stations, have also been reported. Extensive research has been carried out on heaving subsoil by the CSIR and by the NIBR. Distribution of Expansive Clays These soils occur broadly in the central and eastern regions of southern Africa with the most severe conditions present in the highveld areas of the Free State and large areas of Gauteng and the northern regions of South Africa. These areas are indicated in FIGURE 18.1.1. Soils with a 2 :1 clay lattice structure can be classified as either Residual Soils or Transported Soils. Residual Soils Residual soils are developed from basic igneous or argillaceous rocks. Basic igneous rocks of the Bushveld Igneous Complex, Pretoria Group and Ventersdorp Supergroup are present over most of Gauteng and the northern provinces of South Africa, and the active clay horizons associated with these residual soils is generally shallow and close to the surface. Residual soils derived from argillaceous rocks are limited to areas underlain by the Karoo Supergroup which covers a large portion of the southern African landmass. Shales, Mudrocks, Tillites and Varvites of the Ecca, Beaufort and Dwyka Groups have residual expansive soils which can be up to 30 metres in depth. Residual soils developed from Doleritic bedrock in areas of Karoo sediments also exhibit heaving characteristics. Transported Soils Transported soils in the form of Alluvial, Lacustrine, Gulleywash and Hillwash can contain active clays. These deposits are present in southern Africa with the Vereeniging area in the Vaal River floodplain of particular note. 251 FIGURE 18.1.1 Distribution of Expansive Clays and Collapsible Soils Evaluation and Prediction The prediction of total heave values from basic site investigation data is possible using the modified van der Merwe (1964) method, or the method proposed by Weston (1979). Van der Merwe's method incorporates the use of a unit heave approach and does not take into account initial moisture content or in-situ density. Weston's method takes initial moisture content, density and overburden pressure into account, and is based on a statistical approach from measured heave values on road pavement structures. All other methods of prediction require specific laboratory tests on undisturbed samples, and in some cases, in-situ measurement of soil suction and density. Initial methods based on the double oedometer test, developed by Jennings and Knight (1956), are still commonly used, but tend to overpredict total heave, since lateral strain is not considered, and there is a conservative assumption regarding full saturation. Brackley (1975, 1980) has published two methods for estimating the percentage swell. The first method is an empirical correlation based on the factors affecting heave, while the second method is based on soil suction measurements carried out either in-situ, or on undisturbed, unstressed samples. More recent work considering stress path, soil fabric and crack fabric, requires sophisticated testing, such as free swell and swelling pressure measurements, and are seldom used due to their complexity and the problems associated with representative sampling. 252 Differential heave has received little attention, and commonly the assumption of Jennings and Kerrich (1962) is used where differential heave is assumed to be 50% of the total heave. Donaldson (1973) has provided quantitative guidance on the ratio of total and differential heave. The method of swell prediction given by Weston and outlined below takes all the factors governing heave into account, does not require sophisticated laboratory and in-situ testing, and is generally reliable for heave profiles up to 5 metres deep. FIGURE 18.1.2 provides a simple graphical method for the prediction of percentage swell. FIGURE 18.1.2 Percentage Swell after Weston (1979) Engineering Solutions Two methods of soil treatment are in common use in southern Africa: Removal and replacement, and pre-wetting. Soil removal and replacement has been frequently used in areas where a highly expansive, shallow heave profile has developed, mainly in areas underlain by rocks of the Bushveld Igneous complex. This solution is economically feasible for heaving profiles up to 2 metres deep and is highly dependent on the costs and availability of suitable inert replacement material. The alternative solution of pre-wetting has been used for heaving profiles up to 10 metres in depth. This technique involves the drilling of evenly spaced small diameter holes of 200 to 300 mm diameter over the entire area to be treated, and to the full depth of the heaving profile. The holes are filled with sand or coarse aggregate and continually charged with water until full saturation, or an equilibrium in moisture content, has been achieved. The method can be economical, but the amount of time needed to achieve the desired moisture content is unknown. Possible future changes in moisture content and consequent movements of normal spread foundations is another factor which is difficult to predict. These two unknowns often preclude the use of this method. 253 The use of lime columns to break down the 2 : 1 clay lattice has been used in other parts of the world, but has received little attention in southern Africa, and research into this method of soil treatment warrants consideration. Engineered solutions to overcome the problem depend largely on the depth of the expansive profile and magnitude of the heave movements, as well as, the type and sensitivity of the structure proposed. The following three solutions are presently in common use: . Piled Foundation A piled foundation with a suspended and isolated structure, can be used on all profiles and types of development, except perhaps the very deep heaving profiles found in areas such as Kimberley, where depth of the heaving profile can be up to 30 metres. The solution comprises bored socketed piles, or driven cast-in-situ piles, with an expanded base founded on, or within a stable horizon, supporting the suspended and isolated ground floor slab. Various solutions have been developed for isolating the ground floor slab, with the Jackslab method developed and patented by Franki providing the most cost effective method for small and medium size developments. The Jackslab comprises the casting of the reinforced slab directly on the prepared subgrade, and then jacking the slab off the sub-grade, using the piles as reaction for the jacking units. The slab can be jacked to provide the desired isolation and has the advantage of ensuring that all areas of the structure are fully isolated. The methods of design for piles in heaving soil conditions is covered in detail in SECTION 21.4, and is based on the assumption of complete isolation of the structure. . Stiffened Raft Stiffened raft foundations and their design have received much attention and the refinement and sophistication of design and construction methods is at an advanced stage. Their use is generally limited to light or residential structures of one or two storeys, up to 200 m2 in plan area, but these limits can be extended with the use of articulation of the superstructure. Design methods developed by the CSIR, Pidgeon, Lytton and others are complex soil/structure interaction analyses based on the principle of a plate on mound, as shown in FIGURE 18.1.3. There are several patented raft design and construction methods available, and the Boucell, developed by the CSIR, is both cost effective and simple to construct. The Boucell raft also provides a solution to some of the other problem soil types in southern Africa. . Superstructure Articulation Articulated structures on conventional foundations can only be used where estimated total heave values do not exceed 25 mm. 254 FIGURE 18.1.3 Plate on Mound for Stiffened Raft The recommendations of Jennings and Kerrich (1957) developed over 50 years ago, and outlined in TABLE 18.1.1, are still relevant today as a guide to a suitable foundation solution with cost comparisons. TABLE 18.1.1 Types of Construction for Various Heave Magnitudes Type of Construction (modified from Jennings and Kerrich) Estimated Total Heave (mm) Corresponding Maximum Deflection Ratio Estimated Additional Cost Normal - continuous brick walls on strip footings 0-6 1: 4000 0% Modified normal - high fanlights, reinforced footings and lintels 6 - 12 1: 2000 1 - 10% Split construction with reinforced brickwork 12 - 50 1: 500 10 - 30% Piles to limited depth with split construction and reinforced brickwork 50 - 100 1:1000 50% Underreamed piles with suspended floors 100 + No movement 100% Stiffened raft foundations > 12 mm Design parameter 7 - 15% 255 18.2 COLLAPSIBLE SOILS Nature of the Problem Behaviour of soil with a collapsing fabric was first studied by Knight (1961), and the basic concept of collapse settlement is illustrated in FIGURE 18.2.1. FIGURE 18.2.1 Mechanism of Collapse Settlement . . . . Several conditions must be satisfied for collapse settlement to occur: The soil must have a collapsible fabric. Soils of low in-situ dry density which are silty or sandy commonly exhibit a collapsible fabric An initial condition of partial saturation must be present. This condition is applicable to the upper horizons of the soil profile in most areas of southern Africa An increase in moisture content must occur so that a loss of shear strength of bridging colloidal materials can be effected The imposed pressure exerted on the soil fabric by the structure must exceed the overburden pressure Problems associated with construction on collapsible soils are not only confined to buildings with shallow foundation structures, but to roads, airfields and railways, as well as earth dams and reservoirs. Distribution of Collapsible Soils As with expansive soils, collapsible soils occur in both transported and residual soils. Since the problem of collapse can occur in all types of transported soil, foundation problems due to collapse can occur anywhere in southern Africa, but more commonly, and with greater severity in areas where aeolian sands have been deposited. Deep deposits of Kalahari silty sands in the arid western regions produce collapsible soil up to 20 metres in thickness. Residual soils with a collapsible grain structure are mainly confined to Granites of the Basement Complex. These collapsing Granites are confined to old erosion surfaces and areas with an annual water surplus. FIGURE 18.1.1 shows areas where collapsing soils are common. 256 Evaluation and Prediction The identification and quantification of collapse settlement of the soil fabric requires detailed field identification of the collapsible horizons, as well as laboratory or in-situ testing, to quantify the magnitude of collapse settlement. The recording of the soil profile discussed in SECTION 3.1 is the first step in the correct identification of the problem. Careful inspection with a hand lens, as well as identification of the origin of the soil will substantially aid the engineer in identifying the collapse phenomenon. Inspection of the damage to surrounding structures will also be invaluable in alerting the investigator to the problem of collapse. Oedometer testing in the laboratory is the most common method used to predict collapse settlements. The collapse potential test is the simplest of the types of oedometer test used, and provides the engineer with a measure of the collapse potential, but not a parameter for estimating collapse settlement. TABLE 18.2.1 provides a guide for the severity of collapse. TABLE 18.2.1 Collapse Potential Collapse Potential Severity of Problem 0% - 1% No problem 1% - 5% Moderate problem 5% - 10% Problem 10% - 20% Severe problem > 20% Very severe problem Single oedometer testing is used with the sample at natural moisture content, loaded to the anticipated applied stress from the structure, and then soaked. The measured consolidation due to collapse, provides a conservative estimate of the likely collapse settlements. The double oedometer test developed by Jennings and Knight (1958) is a fairly complicated procedure requiring corrections of the consolidation curves and careful interpretation. Plate load testing methods in which a 300 mm diameter plate is loaded to the anticipated bearing pressure, and the surrounding soils then soaked, can be successfully used to predict soil modulus values at natural moisture and soaked conditions. The collapse settlements can be calculated for the specific structure and applied loads from the measured soaked modulus. 257 Engineering Solutions Soil treatment methods for collapsing soils generally revolve around methods of compaction either by removal and recompaction, or by compaction in-situ. For shallow collapsible soil profiles, compaction using vibratory or impact rollers can be considered and has been used successfully for collapsible soil depths up to 1.5 metres. The use of this method has not always been successful and trials should be carried out to ensure the desired compaction can be achieved with the equipment proposed. Dynamic Compaction, described in SECTION 13.2, is a technique which can successfully be used for most collapsing profiles up to 10 metres deep. Other methods of compaction, such as Vibroflotation, have been used successfully and can also be considered. For collapsing profiles > 2 metres, deep dynamic compaction will provide the most cost effective solution for medium and large areas requiring treatment. Like expansive soils, engineered solutions for collapsible soils comprise either piled foundations or stiffened rafts: . . Most methods of piling can be used for collapsible soil profiles, but because the profile is generally partially saturated and slightly cohesive, auger piles or driven castin-situ piles are generally the most economical pile type. The driven cast-in-situ pile has the advantage that compaction of the surrounding collapsing soil is achieved during pile installation; Design methods for raft foundations on collapsing soils have been developed by Lytton (1972). Tromp (1979) has used the method of assuming a ‘soft spot’ of chosen diameter. A ‘soft spot’ is an assumed area of collapsible soil softened by a localised increase in moisture content. Raft types, such as the Boucell, are ideally suited to collapsing soils since the raft is equally stiff in both directions of bending and is therefore not sensitive to the position of the soft spot. 258 18.3 SOFT CLAYS Nature of the Problem As a problem soil type soft clays have no specific connotation to the southern African region and are more widely distributed in other areas of the world, such as Scandinavia. These soils exhibit very low shear strength, high compressibility, and lead to severe time related settlement problems. In southern Africa these clays are often partially saturated and over-consolidated. Some typical undrained shear strengths and compressibility values associated with these soils, which are mainly confined to the eastern seaboard of southern Africa, are given in TABLE 18.3.1. The problems associated with these soils are stability and settlement related. Instability and large settlements for heavy loadings such as road embankments, present engineering problems to infrastructural developments. Most building structures located on these soils demand a piled foundation solution. TABLE 18.3.1 Typical Properties of Soft Clays Typical Shear Strength Values Area Material Description Undrained c u (kPa) Drained c’(kPa) South Coast Black silty clay 10 - 25 5 25o Durban Black silty clay 10 - 25 0 25o Richards Bay Sandy silty clay Black silty clay 15 - 35 10 - 20 10 - 20 5 -15 25o - 30o 20o - 25o φ‘ Typical Compressibility Values Area South Coast Durban Richards Bay Material Description Compressibility m v (m2 / MN) Consolidation c v (m2 /year) Light grey sandy clay Black silty clay 0.3 - 1.0 0.5 - 1.5 5 -10 1-5 Black silty clay 0.5 - 2.0 0.5 - 2.0 Light grey sandy clay Black silty clay 0.2 - 1.0 0.5 - 2.0 5 -10 1-5 Evaluation and Prediction Testing and evaluation of these soils is done using normal site investigation techniques and soil mechanics principles. Testing using small diameter rotary cored boreholes with in-situ SPT and shear vane tests, in conjunction with CPT tests, are the most cost effective methods of investigation. There are numerous empirical and rigorous methods for establishing geotechnical design parameters for these soils from the insitu or laboratory test results. 259 Engineering Solutions There are few methods available to increase the strength or stiffness of soft clays, but dynamic compaction and/or dynamic replacement methods have been used successfully in estuarine sediments, which are generally slightly cohesive, with lenticular deposits of silts and clays. Lime columns have been used in other parts of the world to cement soft clay horizons and enhance their properties. A piled foundation is generally used for medium and heavy structures in areas underlain by soft clays, the piles being founded on bedrock, or dense sand horizons, underlying the soft clays. The choice of pile types suited to soft clay conditions can be assessed using the parameters set out in SECTION 4.0. For piles of low and medium capacity, driven jointed precast piles have been used extensively in the Durban and Richard's Bay areas. Large diameter bored piles founded on, or socketed into bedrock, have been used for heavily loaded structures. In certain areas where a sand stratum overlies the soft clays, it is sometimes possible to found light structures on Franki piles with an enlarged base, or on footings, founded on soil improved by Vibratory Replacement. Negative skin friction due to settlement of soft horizons under surcharge loading can occur in these soft clay profiles, and the additional load imposed on the pile must be taken into account when carrying out the pile design. The alternative is to eliminate the effects of negative friction by isolating the pile shaft using Shell pileslip which is a bitumen product developed specifically for this purpose. This has been done in the Richard's Bay area. For embankments and deep fills where piling is not an economical option, several methods of construction have been successfully used. The installation of vertical drains to shorten drainage paths and rapidly dissipate pore pressures is economical but requires phasing of the construction process and careful monitoring of movements and pore pressures. The use of preloading is both economical and effective if the necessary pre-planning can be implemented. Slope flattening or berms are often used to decrease the shear stress where shear strength is a problem during the construction phase. 260 18.4 DOLOMITES Nature of the Problem Solution cavities within water soluble Dolomitic rock masses result in the formation of sinkholes or subsidences being formed above the cavities due to changes in the groundwater regime. Lowering of the water-table with the resultant removal of water from voided areas of the Dolomitic bedrock causes sub-surface erosion into the voids by infiltration of surface water. Cavities within the Dolomitic residuum move upwards during this erosion process until the cavity reaches the surface resulting in the formation of a sinkhole. A compaction subsidence, as opposed to a sinkhole, occurs where the cavities are filled with highly compressible Dolomitic residuum called WAD. The phenomenon is graphically illustrated in FIGURE 18.4.1. FIGURE 18.4.1 Typical Dolomitic Condition Distribution of Dolomitic Rocks The occurance of Dolomitic conditions is limited to areas underlain by this rock type. Dolomitic rocks of the Chuniespoort, Campbell and Witwatersrand Groups are limited to the northern areas of South Africa, and are shown in FIGURE 18.4.2. FIGURE 18.4.2 Distribution of Dolomitic Rocks in South Africa 261 Evaluation and Prediction The investigation of Dolomitic areas and the evaluation of the stability of these areas requires personnel with expertise and experience in this field of geotechnical engineering. The investigation of Dolomitic areas incorporates both geophysical and direct methods of investigation. Gravity surveys are the most common and reliable geophysical method available and are carried out in conjunction with direct methods such as percussion drilling, auger trial holes and backactor test pits. The use of borehole cameras lowered into small diameter percussion drilled holes facilitates the inspection of areas where cavities are likely. Classification and evaluation of Dolomitic areas is generally done on an empirical basis. Classification of sites is done according to Wagener's (1982) method where the depth of overburden provides the basis for the site being designated as a Class A, B or C site. The average thickness of overburden C from ground level to top of pinnacles gives classifications which reflect the risk of instability as outlined in TABLE 18.4.1. TABLE 18.4.1 Dolomite Classification Class A Pinnacle and boulder Dolomite at or near the surface. C<3m Class B Pinnacle and boulder Dolomite overlain by moderately thick overburden. 3 m < C < 15 m Class C Pinnacle and boulder Dolomite overlain by thick overburden. C > 15 m Engineering Solutions No economical methods of soil treatment have been developed for Dolomitic areas since the detection and backfilling of voided areas is extremely difficult. Control of surface water to prevent erosion into cavities, as well as strict control of groundwater movement, are the most effective methods of prevention. Engineered solutions comprise either the formation of a raft or the use of piled foundations. The use of soil rafts, or mattresses, developed by Wagener (1963) is commonly used, and is suitable for light and medium structures. Piles are generally used for heavy and movement sensitive structures. The four pile types generally suited to the difficult conditions encountered, are Oscillator Piles of large diameter and capacity, pre-drilled precast concrete piles, pre-drilled steel H-piles, and the recently introduced Rotapile. The pre-drilling is carried out, under a head of drilling foam, using large diameter down-the-hole percussion hammers. Driven cast-in-situ piles founded in a dense horizon of sufficient thickness can also be cost effective as a shallow piled solution for light and medium structures. 262 18.5 DISPERSIVE SOILS Particular care should be taken when designing earth dams, drainage channels and lateral support where the soil mass within the structure is dispersive as these soils deflocculate when permeated by relatively pure water and then become susceptible to erosion and piping. 18.6 LIQUEFIABLE SOILS Soil liquefaction occurs when loose sands temporarily change from a solid state to having the consistency of a heavy liquid. Soil liquefaction is a consequence of increasing pore water pressures and corresponding decrease in effective stress induced by loose sands tendency to decrease in volume when subjected to cyclic undrained loading (eg earthquake loading). The vast majority of liquefaction hazards are associated with saturated sandy and silty soils of low plasticity and density. Cohesive soils with clayey content (particle size < 0.005 mm) greater than 15% are generally not considered susceptible to soil liquefaction. Liquefaction typically occurs in cohesionless sands, silt, and fine-grained gravel deposits of the Holocene to late Pleistocene age in areas where the groundwater is shallower than about 15 metres. The resistance of the cohesionless soil to liquefaction will depend on the density of the soil, confining stresses, soil structure (fabric, age and cementation), the magnitude and duration of the cyclic loading, and the extent to which shear stress reversal occurs. Below an SPT of ‘N’ = 15 the risk may be high, and above ‘N’ = 30 soils are usually not prone to liquefy. 263 19.0 ENVIRONMENTAL ENGINEERING Franki Africa offers the following range of services related to the engineering of the environment: .. .. . Core drilling for the interpretation of fracture patterns and aquifers Permeability testing Groundwater sampling and monitoring Sampling and monitoring of surface water bodies Containment / remediation of contaminated areas Core drilling and permeability testing are envirotechnical services that also form part of procedures used in general geotechnical investigations. The techniques adopted for these two procedures are described in detail in SECTION 2: GEOTECHNICAL INVESTIGATION. The other services are specific to envirogeotechnical operations and are discussed in detail in this section. 19.1 GROUNDWATER MONITORING The frequent reporting in the media of major pollution scares, the general tightening of legislation regarding environmental issues, and the greater public awareness of pollution, has highlighted the need to monitor groundwater for potential contamination. For far too long industry has not been aware of the problem of groundwater contamination, or has conveniently ignored it. This situation is now changing and industry has to get its house in order. To firstly ascertain whether the groundwater is contaminated, samples of water have to be obtained. Tests on these samples will show the degree of contamination and the area over which contamination has occurred. The groundwater can also be monitored over a period of time to determine the possible source of the pollutants and other factors such as flow patterns and the effectiveness of remedial measures. This involves the installation of monitoring wells and the use of specialised electronic equipment. Soiltech provides a complete service with respect to sampling and monitoring of groundwater. Installation of Wells The wells are installed using a variety of techniques depending on sub-surface conditions. Rotary percussion or auger drilling are the most common methods of well installation. The work must be carried out to stringent standards to prevent crosscontamination of the samples taken from the various wells. For this reason Franki operates a quality assurance programme which is implemented in accordance with ISO 9001 requirements. Work instructions and quality control procedures for individual contracts are prepared to suit client requirements and engineering specifications, thereby assuring the end-user of the highest quality results. 264 Monitoring Franki collects its water monitoring data via the computerised Grant / YSI 3800 water quality logger unit. This equipment records conductivity, pH, dissolved oxygen, salinity, Eh (ORP) and temperature. The electrode readings are compensated automatically for temperature and pressure. Readings are collected via a sonde, using pump purging techniques within the monitoring wells. Pump purging is carried out along guidelines supplied by the Water Research Commission and Department of Water Affairs and Forestry. The pumping equipment consists of a Grundfos MP1 environmental submersible pump. Field data can be either printed out in tabular form directly from the logger unit, or downloaded onto a PC with the ability then to graph and analyse data further. The ongoing monitoring of water and industrial effluents, can assist in determining the source of pollution incidents, so that remedial action can be taken timeously. The need for expensive and time-consuming laboratory sample testing is reduced, since field staff can determine, on site, whether certain basic parameters lie outside pre-defined limits, allowing for straightforward interpretation and trend analysis. When water samples are required for additional laboratory testing, these are collected and preserved according to the latest specifications for transit to approved analytical laboratories. If necessary, the monitoring can be supplemented by the collection and analysis of contaminated soils using Franki's geotechnical equipment. The various sampling procedures are discussed in SECTION 2.0: GEOTECHNICAL INVESTIGATION. 19.2 MONITORING OF SURFACE WATER The same techniques as described for groundwater monitoring are used for the monitoring of open surface water bodies, except that samples are taken directly from the surface water body. 265 19.3 CONTAINMENT / REMEDIATION With the growth in awareness of the consequences of pollution to groundwater systems, the need to control groundwater flow by the construction of hydraulic barriers has received renewed attention. In groundwater pollution control, the barrier often performs two functions. The first is to isolate the polluted groundwater by controlling the seepage into downstream areas. The second is the containment of the contaminated water, allowing in-situ treatment by de-watering, or recirculation, in the case of stabilisation applications. In the last decade, Franki has constructed various forms of hydraulic barriers to control groundwater flow. Slurry walls, diaphragm walls and steel sheet-pile walls have all been used for this purpose. 19.3.1 SLURRY WALLS Slurry walls are generally employed to control groundwater movement in soil horizons for pollution control measures as listed, or as an hydraulic cut-off for an earthfill dam, or similar structure. They can be constructed under dry stable conditions, as well as, in unstable soil profiles below the water-table. INSTALLATION TECHNIQUE Where the depth of the slurry wall is less than six metres, it can be excavated as a continuous trench using standard trench excavating equipment. For deeper walls, the excavation is carried out in vertical panels utilising grabbing techniques, as described in SECTION 17.5: UNDERSLURRY WALLS. During excavation, the level of the slurry in the trench must be kept to a minimum of 1.5 metres above the level of the surrounding water-table. For the shallow trenches, either a cement / bentonite or a sand / bentonite slurry can be used. With a sand / bentonite slurry, the trench is excavated initially using a conventional bentonite slurry. On completion of the excavation, the sand / bentonite slurry is discharged into the trench using a tremie, and the bentonite slurry pumped to storage. If a cement / bentonite slurry is chosen, the slurry is left in the trench to set after completion of the excavation, eliminating the need to change the slurry. For deeper walls excavated by grab, preference is given to a cement / bentonite slurry due to its self-hardening nature. This makes it possible to excavate into a previously constructed panel without endangering the stability of the previously constructed work. As mentioned, the cement / bentonite slurry is left in the trench to set after completion of the excavation. The trench can vary in width between 400 and 1200 mm with 600 mm being the most common. Depths in excess of 20 metres are achievable using grabbing techniques. With excavations deeper than 12 metres, special precautions are often necessary to ensure the verticality and effective overlapping of adjacent panels. MIX DESIGN The prime consideration of a slurry mix is low permeability which remains stable under the geotechnical and groundwater conditions anticipated on site. For this reason it is essential to carry out a geotechnical investigation of the site, which includes the sampling of the groundwater. 266 Permeabilities of the order of 10 -6 - 10 -8 cm / sec can be achieved with a sand/ bentonite mix, as this generally has a lower permeability than the cement/bentonite alternative. In designing a sand / bentonite mix, it is imperative that a well-graded ‘dirty’ sand is used to prevent the risk of internal piping of the bentonite particles between the coarser grained sand particles. A silt content of at least 20% should be aimed at. With a cement / bentonite slurry, the quantity of cement will depend on the required strength and flexibility of the cut-off wall. A higher cement content will result in a higher strength slurry with increased brittleness and reduced flexibility. The use of slagment in combination with cement, partially offsets these negative factors with the added benefit of a reduction in permeability. The slurry mixing procedure is important and will greatly control the performance of the product. The following basic procedure must be adhered to: .. . Bentonite must be fully hydrated with an uncontaminated water prior to use The cementitious addition should be colloidally mixed with uncontaminated water prior to blending with the bentonite A homogeneous mix must be obtained by thorough blending and mixing prior to introduction to the trench CHEMICAL RESISTANCE In the case of polluted groundwater, the resistance of the bentonite slurry to the contamination in the groundwater must be checked, preferably by testing to ensure the long-term performance of the cut-off wall under site conditions. VARIATIONS IN INSTALLATION TECHNIQUE While a slurry trench can effectively control the horizontal migration of water into or out of the area to be isolated, water flow can occur through the base of the contained area, or into the contained area from the surface. A permeable rock formation below the slurry wall may require grouting to extend the cut-off to competent rock. A clay or similar capping may be necessary to control the inflow of water into the isolated zone, as well as outflow. If a capping for the containment area is envisaged, or the barrier is to be extended above the surface by constructing a concrete or earth wall, attention should be paid to the detail of joining the two barriers systems. In the case of the cement / bentonite slurry, allowance should also be made to accommodate the inevitable cracking that occurs in the upper 300 to 500 mm of the cement / bentonite slurry. 19.3.2 OTHER BARRIER SYSTEMS In addition to the cut-off system noted in 19.3.1, steel sheet piles (see SECTION 17.1), diaphragm walls (see SECTION 17.7) and Cutter Soil Mix (CSM) walls (see SECTION 13.1) are often effective and economical barrier systems. Jet Grouting, Grouting and de-watering provide possible alternatives for controlling groundwater flow, be it for contamination control, water preservation, or flood control. It is recommended that discussions be held with Franki to optimise the barrier best suited to the circumstances of a given situation. 267 20.0 MARINE FOUNDATION ENGINEERING 20.1 PILING FOR MARINE STRUCTURES Piles are used for many marine applications, either as foundations to carry structural loads in poor ground conditions, or as structural members which provide both a foundation and structural element to jetties, quays and dolphins. Piles are also used as tension members, which prevent uplift on marine structures such as dry docks, or as elements carrying horizontal loads such as tie-backs to earth retaining structures, or structures resisting significant horizontal loading from ship berthing. Most piling applications for marine structures involve a shallow phreatic surface and both driven piles and bored piling methods, which are installed with permanent or temporary support to the pile bore, are suitable. For detailed reference of these pile types see SECTION 7.0: PILED FOUNDATIONS. 20.1.1 DRIVEN PILES The most suitable driven pile type for marine structures is the Tube Pile, which is described in detail in SECTION 7.3. Driven tube piles can be installed with a thin-walled casing (6 to 10 mm) and concreted and reinforced in-situ to provide the permanent structural element, with the tube providing a temporary lining. The tube pile can also be installed with a thick-walled casing (10 to 40 mm) where the steel casing forms the primary structural load-bearing element. Reinforced concrete, cast-in-situ, can enhance the structural capacity of the piles and provide additional durability of the pile shaft, particularly above seabed level. Driven tube piles are widely used in a marine environment since they have the advantage of easily providing a supporting member to structures located in the sea. Careful assessment of the corrosive environment must be undertaken to ensure that the durability requirements are met. Cathodic protection or durable coatings are generally essential to ensure long-term durability. A wide range of pile diameters, from 300 mm for minor structures, to 1500 mm for heavy marine applications, can be installed where poor soil conditions underlie the site. Closed-ended bottom driven tube piles are generally not suited to sites where penetration of bedrock or very dense /very stiff soil is required. Penetration of hard layers and limited penetration into rockhead can be achieved with open driven piles. Rocksockets of up to 3 metres can be achieved using heavy chiselling and grabbing below the driven casing. This method of penetrating rock is reasonably economical for soft or highly jointed medium hard rock. For massive hard rock more economical methods of penetration should be considered. Driven precast concrete piles, outlined in detail in SECTION 7.2, can be effectively used for marine applications, and can perform a similar function to the driven tube pile. Precast concrete pile shafts provide a high quality, durable product for marine applications, but are limited in their use by the size of member that can be installed, as well as the general requirements that joints within the pile shaft should be located below the groundline. It is generally advisable to install precast piles without joints 268 where they are required to provide structural support above seabed level and prestressed single length precast concrete shafts are preferable. This, together with the limitation on size of member that can be handled, preclude the precast pile from being used on marine structures in deep water, or those which are heavily loaded. Tubular precast concrete piles can be considered for these conditions, but this pile type has not been used in the African region. Driven steel H-piles can be used in similar applications to the precast concrete pile, but the pile type is susceptible to corrosion and therefore less cost-effective than the driven tube pile. Details of the driven H-pile are given in SECTION 7.4, but it must be pointed out that the supply of heavier sections (350 x 350) is limited, and careful investigation into the supply situation must be undertaken if this solution is to be considered. FIGURE 20.1.1 shows typical marine structures where driven tube piles are often utilized, and PLATE 20.1.1 shows tube piles being installed for a fishing quay in Mauritius. PLATE 20.1.1 Installing Driven Tube Piles for Fishing Quay, Mauritius 269 (a) Suspended Deck Quay (b) Suspended Deck Jetty (c) Access Trestle FIGURE 20.1.1 Marine Structures Suited to the Use of Driven Tube Piles 270 20.1.2 BORED PILES Open drilled auger piles, or CFA piles, are generally not suitable for use in a marine environment. Bored piles which are installed with a temporary or permanent steel liner, or piles drilled under bentonite usually with a steel liner which is temporary or permanent over the upper zones, are suitable for a wide variety of marine structures. Piles of 600 to 2500 mm diameter can be installed to depths of up to 70 metres, using modern piling rigs and one of the many installation techniques available. An economical bored pile solution for heavy marine structures is the auger underslurry pile, covered in SECTION 7.6, where a thin-walled casing is used as a permanent liner and is vibrated a few metres below the seabed level and extends to the required cut-off level, which should be a minimum of 2 metres above maximum tide level. The auger underslurry solution can be installed in a wide variety of soil and rock conditions with the use of appropriate drilling tools. Where extensive penetration into hard rock layers is required, a more economical and quicker piled solution would be the Rotapile described in SECTION 7.11. Piles from 300 mm to a maximum of 1000 mm can be rapidly installed into hard rock using propriety casing systems. The casings, 8 to 12 mm thick, are installed permanently, and can extend from the required cut-off level above maximum tide level down to the pile-toe. It can be advantageous, when the required penetration into rock has been achieved, to lift the permanent liner approximately 2 x pile diameters, before the pile shaft is reinforced and concreted in-situ. Where large diameter piles up to 2.5 metres in diameter must be installed in hard rock conditions, the Oscillator pile described in SECTION 7.9 should be considered. A thinwalled permanent liner is installed from cut-off level to a few metres below seabed level, to facilitate concreting of the pile above seabed level. All piles installed in a marine environment should be concreted using tremie techniques. Due to durability considerations, concrete strengths in excess of 30 MPa should be used, but strengths in excess of 40 MPa should be avoided, unless high quality aggregates are available and the required cement content and workability can be achieved. Where piles are permanently lined normal reinforcement can be used, but where no liner is utilized, it is advisable to use coated reinforcement if aggressive marine conditions are prevalent. 271 20.2 EARTH RETENTION FOR MARINE STRUCTURES Earth retention is required on most marine structures, but more particularly for quays, jetties, dry docks and slipways. A wide variety of earth retention systems are available and the choice should be based on the following key parameters: .. .. .. .. Seabed Conditions Geotechnical Considerations Environmental Requirements Sea State Durability and Operational Requirements Materials Availability Earth retaining structures are classified as solid structures, with a vertical berthing face, and are generally divided into two categories: Sheet-pile Walls or Reinforced Concrete Cast-in-situ Walls Gravity Walls 20.2.1 SHEET-PILE WALLS Steel sheet-pile walls are the most widely used wall element in quays, jetties, dry docks and slipways. They are relatively light and easy to handle and can be installed in a wide variety of soil conditions due to their low displacement characteristics. The most commonly used interlocking sections are the U and Z types. Several composite wall sections are also available usually combining the U and Z sections with H-piles, box piles or tubular piles. Typical details and arrangements for high modulus steel sheetpile walls are given in FIGURE 20.2.1. Steel sheet-pile walls can be designed for a wide range of retention depths, with composite sections being suitable for the deeper retention depths, generally in excess of 10 metres. Precast reinforced concrete sheet-pile walls can be considered where relatively shallow retention depths are required. The weight and difficulties in extending the length of individual elements together with installing the piles without damage in hard driving conditions often preclude the use of this solution. Interlock of adjoining segments is not as effective as steel sheet piles and post grouting of the joints is generally required. Typical details and arrangements of precast concrete sheet piles is given in FIGURE 17.1.3. 272 (a) High Modulus Composite Sheet Piles (b) Combi Sheet-pile Wall FIGURE 20.2.1 Typical Details of Composite Sheet-pile Walls Sheet-pile walls can be installed as cantilever structures for shallow depths of retention, but are generally installed with tie-backs, or relieving platforms can be incorporated into the design to reduce earth pressures for jetty and quay construction. Tie-backs typically consist of steel ties fixed to a deadman anchorage or anchor walls. Typical arrangements of sheet-pile wall construction are given in FIGURE 20.2.2, with FIGURE 20.2.3 giving typical details for a steel sheet pile quay wall. A steel sheet pile quay wall with sheet-pile anchor wall being installed at a shipbuilding yard in Mauritius, is illustrated in PLATE 20.2.1. Sheet wall structures are most suitable when the ground below the dredged level is of medium or dense granular soil, or firm and stiff cohesive soil. Sheet walls may be used in weak soils, but excessive penetration depth is required, and they may be unsuitable if extensive depths of soft clay or loose sands are present. Where the seabed comprises mainly rock horizons, pre-treatment of the seabed or pile-toe will be required to allow sheet piles to be installed. 273 (a) Cantilever Wall (b) Wall with Relieving Platform (c) Tied-back Wall FIGURE 20.2.2 Typical Sheet-pile Wall Construction 274 x x FIGURE 20.2.3 Typical Arrangement of a Steel Sheet Pile Quay Wall PLATE 20.2.1 Partially Completed Steel Sheet Pile Quay Wall 275 20.2.2. DIAPHRAGM WALLS AND CAST-IN-SITU CONCRETE PILE WALLS Diaphragm walls and secant or contiguous pile walls are economical to build where there is a sufficient width of existing ground or an artificial embankment is available for use as a working area. They may be built in a wide variety of soil conditions and can be installed in rock with suitable modern equipment. These solutions are not suitable where flowing water or artesian conditions are present. Diaphragm walls are constructed in panels which can either be straight or T-panels in wall widths between 0.6 and 1.5 metres. Typical arrangements of wall panels are given in FIGURE 20.2.4. For a more detailed description of diaphragm walls reference should be made to SECTION 17: DIAPHRAGM WALLS. (a) Straight Panels (b) T-Panels (c) Long T-Panels Supporting Relieving Platform, with Intermediate Straight Panels FIGURE 20.2.4 Arrangement of Diaphragm Wall Panels 276 Contiguous or secant pile walls are constructed using pile diameters between 0.6 and 1.8 metres, with secant piles spaced at 0.75 to 0.9 x pile diameter, to ensure continuity of the wall. Contiguous piles should only be adopted where there is no risk of soil migration through the pile joint. Cast-in-situ concrete sheet walls, like steel sheet pile walls, can be tied-back, and can also incorporate relieving platforms. A typical arrangement of a T-panel diaphragm wall panels with relieving platform is shown in FIGURE 20.2.5. The use of a T-shaped diaphragm wall to construct a dry dock at a shipbuilding yard, in Mauritius, is shown in PLATE 20.2.2. FIGURE 20.2.5 Typical Use of Diaphragm Walls with Relieving Platform for Quay Construction 277 PLATE 20.2.2 Diaphragm Wall Used for a Dry Dock at a Shipbuilding Yard, Mauritius 20.2.3 GRAVITY RETAINING STRUCTURES Gravity structures are used where the seabed soils are of good quality, and are generally constructed in water, using methods unique to maritime works. Large precast elements are lifted or floated into position and placed on a prepared seabed. Where gravity structures are built in the dry, they are placed behind a cofferdam on a prepared seabed, or placed in an excavation protected from water ingress by de-watering and cut-off walls. Gravity walls used in maritime works are generally required to retain reclaimed ground, the quality of which can be selected. It is usual to use rubble or a free draining granular fill behind the wall, so that the effects of tidal lag and earth pressures are minimised. Many of the harbours in the southern African region have been constructed using gravity structures, which provide good durability and require limited maintenance. These structures are susceptible to movements if adequate scour protection has not been specified, and present difficulties if future berth deepening is required. Problems have also been encountered where gravity structures have been used in poor ground, or where inadequate seabed preparation has been undertaken. A wide variety of Concrete Blockwork wall structures have also been utilised and include solid blocks or hollow blocks with granular void filling. Typical examples of this type of wall are shown in FIGURE 20.2.6. Precast reinforced concrete retaining walls with counterforts have also been successfully used in the African region. The use of this type of construction for a deep berthing facility in Angola, is illustrated in PLATE 20.2.3. 278 - FIGURE 20.2.6 Typical Concrete Block Construction for Quay Walls 279 PLATE 20.2.3 Placement of Counterfort Precast Concrete Wall Panels in Angola 20.2.4 COFFERDAMS AND CAISSONS Cofferdams consist of open cellular structures placed on a seabed prepared at the required berthing level, with the structure projecting above high water level. The cells are filled with free draining granular material, and a capping structure of solid reinforced concrete is generally placed over the cofferdam sheet piles. Fenders, bollards and service ducts can be placed in the cast-in-situ concrete capping element. Where precast reinforced concrete caissons are used, they can be rectangular, circular, or cloverleaf in shape. The elements are constructed in the dry, floated into position, and then placed on a prepared seabed. Cellular steel sheet pile structures consist of cells formed by interlocking straight-web steel sheet piles driven with the use of guide frames. The cells are filled with granular soils in a carefully designed sequence and capped with a solid reinforced concrete structure. Deep-water berths can be constructed using this type of structure where the average width of the cofferdam should be greater than 0.8 x the retained height. Where soft clay or loose sand horizons are evident below seabed level, the soft material should be dredged before backfilling of the cells commences. Care should be exercised when using these structures with deep horizons of poor quality in-situ material, since compression of the cells can occur with consolidation of these horizons. Details of typical cofferdam structures using precast concrete elements as well as sheet pile cellular structures, are shown in FIGURE 20.2.7. 280 (a) Circular Cells (b) Diaphragm Cells (c) Typical Cross-section FIGURE 20.2.7 Typical Caisson and Cofferdam Configurations 281 20.3 SOIL IMPROVEMENT FOR MARINE APPLICATIONS Harbours are often located on, or near rivermouths and estuaries, and are frequently associated with deep transported soil horizons. In many instances, these horizons exhibit low strength, and are highly compressible, which presents engineering challenges for the heavy loadings applied by many marine structures. There have been great advances in developing effective and economical soil improvement systems over the last 20 to 30 years, and many of these developments can be utilised for marine structures. All of the soil improvement techniques outlined in SECTION 13.0 can be considered for a wide variety of marine structures. Most of these techniques were developed initially for land based operations, but have recently been adapted for a marine environment. 20.3.1 SOIL COMPACTION TECHNIQUES FOR MARINE WORK Vibratory compaction of loose granular soils below seabed level is the most efficient and cost effective soil improvement methodology for this soil type in a marine environment. The process described in detail in SECTION 13.1 can easily be implemented below significant water depths, and there is no major cost penalty in completing the process over water. Dynamic compaction, as described in detail in SECTION 13.2, has been successfully adopted on the landward side of marine structures, and the process provides a competent and uniform sub-grade for pavement structures behind quays, dry docks and similar marine structures. The improvement in the properties of retained soils using Dynamic Compaction provides significant reduction of earth pressures. Compaction of loose sands in-situ is often mandatory in seismic areas where the risk of liquefaction can be catastrophic, as demonstrated in the severe seismic event at Kobe, in Japan, in the 90’s. 20.3.2 SOIL REPLACEMENT TECHNIQUES FOR MARINE WORK Vibratory replacement methods have been developed to improve soft cohesive soils in-situ beneath the seabed and provide adequate founding for gravity structures. Vibratory Replacement can also be effectively used on reclaimed areas, where excessive settlements are likely beneath large reclaimed areas, which are required for container and bulk storage on the landward side of berthing facilities. Vibratory replacement methods are described in detail in SECTION 13.4 and can be effectively combined with preloading to reduce settlements to acceptable levels for on-grade pavement structures where significant depths of soft compressible materials are present. 282 20.3.3 SOIL CEMENT COLUMNS FOR MARINE WORK Deep soil mixing methods have been extensively used in Japan and other regions characterised by poor seabed conditions. The installation of soil / cement columns, using deep mixing methods, is effective in significantly improving both granular and fine grained soils, and can eliminate the potential for liquefaction in areas subject to seismic activity. This technique has been successfully adopted on large marine projects, such as the repair of Kobe Harbour and the large reclamation work associated with the construction of the Kansai Airport, at Osaka, in Japan. The formation of jet grout columns, described in detail in SECTION 13.8, has been effectively used for remediation work on existing marine structures, and is covered more fully in SECTION 20.5. 283 20.4 CONSTRUCTION METHODS FOR QUAYS AND JETTIES Marine construction is most easily and economically carried out using land-based techniques, but in many instances this is not feasible and construction over the water is unavoidable. Three methods of construction can be considered for a water-based operation: .. . Work sequentially from the permanent structure; Install a temporary working platform to facilitate the construction of the permanent structure; Use floating equipment to install the permanent structure There are many factors which govern the choice of construction methodology and each project will have a unique solution which is optimal from a technical and economical viewpoint. Whatever solution is adopted, it is essential that a fully detailed and well executed construction sequence is implemented. This aspect of marine engineering is vital on projects which require water-based construction methodologies. 20.4.1 LAND-BASED CONSTRUCTION METHODS A land-based operation facilitates the consideration of a much wider range of both marine structural and foundation solutions. In general, the full range of geotechnical solutions can be considered since there is generally no requirement to adopt a preformed pile or earth retaining structure. If large open areas of reclaimed or onshore land are available for the project, then a wide range of construction techniques and sequences can be considered when planning the works. 20.4.2 WATER-BASED CONSTRUCTION METHODS The use of the completed permanent structure as a construction platform provides a low-risk working environment, but severely restricts the construction sequence and programme, as well as providing additional parameters that must be considered in the design of the permanent works, since the marine construction equipment is heavy and can transmit large horizontal loads. The construction of a spooling jetty in Angola illustrates this construction method in PLATE 20.4.1. The use of temporary construction platforms is often used for piers and jetties in relatively shallow water and provides greater flexibility for the construction of the permanent works. The temporary construction platform can be used for the construction of both the foundations and superstructure, and these operations can be carried out separately. PLATE 20.4.2 shows the construction of an oil pipeline trench near Durban, using this technique. 284 Large barges which can carry heavy construction equipment are generally used where deep water is present and / or the structure is placed well off the shoreline. Floating equipment is not suitable for working in the surf zone, or where a severe sea state could develop in a storm, or where cyclonic conditions are likely. There have been many instances of severe damage or loss of equipment during these events. The use of jack-up barges which provide a stable platform are more suited to working in moderate water depths than floating barges, but it is essential to ensure that the barge can be elevated above potential swells at all times. PLATE 20.2.3 shows the use of a large floating barge for the construction of a precast concrete retaining structure for a quay in Angola. PLATE 20.4.1 The Construction of a Spooling Jetty in Angola 285 PLATE 20.4.2 The Construction of an Oil Pipeline Trench, in Durban, South Africa PLATE 20.5.1 Jet Grouting in Dar es Salaam Harbour, Tanzania 286 20.5 REHABILITATION OF QUAYS AND JETTIES Many of the harbour facilities in the African region are either in a poor state of repair, or are inadequate to cater for the large marine vessels now in operation. There is currently a need for either repair and/or deepening of these facilities to accommodate modern vessels and on-shore facilities. Inadequate scour protection on many of these installations has resulted in severe undermining of gravity structures by large modern vessels with powerful bow-thrusters. Where existing concrete or steel structures have deteriorated extensively, there is often little alternative to a full replacement solution. Where foundation strengthening or stabilisation is required then soil improvement solutions such as jet grouting or reticulated micropiles can be used as an effective method of rehabilitation. Where deepening of existing gravity structures is required, or where severe scour has occurred and the seabed is severely weakened, the use of reticulated micropiles or jet grouting combined with the installation of sheet piles in front of the wall toe can be an effective and economical method of rehabilitation. PLATE 20.5.1 shows the successful use of jet grouting to stabilise erosion behind the existing quay in the Dar es Salaam harbour. Applicable Norms As the subject of Marine Engineering covers many aspects of Civil Engineering, the topic cannot be adequately covered in one book. The reader can get further details on this topic at the following links / references: BS 6349 - 1: Maritime Structures: Part 1 – General Criteria (2000) BS 6349 - 2: Maritime Structures: Part 2 – Quay Walls, Jetties and Dolphins (1988) BS 6349 - 3: Maritime Structures: Part 3 – Drydocks, Slipways, etc (1988) BS 6349 - 4: Maritime Structures: Part 4 – Fenderings and Moorings (1994) BS 6349 - 5: Maritime Structures: Part 5 – Dredging and Reclamation (2000) BS 6349 - 6: Maritime Structures: Part 6 – Inshore Moorings and Floating Structures (2000) BS 6349 - 7: Maritime Structures: Part 7 – Breakwaters (2000) Thoresen, Carl A.: Port Designers Handbook – Recommendations and Guidelines (Thomas Telford 2003) 287 21.0 DESIGN AIDS: PILING The capacity of a piled foundation to support load is governed both by the structural strength of the pile itself and by the strength of the soil surrounding the pile shaft and the base of the pile. The structural strength of the pile is controlled using materials with known properties and the design is carried out using structural design principals assuming the pile to be a column subjected to vertical and lateral loading with the soil providing restraint. The capacity of the soil to carry the loads transferred to it by the pile is influenced by several factors, the most important of which are: .. . The soil type and its stress history The strength and stiffness of the soil The method of installation of the pile which results in changes to the stress regime and strength of the soil surrounding the pile 21.1 PILE CAPACITY TO RESIST COMPRESSIVE LOAD 21.1.1 PILE BEHAVIOUR UNDER LOAD The behaviour of a pile under load is a complex soil/ pile interaction problem and rigorous methods have not, as yet, been developed to model pile behaviour. Structural loads are generally imposed at the head of the pile. This load is transmitted along the pile shaft and transferred into the surrounding soil. At small loads, the transfer of load occurs almost entirely along the pile shaft in friction. With increased load, friction transfer approaches the ultimate resistance value while an increasing share of the load goes onto the base of the pile. Full pile shaft capacity is mobilised at relatively small deflections (less than 15 mm) while full pile base capacity is mobilised at relatively large deflections (approximately 10 % of the pile base diameter). The proportioning of load transfer into the pile shaft and pile base will depend on several factors, the most important of which are the pile type and geometry, method of installation, and soil profile. FIGURE 21.1.1 shows the idealised behaviour of a pile under increasing load. From start to point A the load is resisted almost entirely by friction on the pile shaft. Between A and B the friction still increases slightly but reaches a maximum, whereas the end-bearing resistance starts to build-up. If the load is removed at this stage the pile-head will recover to virtually its original position indicating an elastic behaviour. At point C, where the pile-toe deflection is approximately ten percent of the pile base diameter, the end-bearing resistance has reached its ultimate resistance value and the pile is on the point of failure. Small increase in load from C to D results in a large increase in the pile-head deflection as well as pile-toe deflection. On removal of the load the pile will not recover to its original position due mainly to the plastic deformation at the base of the pile. 288 FIGURE 21.1.1 Idealisation of Pile Behaviour The static calculation of a pile's ultimate capacity considers the contribution of pile shaft and pile base separately, and the basic equation for the ultimate pile resistance is given as: (21.1a) Qp = Qs + Qb The static method of calculation can be used for both soil displacement and replacement type piles while dynamic methods of calculation and analysis are limited to driven displacement type piles. See SECTION 21.4. EQUATION (21.1a) ignores the mass of the pile itself in contributing to the applied load, which is a valid assumption for normal applied loading and pile geometry. 289 21.1.2 STATIC CALCULATION OF PILE CAPACITY EQUATION (21.1a) outlines the procedure adopted in determining the ultimate pile capacity. The implicit assumption that the pile's shaft friction capacity (Qs ) and base end-bearing capacity (Qb ) are not interdependent, is valid for normal pile geometry. The evaluation of the ultimate shaft capacity (Q s) is carried out by integrating the pile/soil shear strength along the shaft using the equation: τ = ca + σn tan φs Where: τ ca σn φs = = = = (21.1b) pile/soil shear strength at a particular point on the shaft pile shaft adhesion normal stress between pile and soil angle of friction between pile and soil The ultimate base capacity is calculated using bearing capacity theory with the equation: Q b = A b (cuNc + σv Nq + 0.5γdNγ ) Ab = cu = σv = γ = d = Nc , Nq & N γ = Where: .. .. (21.1c) area of base cohesion of soil vertical stress in soil at pile base unit weight of soil pile diameter bearing capacity factors dependent on soil properties and pile geometry Calculation of pile capacity is divided into four main soil categories: Cohesive Soils Non-cohesive Soils c - φ Soils Rock A further classification based on the method of installation of the pile is generally used in the static method of calculation. Two broad methods of pile installation are considered: Soil Displacement Piles and Soil Replacement Piles. In this section, Soil Displacement Piles will be referred to as Driven Piles, and Soil Replacement Piles as Bored Piles. Values of strength and compressibility parameters for the various types of soil and rock are given in SECTION 3.3. 21.1.2.1 PILES IN COHESIVE SOILS For piles in clay, the undrained resistance is generally taken to be the critical value. The undrained shear strength cu is used for the calculation of both the ultimate shaft and base capacities. In stiff over-consolidated clays the drained rather than the undrained resistance may be the critical value and effective stress parameters can be used. 290 BASE CAPACITY The base end-bearing capacity of both driven and bored piles is given by the equation: Q b = Nc c u A b Where: cu = Ab = Nc = (21.1d) undrained cohesion at the pile-toe area of pile base bearing capacity factor generally = 9 for penetration of at least five pile diameters into the bearing stratum. Nc values up to 20 have been measured for driven piles with an expanded base. FIGURE 21.1.2 shows the variation of Nc with the depth of penetration after Skempton (1951). / FIGURE 21.1.2 Bearing Capacity Factors after Skempton (1951) 291 SHAFT CAPACITY EQUATION (21.1e) gives the ultimate pile shaft capacity in cohesive soils: Q s = ca As Where: ca As = = (21.1e) average pile/soil adhesion over pile shaft length surface area of pile shaft The calculation of the ultimate pile shaft capacity in clay is influenced by the nature of the cohesive soil, as well as the method of installation and type of pile. Driven Piles The traditional method of calculating the static pile load capacity uses the undrained pile/soil adhesion ca and the undrained shear strength c u which has been studied by several authors for driven displacement piles and is generally equal to or greater than unity for soft clays and decreases markedly with an increase in the undrained shear strength. The relationship giving the shaft adhesion factor α defined in EQUATION (21.1f) for varying shear strengths of clay is given in FIGURE 21.1.3. c α = ca (21.1f) u FIGURE 21.1.3 Pile Adhesion Factors after Tomlinson (1970) for Driven Piles in Clay New research at the Imperial College, London, on driven piles in clay and sand, has led to the development of a new design methodology which is presented under SECTION 21.1.3: STATIC CALCULATION USING IN-SITU TESTING. 292 Bored Piles The ultimate skin friction of bored piles is calculated using EQUATION (21.1e). The shaft adhesion factor α relating the pile soil adhesion ca and undrained shear strength cu has been extensively studied locally and abroad for both residual and transported clay soils. Values of α once again vary considerably, but α is generally between 0.2 and 0.8, with a trend of α increasing in value with a decrease in undrained shear strength. A tabulation of typical values is given in TABLE 21.1.1. If accurate values of α are required for the determination of pile capacity, pile testing will be required to determine the value for the particular site, or measured values for similar founding conditions used. TABLE 21.1.1 Typical Values of Factor α and Pile Adhesion c a Undrained Shear Strength cu (kPa) SPT ‘N’ Dutch Cone Point q c (kPa) α Factor Pile Adhesion (kPa) < 10 <2 < 150 1.0 < 10 10 - 20 2-4 150 - 300 1.0 15 20 - 30 4-6 300 - 450 1.0 25 30 - 40 6-8 450 - 600 0.9 32 40 - 50 8 - 10 600 - 750 0.8 38 50 - 60 10 - 12 750 - 900 0.7 41 60 - 70 12 - 14 900 - 1050 0.6 42 70 - 80 14 - 16 1050 - 1200 0.55 42 80 - 90 16 - 18 1200 - 1350 0.50 43 90 - 100 18 - 20 1350 - 1500 0.45 43 100 - 110 20 - 22 1500 - 1650 0.40 43 110 - 120 22 - 24 1650 - 1800 0.38 45 120 - 130 24 - 26 1800 - 1950 0.36 46 130 - 140 26 - 28 1950 - 2100 0.34 47 140 - 150 28 - 30 2100 - 2250 0.32 48 150 - 170 30 - 31 2250 - 2750 0.30 50 170 - 190 31 - 32 2750 - 3250 0.29 53 190 - 210 32 - 35 3250 - 3750 0.28 57 210 - 230 35 - 38 3750 - 4350 0.27 61 230 - 250 38 - 42 4350 - 5000 0.26 65 250 - 300 42 - 50 5000 - 6300 0.25 75 300 - 400 50 - 65 6300 - 8800 0.22 85 400 - 500 > 65 8800 - 12000 0.20 100 293 21.1.2.2 PILES IN COHESIONLESS SOILS The calculation of the capacity of piles in cohesionless soils is generally governed by the internal angle of friction of the soil φ’ as well as the method of installation and type of pile. BASE CAPACITY Driven Piles The ultimate base capacity of driven piles in cohesionless soils is given by the equation: Qb = Nq Po’ Ab Where: Nq = Po’ = Ab = (21.1g) bearing capacity factor given by Berezantsev et al (1961) vertical effective stress in soil at pile-toe level area of base The relationship of Nq to the internal angle of friction of the soil, φ’, is given in FIGURE 21.1.4. The internal angle of friction chosen should consider the soil density over a depth of four diameters above the pile-toe and one diameter below the pile-toe, as well as pile installation effects on the soil surrounding the base. The relationship of soil density, SPT ‘N’ value, and φ’ is given in SECTION 3.3. The relationship of φ’ to the bearing capacity factor has been given by Berezantsev et al (1961) in FIGURE 21.1.4. FIGURE 21.1.4 Bearing Capacity Factors in Cohesionless Soils after Berezantsev et al (1961) 294 Meyerhof (1959) published the following bearing capacity factors indicating the significant increase in base capacity that can be expected when one considers the increase in friction angle due to the ground improvement achieved by the bulbous base compaction. This is clearly shown in FIGURE 21.1.5. φ’ φ’ φ’ φ’ φ’ FIGURE 21.1.5 Bearing Capacity Factor Enhancement after Meyerhof (1959) Bored Piles The base capacity of bored piles in cohesionless soils is difficult to predict and cannot be relied upon below the water-table due to disturbance of the soil during pile installation. A low value of φ’ = 28° to 30° can be considered and EQUATION (21.1g) used. For normal pile design where soil disturbance is likely, the contribution of the base to the ultimate load capacity should be ignored. 295 SHAFT CAPACITY The ultimate pile shaft capacity of piles in cohesionless soils is given by the equation: Q s = 0.5 K s P’d tan δ A s Where: Ks P’d δ As = = = = (21.1h) coefficient of earth pressure vertical effective stress at pile-toe level angle of friction between pile and soil pile shaft surface area Driven Piles The estimation of Ks tan δ is presently not well defined and there is substantial evidence that, in a uniform soil profile, shaft friction reaches a limiting value at a critical depth Z c. Poulos (1980) simplified this approach by defining the critical depth Zc in terms of a maximum vertical effective stress at a depth Zc. Poulos then proposed a method of relating Z c / d (where d = pile diameter / breadth) to φo given in FIGURE 21.1.6(a). The value of φo is the soil angle of friction after pile installation and is related to φ‘ the soil angle of friction before pile installation as follows: φo = 0,75 φ’ + 10o (21.1i) Values of K s tan δ for driven piles are plotted in FIGURE 21.1.6(b). Tomlinson (1977) proposed a simplified approach in determining Ks tan δ and values are tabulated in TABLE 21.1.2. TABLE 21.1.2 Ks tan δ values after Tomlinson (1977) Pile Type δ Ks Low Relative Density High Relative Density Steel 20o 0.5 1.0 Concrete 0.75 φ‘ 1.0 2.0 Wood 0.67 φ‘ 1.5 4.0 Bored Piles Poulos proposed using φ° = φ’ in determining values of Ks tan δ for bored piles plotted in FIGURE 21.1.6(c). In determining the critical depth Z c Poulos proposed using φ° = φ’- 3° in calculating Z c /d values given in FIGURE 21.1.6(a). Touma and Reese (1974) used a similar approach but proposed using δ = φ’ and K s = 0.7 for bored piles. 296 FIGURE 21.1.6 Values of K s tan δ after Poulos (1980) 21.1.2.3 PILES IN C - φ SOILS Where the soil is a sandy clay, a clayey sand or a sand silt and there is appreciable frictional as well as cohesive characteristics, the pile capacity should be derived using both these characteristics. Where φ’ is less than 25° the soil should be considered as primarily cohesive and designed as a φ’ = 0 soil. Where c u is less than 30 kPa the soil should be considered as non-cohesive with c u = 0. The approach used in calculating the ultimate capacity of piles in c - φ soils follows the principles and methods noted above, where the adhesion (cohesive soils) and the friction (cohesionless soils) are calculated separately and added together in calculating the shaft friction. The end-bearing capacity is calculated using Terzaghi's (1967) coefficients Nc and Nq in the equation: Q b = A b[1.3c uNc + P’o(Nq - 1) + 0.4γdNγ] 297 (21.1j) 21.1.2.4 PILES IN ROCK The calculation of the pile capacity in rock is generally based on the unconfined compressive strength of the intact rock qa. BASE CAPACITY A number of approaches can be used in estimating the capacity of piles founded on or within a rock mass. Strong rock with an unconfined compressive strength qa > l00 MPa exhibits brittle behaviour, while weaker rocks exhibit plastic or ductile behaviour and large movements are required to mobilise the full pile capacity. Jointing of the rock mass will also reduce the ultimate capacity, with vertical and open jointing requiring careful attention. It is generally accepted that the ultimate base capacity of either driven or bored piles is between 4 and 11 times qa. The value of 5qa is regarded as a reasonable value for the base resistance in rocks where the effects of jointing are not significant and the intact rock strength governs the pile capacity . For piles socketed into bedrock the base resistance increases. Where the socket length to pile diameter ratio exceeds 2.0, bearing capacity failure of the pile-toe cannot be effected. The ultimate base resistance can exceed 20qa, with crushing of the rock occurring beneath the pile-toe. A conservative empirical approach to the load capacity of piles founded on rock can be used, and the specified allowable bearing pressures outlined in Codes of Practice can be adopted. Typical allowable end-bearing values of 0.5qa are often stipulated, which is far below those recommended above. The values given in these codes are often conservative and the Factor of Safety against failure is generally well in excess of 3.0. Typical values given in these Codes are tabulated in TABLE 21.1.3 for shallow foundations on rock. TABLE 21.1.3 Allowable Bearing Pressure on Rock Recommended End-bearing Values from CP2004 (1972) Type of Rock Allowable End-bearing Value on Unweathered Rock MPa Hard igneous and gneissic rocks in sound condition 10.0 Hard limestones and sandstones 4.0 Schists and slates 3.0 Hard shales, hard mudstone and soft limestones 2.0 Soft shales and soft mudstones 0.6 - 1.0 Hard sound chalk, soft limestone 0.6 Allowable Bearing Values from New York City Building Code (1968) Type of Rock Allowable Bearing Values MPa Hard sound rock 5.8* Medium hard rock 3.85* Intermediate rock 1.95* Soft rock 0.77 * These values can be increased by 10% of basic value for each 300 mm of embedment of foundation into sound rock which has not been loosened by blasting or other means, and provided the loaded area is below the adjacent rock surface. Increased bearing values must not be more than twice basic values. 298 SHAFT SOCKET CAPACITY When piles are socketed or driven into rock, load transfer occurs through pile/rock adhesion within the socketed portion. The skin friction mobilised in the socket is a function of the strength of the rock, the method of installation and the jointing of the rock mass. The ratio of ultimate skin friction to unconfined compressive strength for medium hard rock (10 MPa < qa < l00 MPa) is generally between 0.05 and 0.1. The effects of smear and roughness on the socket capacity requires careful consideration where high socket friction values are anticipated. For very soft and soft rock the ratio of the ultimate skin friction to unconfined compressive strength for bored piles is given in FIGURE 21.1.7. Since the ultimate capacity of piles founded on, or within rock, is generally not the governing factor, methods of design should be based on limiting pile-head movements to values acceptable to the proposed structure. Design methods for piles in weak rock have been developed by Williams (1980), in Australia, and Rowe (1987), in Canada. These methods of design are generally applicable to soft rock of the Karoo Supergroup, as well as soft rocks of Miocene and Cretaceous age occurring in southern Africa. FIGURE 21.1.7 Side Shaft Resistance Values after Williams (1980) 299 The rock mass factor can be estimated from the TABLE 21.1.4. TABLE 21.1.4 Rock Mass Factor j RQD % Fracture frequency/m Mass factor j = Em/Ei 0 - 25 15 0.2 25 - 50 15 - 8 0.2 50 - 75 8-5 0.2 - 0.5 75 - 90 5-1 0.5 - 0.8 90 - 100 1 0.8 - 1.1 The ultimate shaft friction can then be estimated from the equation: fs = α.β.qa (21.1k) 21.1.3 STATIC CALCULATION OF PILE CAPACITY USING IN-SITU TESTS The use of the Cone Penetration Test (CPT), Standard Penetration Test (SPT) and the Pressuremeter Test results to predict pile capacity have been used worldwide. Empirical relationships between in-situ test values and the ultimate base and shaft resistance values of piles in cohesive, non-cohesive soils and soft rock for both displacement and non-displacement piles have been developed locally, and in other parts of the world. The use of the pressuremeter test is limited in the southern African region and the application of this test in determining pile capacity will not be covered in this text. For details of its application, reference should be made to Baguelin et al (1978). 21.1.3.1 CONE PENETRATION TEST Recent developments in the study of pile behaviour at the Imperial College, London, has led to the development of the ICP design methodology based on the Cone Penetration Test (CPT). The full methodology is outlined in the publication described by Jardine et al (2005). NON-COHESIVE SOILS BASE CAPACITY The ultimate base capacity of driven piles in sands is related to the average point resistance value q c over a depth of 1.5 pile diameters above and below the pile-toe. The ICP design methodology relates the ultimate base resistance to the CPT point resistance as a function of the pile and cone diameters as follows: qb = qc [1 - 0.5 log (D/DCPT)] (21.1l) The above relationship for closed-ended driven piles is halved for fully plugged openended piles. The ultimate base resistance for unplugged open-ended piles is equal to the CPT point resistance with the base capacity developed on the annular area. FIGURE 21.1.8 provides a guide to predicting conditions of plugging for varying pile diameters. 300 - FIGURE 21.1.8 Plugging Criteria for Open Driven Tube Piles in Sand after Jardine et al (2005) 301 SHAFT CAPACITY The mobilised pile shaft friction reduces with distance from the pile-toe and is graphically represented by Tomlinson (2001), in FIGURE 21.1.9, using the ICP method of design given by Jardine et al (2005). - FIGURE 21.1.9 Shaft Friction Resistance versus q c after Tomlinson (2001) 302 COHESIVE SOILS The ICP method given by Jardine et al (2005) provides a detailed and somewhat complex method of calculating the ultimate base and shaft capacity of driven piles in clay soils. The procedures outlined below summarise the methodology, but detailed reference to the original publication should be made. BASE CAPACITY Driven Piles The ICP method proposes the following equation in calculating the ultimate base resistance: qb = α qc Where: qb qc α = = = (21.1m) ultimate base resistance average point resistance at pile-toe level coefficient dependent on pile type and loading as given in TABLE 21.1.5 TABLE 21.1.5 End-bearing Capacity Coefficient α Loading and End Condition of Tube α Undrained loading, closed-ended piles 0.8 Drained loading, closed-ended piles 1.3 Undrained loading, open-ended plugged piles 0.4 Drained loading, open-ended plugged piles 0.65 Undrained loading, open-ended unplugged piles 1.0 Drained loading, open-ended unplugged piles 1.6 SHAFT CAPACITY Diven Piles The ICP design method of estimating the ultimate pile shaft capacity is given by the following equation: Qs = πDΣτf.dz Where: Qs τf = = ultimate shaft capacity a function of a Mohr Coulomb failure rule is given by equation: τf = σrf tanσf D = (21.1n) pile shaft diameter 303 21.1.3.2 STANDARD PENETRATION TEST (SPT) The SPT test is outlined in SECTION 2.0. The SPT ‘N’ value is however the most widely used in-situ test parameter for predicting pile capacity and is less prone to refusal than the CPT. The test will generally provide a conservative estimate since poor execution of the test will generally under-predict the in-situ soil strength. Like the CPT test, the SPT ‘N’ value can be used to estimate the soil shear strength using methods outlined in SECTION 3.3. The methods outlined in SECTION 21.2 can be used with the shear strengths calculated from the ‘N’ value. Direct correlation with ultimate base and shaft capacity values has been put forward by several authors. NON-COHESIVE SOILS Meyerhof (1956) proposed the correlation of ‘N’ versus ultimate base capacity for driven piles as: Qb = 400 N’ Ab (kN) where Ab is in m2 Where: Qb = Ab = N’ = (21.1o) ultimate base capacity in kN area of base in m2 average SPT ‘N’ value above and below base Meyerhof proposed a similar method of correlation for shaft friction capacity but noted that the ultimate shaft capacity for low displacement piles (H-piles) is approximately half that of medium and large displacement piles. The following correlations for ultimate shaft friction capacity are proposed for average SPT ‘N’ over the pile shaft length = N Where: Qs As N = = = Q s = 4 N A s (kN) (full displacement) (21.1p) Q s = 2 N A s (kN) (low displacement) (21.1q) ultimate shaft capacity in kN pile shaft area in m2 average SPT ‘N’ value over length of shaft COHESIVE SOILS The ultimate base and shaft resistance values can be obtained from the SPT ‘N’ value by using the correlation of SPT ‘N’ to the undrained shear strength C u and applying the calculation methods outlined in 21.1.2.1. 21.1.3.3 APPROXIMATE DIRECT CORRELATIONS WITH SPT AND CPT For preliminary design of piles TABLE 21.1.6 outlines approximate methods for determining ultimate shaft capacity values for piles in non-cohesive and cohesive soils. Factors are tabulated for various pile types and these should be multiplied by the test value (qc in MPa) to obtain the ultimate shaft capacity in kPa. TABLE 21.1.7 outlines a similar procedure for preliminary estimation of ultimate base capacity values in kPa. It should be noted that for end-bearing values used in the tabulation it is assumed that the piles are founded a minimum of five pile base diameters into the founding horizon and the average test value is taken over a depth of four pile base diameters above and one base diameter below pile-toe. 304 TABLE 21.1.6 Factors for Calculating Ultimate Shaft Capacity Using In-situ Tests Pile Auger Auger U/S CFA Oscill. Precast Tube Test Franki Wet Shaft Franki Forum Forum Ram Wet Ram Shaft Shaft Shaft Piles in Non-cohesive Soils CPT qc 5 5 5 5 8 8 8 12 5 8 SPT ‘N’ 2.5 2.5 2.5 2.5 4 4 4 6 2.5 4 Max (kPa) 125 80 125 125 150 150 150 200 125 150 Piles in Cohesive Soils CPT qc 10 10 10 10 15 15 15 30 10 15 SPT ‘N’ 2.5 2.5 2.5 2.5 3.0 3.0 3.0 4.5 2.5 3.5 α 0.4 0.4 0.4 0.4 0.6 0.6 0.4 0.6 0.4 0.5 Max (kPa) 150 80 150 150 100 100 150 200 150 150 TABLE 21.1.7 Factors for Calculating Ultimate Base Capacity Using In-situ Tests Pile Auger Auger U/S CFA Oscill. Precast * 0.5qc * 300 * 0.5qc * 300 * 0.5qc * 300 * 0.5qc * 300 8000 8000 8000 8000 Tube Test Franki Wet Shaft Franki Forum Forum Ram Wet Ram Shaft Shaft Shaft Piles in Non-cohesive Soils CPT qc SPT ‘N’ Max (kPa) 1.0qc 1.0qc *** 1.2qc *** 1.2qc 15000 ** 1.0qc ** 400 ** 15000 ** 1.0qc ** 400 ** 15000 400 400 500 500 20000 15000 15000 Piles in Cohesive Soils CPT qc 0.45qc 0.45qc 0.45qc 0.45qc 0.45qc 0.45qc 0.60qc 0.60qc 0.50qc 0.50qc SPT ‘N’ 50 50 50 50 50 50 60 60 50 50 Nc 9 9 9 9 9 9 9 - 20 9 - 20 9 - 12 9 - 12 Max (kPa) 4500 4500 4500 4500 4500 4500 6000 6000 5000 5000 * Very low base resistance values are likely for bored piles below the water-table in cohesionless soils due to installation effects and the contribution of the base to the load capacity should be conservatively ignored. ** If the base of the tube cannot be sealed against water ingress the contribution of the base to the load capacity should be conservatively ignored. *** Meyerhof indicates that ultimate base capacity = 2qc can be achieved and values given are conservative. 305 21.1.4 CALCULATION OF PILE CAPACITY USING DRIVING FORMULAE One of the oldest methods of estimating the load capacity of driven piles is the use of semi-empirical driving formulae. Most of these formulae derive the ultimate pile capacity from the energy input during driving and the pile set. A suitable factor of safety is then applied to this ultimate capacity to arrive at a safe working load. Alternatively, a formula can be used to calculate the required set the pile must be driven to, so as to provide the required safe working load. The following four dynamic formulae are regarded as the most reliable: Hiley Qult = S + 0.5(C1 + C2 + C3) Where: λ e 1and + and: ) Cd (W+n2 Wp ) 1+ C d = 0.75 + 0.15 λe Cd (W + Wp ) λe = W = = = WHL AES2 ef WH [ ] (21.1t) 0.5 Qult = W(0.27 + 0.3H) C3 S+ 2 Qult = Wp = ef = W = H = L = A = Ep = S = C1 C2 C3 ) Wp S + 2efWHL AEp Cornfield (21.1r) (21.1s) ku = C d(1 + Qult = Danish Where: x Qult = WH kuS Janbu ku = C d(1 + ef WH for bottom driven piles. ultimate load capacity (kN) weight of the pile (kN) efficiency of the hammer blow weight of the hammer (kN) hammer drop (m) length of the pile (m) area of the pile shaft (m2) modulus of elasticity of pile shaft (kN/m2) set taken as the average penetration per blow over the last ten blows (mm) temporary compression of pile-head and cap (m) temporary elastic compression of the pile shaft (m) temporary quake of the ground (m) 306 (21.1u) Certain inaccuracies can occur with the calculation of pile capacity using a dynamic formula. Firstly, the quantum of energy available from the hammer and the efficiency of energy transfer from the hammer into the pile, are not known with any degree of accuracy unless measured with electronic equipment. Secondly, the effects of the stress history of the soil and changes in pore pressure during driving, are also not taken into account. The generation of positive pore pressures during driving will decrease the resistance to penetration temporarily, whereas the opposite effect is experienced when negative pore pressures are generated. The latter can result in an overestimation of pile capacity and, to avoid this risk, the set on a selected number of piles should be checked about 48 hours after driving. Despite their shortcomings, dynamic formulae are still widely used and the calculated set is a simple form of site control. For long slender piles, the Wave Equation provides a more accurate form of analysis. Piling Driving Analysis by Wave Equation The wave theory analysis provides a means of relating the ultimate pile capacity to the pile set by considering a stress wave transmitted down the pile shaft. A finite difference method of wave analysis was originally developed by Smith (1960). A computer is required to carry out the analysis in a reasonable period of time. There are several commercially available computer software packages which carry out this type of analysis and Bowles (1974) gives the Fortran code for such a program. With the wave equation analysis there are a number of input parameters, and the problem of determining these accurately for someone not experienced with the program, is a negative aspect. The parameters include characteristics of the pile, the hammer, the hammer efficiency, spring constants for the helmet packing materials, percentage of resistance provided by the pile-toe, the ground quake, soil spring constants, and damping constants. With the correct input, the program provides good results. 307 21.2 PILE CAPACITY TO RESIST UPLIFT LOAD Piles are an effective and economical means of providing resistance to uplift loads. The resistance to uplift forces can be generated purely by skin frictional forces along the pile shaft, or by forming base enlargements which anchor the pile at depth below the surface. Uplift Load Capacity and Deflections The uplift capacity of friction piles can be calculated using similar principles to compression piles outlined in SECTION 21.1. The frictional resistance in uplift is approximately 70% of that for compression loading for piles where D/B is less than 10 and increases to 100% where pile depth to base diameter D/B is > 20. Jardine et al (2005) showed that the effect of contraction of the pile shaft in tension loading provides a reducing effect on the radial stresses, confirming that some reduction in capacity can be expected. The uplift capacity of piles with base enlargement, such as under-reamed augered or DCIS piles with an enlarged base, can be calculated using the method proposed by Meyerhof and Adams (1968). This method of calculation is graphically illustrated in FIGURE 21.2.1 and given by EQUATION (21.2a) and (21.2b). For D<H Qu = πcBH + s(π /2) γ B D2Ku tan φ + W (21.2a) For D>H Qu = πcBH + s(π /2) γ B (2D - H) HKu tan φ + W (21.2b) Where the value of H, m and the shape factor s = 1 + mH/B are given in the following table. Values of Ku , the earth pressure coefficient in uplift, are shown in FIGURE 21.2.1. φ 20o 25o 30o 35o 40o 45o 48o HB 2.5 3 4 5 7 9 11 M 0.05 0.1 0.15 0.25 0.35 0.50 0.60 s(max) 1.12 1.30 1.60 2.25 3.45 5.50 7.60 The ultimate uplift capacity Qu should not exceed the sum of the ultimate shaft capacity and the ultimate bearing capacity of the annular ring calculated using methods outlined in SECTION 21.1. This value should be reduced for piles with shallow embedment. 308 φ FIGURE 21.2.1 Uplift Capacity in a c - φ Soil Pile-head movements due to uplift loading are small over the working load range of applied load since all the load is generally carried in skin friction. Rapid increase in pile-head deflection occurs as the pile reaches its ultimate capacity. For normal structures and loading it is not necessary to specifically check movements if the piles have been adequately designed. 309 21.3 PILE CAPACITY TO RESIST LATERAL LOAD The analysis and design of piles subject to lateral loading is a complex subject and a detailed design approach is only required for structures subject to significant horizontal and seismic loads. For normal structures where the lateral loads are a minor load case and often transient, a simplified approach is sufficient. General methods of analysis and design will be given and a simplified method of analysis will be outlined in detail. Lateral Load Capacity and Deflections Three methods of calculating the ultimate lateral load resistance of piles are outlined below. The working lateral load capacity can be obtained as follows: . . The ultimate lateral capacity can be calculated and divided by an appropriate factor of safety to establish the working lateral load capacity The pile shaft movements and pile-head deflections can be calculated for a range of horizontal loads and the working load established at a load where movements are within acceptable limits. The ultimate lateral load capacity of a pile is governed by the fixity of the pile-head and the relative flexibility of the pile shaft versus the soil stiffness. The pile can behave as a short rigid element or as an infinitely long flexible member. Broms (1964) has presented detailed methods of analysing the ultimate lateral capacity for both cohesive and non-cohesive soils as well as for free and fixed head piles. The failure mechanisms, soil reactions and distribution of pile shaft moments are shown graphically in FIGURE 20.3.1. As an alternative to Brom's method of analysis the method proposed by Brinch Hansen (1961) can be used where the ultimate soil resistance at a depth below the ground surface is given by: PH = Po’Kq + cu Kc Where: (21.3a) PH = ultimate lateral soil resistance Po’ = vertical effective overburden stress c u = undrained cohesion K c and Kq are factors given in FIGURE 21.3.2 which are a function of φ’ and a ratio of depth to pile diameter The soil resistance and point of rotation can be calculated iteratively by taking moments about the point of load application. The ultimate lateral capacity can be calculated statically by considering horizontal equilibrium, based on the point of rotation as calculated. 310 FIGURE 21.3.1 Pile Behaviour Under Lateral Load after Broms (1964) φ φ FIGURE 21.3.2 Lateral Resistance Factors after Brinch Hansen (1961) 311 A third more detailed method has been proposed by Wang and Reese [FHWA-RD-97130 (1998)]. This method shows that the values obtained from the Broms method are adequate for piles spaced less than 3 diameters apart when the piles act as a wall, but are conservative when considering piles at greater spacing. Three possible failure mechanisms are considered by the Wang /Reese equations for ultimate passive resistance of cohesionless soils. The first mechanism is a simple passive wedge failure in front of a single pile with no interaction with adjacent piles. The second mechanism considers the limit equilibrium failure surfaces interacting, for deep or closely spaced shafts. Thirdly, a plastic flow around the shaft is considered. The limiting value used in design is the least force obtained, considering all three mechanisms. To obtain the equations for cohesive soils refer to the original paper by Wang and Reese (1986). As an alternate to this approach, the pile can be modeled as a structural member, either fixed or free at the pile-head or pile-toe, with linear elastic springs forming the horizontal soil reaction. Values of linear horizontal sub-grade moduli are given in SECTION 3.3. No account of soil yielding or non-linear behaviour is taken into account with this method. FIGURE 21.3.3 shows the proposed model graphically. FIGURE 21.3.3 Euler Spring Model of Laterally Loaded Pile 312 A simple method of analysis for piles subjected to small horizontal loading, which have principally been sized for vertical loading, is to assume a point of virtual fixity at a depth below the ground surface and calculate the maximum pile-shaft moment, with the pile acting as a cantilever above the point of virtual fixity as shown in FIGURE 21.3.4. For free head piles Mu = H (e + Z f) (21.3b) For fixed head piles Mu = H/2 (e + Z f) (21.3c) Where: Zf Zf = = 3.0 m for loose sands and soft clays or 1.5 m for dense sands and stiff clays FIGURE 21.3.4 Simplified Model for Calculating Maximum Pile Shaft Moments The accurate calculation of deflection of a pile under lateral load is complex and can only be carried out by numerical methods. Sub-grade reaction theory should be used with caution as the value used for the springs are area dependent and hence a function of the spacing of the nodes used in the model. Various approaches to this analysis can be taken: .. .. Sub-grade reaction theory and ‘p - y’ or ‘z - t’ curves (MPile) Elastic continuum analysis described by Poulos et al (1980) Boundary element methods (Repute) 3-D Finite element methods (Plaxis/FLAC) If accurate deflection predictions are required, a detailed site investigation and in-situ test programme should be implemented to obtain the necessary soil stiffness parameters. A full scale load test by jacking two piles apart can be relatively simple and inexpensive, and should also be considered where movements are critical. 313 21.4 THE DESIGN OF PILES FOR HEAVING SUBSOIL CONDITIONS Heaving clays and the problems associated with them are covered in detail in SECTION 18.1. The foundation solutions for heaving conditions are also outlined in this SECTION. The capacity of piled foundations to resist the heave uplift forces induced by the movement of the soil within the expansive horizons is an integral part of pile design in the southern African region. The design methods are based on the assumption that the superstructure is fully isolated from the heaving subsoil and the method of analysis is based on the design theory proposed by Collins (1953) and confirmed by Blight (1984). The method of analysis and research was based on expansive alluvial clays in the Vaal River floodplain. Heave Uplift Forces Collins analysis was based on the methodology outlined in SECTION 21.1 for pile shaft resistance in cohesive soils. The pile soil shear strength is given by EQUATION (21.1b) where: (21.4a) τ = c’ + σn tan φs Collins research showed the φ s = φ’ and σn = K σ v with K varying between 0.5 and 2.0. For the Leeuhof clays Collins proposed to use K = 1.0 with φ’ measured in a drained test as the most appropriate parameters used in design. In addition Collins proposed that c’ = Cohesion measured in the drained test. To obtain the total heave uplift over the full depth of the heave profile EQUATION (21.1c) should be integrated over the heave zone of depth H. The nett uplift on the pile shaft for a loaded pile is given by the equation: P + T = 0.5πD (2c’H + KγH2 tan φ’) Where: P T D c’ φ K H γ = = = = = = = = (21.4b) compression load on pile (including pile weight) nett heave uplift force pile diameter drained cohesion soil friction angle (drained) coefficient of earth pressure depth of heave profile unit weight of expansive soil The solution of this equation results in a force distribution down the pile shaft illustrated in FIGURE 21.4.1(a). FIGURE 21.4.1(b) shows typical heave uplift forces for various depths of heave and pile diameter given by Collins for the Leeuhof clays. The nett heave uplift on a pile calculated using EQUATION (21.4b) must be resisted by the portion of the pile shaft anchored in the stable horizon underlying the heave zone of depth H. The minimum axial compressive load P, should incorporate only the dead load portion of the applied pile load in calculating the nett uplift force on the pile. 314 (a) (b) FIGURE 21.4.1 Heave Uplift Force on a Pile after Collins (1953) It should be noted that research into the applicability of the methods proposed by Collins to residual expansive soils has not been carried out and conservative assumption should be made particularly for deep heave profiles comprising stiff residual clays. Further research into pile heave uplift is needed to enhance the early research carried out by Collins. Pile Anchorage There are three methods for anchoring a pile in a stable stratum: .. . The formation of an underream below the heaving zone (not commonly used); The formation of a straight sided socket within the stable horizon; The formation of an enlarged base with a Franki or a Forum bored pile Underream The resistance of the underream can be based on the calculation of the bearing capacity of the annulus of the base enlargement as well as the shaft resistance of the soil between the base and the heaving zone. The combined resistance of the base and shaft must not exceed the shear resistance of the intact soil on the perimeter of the base enlargement over the height of the stable soil horizon above the base enlargement. The design methods for piles subjected to uplift loads outlined in SECTION 21.2 should be used to calculate the ultimate resistance of the socket or base enlargement to the maximum nett heave uplift forces. The principles noted above are graphically illustrated in FIGURE 21.4.2. 315 FIGURE 21.4.2 Capacity of Pile Base Enlargement The pile shaft should be designed to accommodate an ultimate uplift force of 1.5 times the nett heave uplift force calculated, and the piles should be reinforced to resist the nett uplift forces without excessive cracking of the pile shaft concrete. It should be noted that in heaving conditions axial compressive stresses in the pile shaft should be maximised in order to limit the tensile uplift stresses near the base of the heaving profile. For deep heaving profiles where the calculated nett heave uplift forces are excessive, the use of an isolation layer surrounding the pile shaft should be considered. The annular infill material used should be low strength material such as Vermiculite and the use of sand, which could densify with time, should not be considered. Methods of construction for the voids and infill materials are outlined in SECTION 7.6: AUGER PILES. Socket The capacity of the socket within the stable horizon for a straight shafted bored pile solution, should use the principles outlined for either cohesive soils or rock and outlined in SECTION 21.1. Enlarged Base The capacity of an enlarged base is calculated in the same manner as that for an underream. As the enlarged base on a Franki or Forum bored pile is formed in the ground, its diameter cannot be measured. The volume of compacted concrete in the enlarged base can, however, be estimated using the volume of concrete as measured loose in the skip, multiplying by the number of skips and applying a compaction factor of 0.85. From this the diameter of the enlarged base can be estimated, assuming the base is perfectly spherical. 316 21.5 FACTORS OF SAFETY The traditional approach in determining the Working Load Capacity of a pile is to consider this load to be a proportion of the estimated or measured Ultimate Load Capacity of the pile. This ratio is defined as the Factor of Safety and is generally chosen by the designer with a value of between 2.0 and 3.0. . . . The Factor of Safety chosen is required for the following reasons: To ensure pile-head differential and total settlements at serviceability conditions are acceptable; To allow for natural variations of the soil profile as well as the uncertainties of the calculation/installation method; To ensure the working stresses on the pile shaft are within safe limits for the methods of construction. Since the movement required to mobilise ultimate shaft capacity is small, the use of a lower Factor of Safety of 2.0 can be applied to the ultimate shaft friction capacity. On the other hand, large displacements are required to mobilise the ultimate end-bearing capacity, and a larger Factor of Safety of 3.0 should be used. An overall Factor of Safety of 2.5 is commonly regarded as an acceptable value for piles of average geometry carrying load in both shaft friction as well as end-bearing and has become the norm in southern Africa. In assessing the required Factor of Safety the designer should consider the type of structure supported by the piled foundation with particular reference to the likely settlements that can be tolerated by the structural frame and finishes. TABLE 21.5.1 gives indicative acceptable settlement values for various types of structure. Instead of using Factors of Safety to limit pile-head deflection, the designer can use pile modelling and work directly with deflection. As the model can accurately predict the full load/deflection curve, the designer can decide the working load of the pile at which the settlement of an individual pile is acceptable, taking into account any group effects. 317 TABLE 21.5.1 Typical Allowable Settlement of Structures Indicative Acceptable Settlement (mm) Type Very sensitive machinery Monumental buildings, Very heavy machinery, Grain elevators, Concrete storage bins, Water towers, Retaining walls Overhead cranes, Delicate equipment, Bridges, Hangars, Buildings > 10 storeys Simply supported bridges, Steel tanks, Docks and piers, Concrete and steel framed buildings < 10 storeys First sign of cracking in rigid panel walls Factories, Stores, Warehouses, Single storey buildings without rigid infilling, Highway structures Structural damage to frames Limiting Differential Total Total Angular Settlement Settlement Settlement Distortion under WL under WL under 2WL Nett Settlement after Removal of 2WL 1/750 1/600 3 12 36 12 1/400 3 6 18 12 1/300 6 18 36 24 12 36 48 36 1/300 1/200 1/500 318 21.6 LIMIT STATE DESIGN IN GEOTECHNICAL ENGINEERING Geotechnical design in southern Africa has traditionally been carried out using working load design methods. These methods use unfactored values for loads and resistances. Provision for safety is made by way of a global Factor of Safety (FOS). In general terms, the design is considered satisfactory if the following condition is satisfied: Loading < Resistance FOS (21.6a) In Limit State Design, the two main limit states considered are the serviceability and ultimate limit states that control the functionality and strength of the structure respectively. Other limit states may also have to be considered such as accident, overall stability, durability, etc. In the verification of the ultimate limit state, partial factors are applied to actions (loads or deformations), material properties and /or resistances to obtain design values for the effect of actions Ed and for the corresponding resistance Rd. The design is satisfactory if: Ed < Rd (21.6b) The advantage of this approach is that the partial factors can be chosen to reflect the uncertainty attached to a particular parameter, rather than lumping the provision for safety into a single Factor of Safety. The intention is that a reasonably consistent level of reliability will be obtained for all loading conditions and material types. The European Code of Practice EN1997-1, Eurocode 7: Geotechnical Design - Design Rules, permits three approaches to the verification of the ultimate limit state by design calculation. These approaches have been described by Frank et al (2004) as (1) ‘action and material factor’, (2) ‘action effect and resistance factor’, and (3) ‘action effect and material factor’ approaches. In the latest version of SANS 10160: Basis of Design and Actions for Buildings and Industrial Structures, the use of Design Approach (1) ‘the action and material factor’ approach, was being advocated. For most routine designs, this approach involves verification of two ultimate limit states by means of independent sets of calculations. The first limit state, known in SANS 10160 as the STR limit state, will frequently control the design of the structure while the second, the GEO limit state, often controls failure in the ground. A variation of the STR limit state, known as the STR-P limit state, is used for self-weight dominated structures. For typical limit equilibrium type calculations (such as bearing capacity, earth pressure or slope stability), the input variables used in the calculation (eg actions, material properties, etc) are factored in accordance with the partial action and material factors given in TABLE 21.6.1. SANS 10160 is not a geotechnical design code but a ‘basis of design’. As such, it does not deal directly with the design of geotechnical structures such as foundations, slopes, retaining walls or piles, but makes reference to the relevant sections of Eurocode 7. 319 Eurocode 7 gives three basic methods for the design of axially loaded piles. The three methods are based on load testing of piles (static or dynamic), analysis of pile driving records (driving formula or wave equation analysis), and calculations using ground test results. In the first two methods, the characteristic value of the compressive resistance of the pile is determined from the mean of the measured or estimated pile capacities and the minimum pile capacity, each divided by a correlation factor. The value of the correlation factors depends on the type of test and the number of piles tested. When calculating the resistance of a pile using ground test results, the shaft and base resistance of the pile are calculated by means of recognised analytical models, using unfactored soil strength parameters. The resistance obtained is divided by the appropriate resistance factor given in TABLE 21.6.1 and a model factor (> 1.0). The model factor is introduced to ensure that the calculated resistance of the pile agrees with the resistance obtained from static load tests to be carried out on the site. The value of the model factor will vary depending on the calculation method used, the site conditions, and the type of pile. No typical values are given in Eurocode 7. Clearly the intention is that static load tests should be used as the ultimate verification of pile capacity. Partial Action Factors TABLE 21.6.1 Partial Factors STR STR-P GEO 1.2 0.9 1.35 0.8 1.0 1.0 1.6 / 1.3 (1) 1.6 ψ l (3) 0 1.0 0 (2) 0 1.25 1.3 ψ l (3) / 0 (1) 0 1.0 1.0 1.0 1.0 1.0 1.0 1.0 1.0 1.0 1.0 1.25 1.25 1.4 1.4 1.0 ACTIONS Permanent Actions Unfavourable Favourable Variable Actions Leading – Unfavourable Accompanying – Unfavourable All – Favourable Partial Material and Resistance Factors SOIL PARAMETERS Angle of shearing resistance (4) Effective cohesion Undrained shear strength Unconfined strength Weight density φ’ c’ cu qu γ RESISTANCES Pile base driven / bored / CFA Pile shaft – compression Pile combined – comp driven/bored/CFA Pile shaft – tension Pre-stressed anchors 1.0 / 1.25 / 1.1 1.0 1.0 / 1.15 / 1.1 1.25 1.1 1.0 / 1.25 / 1.1 1.3 / 1.6 / 1.45 (5) 1.0 1.3 (5) 1.0 / 1.15 / 1.1 1.3 / 1.5 / 1.4 (5) 1.25 1.6 (5) 1.1 1.1 (5) NOTES: (1) Values apply to variable actions other than wind, and wind respectively (2) For the STR-P combination, only permanent actions and the leading variable action are combined. Accompanying variable actions not considered (3) ψ l is an action combinations factor (4) Factor applies to tan φ’ (5) Resistance factor applied to pile and anchor capacities are calculated using unfactored ground parameters. In the case of unfavourable actions on piles (eg due to downdrag or transverse loading on the piles) the resistance factor is applied to the actions calculated using factored ground parameters 320 This description of Limit State Design in geotechnical engineering is a highly abbreviated summary of a complex subject and is not intended for use by geotechnical designers. Reference should be made to the following publications for further guidance: . . . . . EN 1990:2002. Eurocode – Basis of Structural Design, European Standard. European Committee for Standardisation, Brussels. EN 1997-1:2004. Eurocode 7: Geotechnical Design – Part 1; General Rules, European Standard. European Committee for Standardisation, Brussels, Belgium. Frank, R., Bauduin, C., Kavvadas, M., Krebs Ovesen, N., Orr, T. and Schuppener, B. (2004). Designers’ Guide to EN 1997-1, Eurocode 7: Geotechnical Design – General Rules. Thomas Telford, London, England. Orr, T.L.L., (Editor) (2005). International Workshop on the Evaluation of Eurocode 7, Proceedings. Trinity College, Dublin, Ireland. SANS 10160 (Draft - 2007). Basis of Structural Design and Actions for Buildings and Industrial Structures, Code of Practice. South African Bureau of Standards, Pretoria, South Africa. 321 21.7 ANALYSIS AND DESIGN OF PILE GROUPS The analysis of individual pile forces and moments in a pile group subjected to a combination of loads can be simplified using elementary statics, or made highly complex using an elastic continuum finite element analysis. The choice of the method of analysis will depend on the type and requirements of the structure, the nature of the loading, and the knowledge of the sub-surface condition and soil parameters. Basic Principles No matter how complex the analysis, the basic principle of static equilibrium must be satisfied and it is always good practice to perform a static equilibrium check on all methods of analysing the pile group. With the rapid advance of computational methods and the free availability of specialised computer software, it is generally more time consuming simplifying a pile group subjected to a variety of applied loads, than it is carrying out a detailed analysis using for instance a frame programme with spring supports. A basic static analysis is generally adequate for pile groups with the principal loading being axial, and considerations of load distribution, group action and settlement require careful consideration, rather than the analysis of individual pile forces and moments. Methods of estimating settlement of pile groups are given in SECTION 21.10. The load-carrying capacity of a pile group can be significantly less than the sum of individual pile capacities, particularly for friction piles with relatively close spacing or piles bearing on a relatively thin founding horizon. In the case of friction piles the group should be considered as acting as a block and the block capacity should be calculated using the method proposed by Terzaghi and Peck (1967) and shown in FIGURE 21.7.1. For pile groups with more than approximately 10 piles this check on group capacity should be carried out. There have been several reported cases of pile group failure where piles founded on a thin competent horizon have performed adequately for a single pile test, but the group capacity has been severely reduced due to a weak underlying horizon as shown in FIGURE 21.7.2. It should also be noted that pile-caps are generally designed as rigid members, thus ensuring equal settlement of individual piles in the group. Due to interaction of piles in a group the load distribution on the piles in the group is not equal and the outer piles will carry more load in sands, while the inner piles will carry higher load in clays. The usual assumption of equal load distribution for vertically loaded groups with a rigid pile-cap is still valid, since the overall factor of safety and settlement will be satisfactory if the individual piles are adequately designed. Bastile (2003) however highlights that a flexible pile-cap results in lower pile loads at the periphery than assumed in rigid models, and that using this can lead to more economical designs. Piled Rafts The advent of finite element software has also allowed the benefit of the complex interaction between the piles and rafts to be better understood and optimized. In piled raft design, piles are designed with their skin friction fully mobilized resulting in maximum settlement reduction where they are placed. The remainder of the load is designed to be carried by the raft, resulting in optimal use of the combined system. Methods of estimating the load-settlement behaviour of piled rafts have been presented by Poulos and Davis (1980) and Randolph (1994). 322 FIGURE 21.7.1 Pile Group Acting as a Block Foundation after Terzaghi and Peck (1967) (a) (b) FIGURE 21.7.2 Possible Failure of a Pile Group Founded on a Thin Bearing Stratum after Tomlinson (1977) Pile Groups Subjected to Combined Loading As noted, analysis of this type of group should be carried out with the use of a specialised computer program. If the analysis requires accurate estimates of pile group defections, as well as individual pile forces and moments, a finite element or nonlinear sub-grade reaction analysis is required with well-defined soil parameters for the elastic continuum. A typical program that can perform this type of analysis has been developed by Randolph (1989) and is called PIGLET. Reese (1987) has also developed a suite of programs using non-linear load transfer functions to model the soil reaction for both compressive and lateral loads. 323 FIGURE 21.7.3 Idealised Model of a Pile Group Using Frame Analysis For the routine analysis of pile groups where deflections are unlikely to require accurate assessment and the calculation of pile forces and moments is of primary concern, a three-dimensional frame analysis program, using Euler springs derived from horizontal sub-grade reactions to model the soil, is generally satisfactory. FIGURE 21.7.3 graphically illustrates a typical pile group idealisation using this type of analysis. The analysis can accurately model the geometry of the structure (including raking piles), and can easily accommodate a specified free-standing height (scour condition), but the group settlement effects are not directly taken into account, unless the spring reactions used are reduced to take group effects into account. The choice of modulus of sub-grade reaction values for the horizontal soil reaction is given in SECTION 3.3. The modelling of the vertical soil reaction should be based on the estimated or measured individual pile load/deflection characteristics over the working load range of the pile. Simplified hand computational methods of analysis were described in the first and second editions of this book, and should a suitable computer program not be available to the designer, reference to these methods can be made. 324 21.8 SETTLEMENT OF A SINGLE PILE AND PILE GROUPS SINGLE PILE SETTLEMENT Before the advent of modern computational methods, the analysis of the settlement of a single pile was based on empirical correlations or test results. With the advent of the computer, three approaches to the analysis of single pile settlement have been employed: .. . Load Transfer Function Methods Elastic Theory Methods Finite Element Methods Load Transfer Functions Piles can transfer load into the soil by means of shaft friction and end-bearing. A certain amount of movement of the pile is necessary for the full development of these two components. The empirical expression relating the resistance to the movement is known as a load transfer function. There are unique functions for both the shaft friction resistance, as well as the end-bearing resistance, for the range of pile types in various types of soil. Load transfer functions have been developed by Everett (1991), using a large database of pile test results carried out by Franki, on the principal pile types in various soil conditions throughout southern Africa. Typical shaft and base transfer function curves used in the analysis are shown in FIGURE 21.8.1(a) and (b) respectively. More recently the method developed by Flemming (1992) based on the original work by Chin (1970), has proven to be very reliable and accurately predicts the full load settlement curve. The ultimate skin friction of the pile and the ultimate end-bearing capacity are the required inputs, along with the base stiffness, centroid of friction, pile length and curvature factor. A typical plot is shown in FIGURE 21.8.2. The method is very useful in back-analysing pile test data and allows actual performance of the pile shaft and base to be evaluated. 325 (a) Shaft Load Transfer Function (b) Base Load Transfer Function FIGURE 21.8.1 Load Transfer Function after Everett (1991) 326 -- - -- - FIGURE 21.8.2 Prediction Pile Settlement Performance using Fleming (1992) 327 The elastic and finite element methods of analysis for single piles have been employed by several authors and reference should be made to Poulos (1986) for a full description and comparison of these methods. PILE GROUP SETTLEMENT The accurate prediction of the settlement of pile groups requires well defined soil parameters as well as knowledge of the rate and nature of the loading of the foundation. Many methods can be used in approaching the settlement analysis and it is prudent to use more than one of these methods to assess the likely range of predicted settlement values. The following methods can be used for calculating pile group settlement: .. .. Empirical relationship between single pile settlement and number of piles in group Equivalent raft method Pile interaction analysis 3-Dimension Finite Element analysis The general classification of soils into cohesive and non-cohesive is used for the calculation of group settlement. Empirical Method Skempton developed an empirical correlation of the ratio of pile group versus single pile settlement for various widths of pile group in sands, as shown in FIGURE 21.8.3. FIGURE 21.8.3 Pile Group Settlement in Sand after Skempton et al (1953) 328 Equivalent Raft Method The equivalent raft method of analysis can be utilised for cohesive, non-cohesive and layered profiles. The method comprises the assumption that the pile group effectively transfers its load into the subsoil as an equivalent raft at a depth D below the surface. End-bearing piles transfer all their load at the full depth of the pile. For fully frictional piles the assumption is made that the transfer of load is at a depth D = two thirds of the pile length. Where the piles are founded at a depth D below the top of a dense horizon with a compressible horizon overlying, the equivalent raft is assumed to be at a depth two thirds D below the top of the dense horizon. The load is assumed to spread at an angle of 1: 4 from the pile-head to the equivalent raft as illustrated in FIGURE 21.8.4. .. . . The equivalent raft settlement can be calculated in several ways: Use of semi-empirical methods Use of elastic solutions for settlement of footings Calculate strains of each soil layer for increased stress from equivalent raft load distribution The calculated settlement is corrected for depth of founding and rigidity FIGURE 21.8.4 Equivalent Raft Method after Tomlinson (1977) 329 Semi-empirical Methods The estimation of settlement of shallow foundations in non-cohesive materials is generally approached statistically or empirically. Burland et al (1977) related settlement to applied footing pressure for various widths of footing and soil density. FIGURE 21.8.5 illustrates the method graphically. Schmertmann (1970) used semi-empirical strain influence methods of estimation using CPT results, and this method should be used where CPT tests have been used as a method of investigation. More recently Burland and Burbridge (1985) proposed a method using a statistical approach. FIGURE 21.8.5 Settlement of Footings on Sand after Burland et al (1977) Elastic Solutions The settlement of the equivalent raft can be calculated directly from the applied pressure and estimated using elastic solutions for settlement of loaded areas outlined in Poulos and Davis (1974). The soil modulus (drained) chosen should be based on recommendations covered in SECTION 3.3. 330 Layer Strains The stress distribution beneath the equivalent raft can be calculated using Boussinesque or a two layer stress distribution similar to that for a circular footing comparatively outlined in FIGURE 21.8.6. The soil horizons up to a depth of four times the width of the equivalent raft can be assigned drained soil modulus values calculated using recommendations set out in SECTION 3.3. Settlement of each horizon can be calculated using the stress increase and soil stiffness assigned to that layer. These settlements are then summated to calculate the total equivalent raft settlement. FIGURE 21.8.6 Stress Distribution Beneath a Circular Footing after Poulos and Davis (1974) 331 Depth of Founding Correction The total equivalent raft settlements calculated, using methods outlined, should be corrected for depth of founding and rigidity of the pile-cap. A major advantage in the use of piled foundations over shallow foundations, is the reduction in settlement due to the transfer of the load well below the surface. Fox (1948) provided correction factors for this effect, and FIGURE 21.8.7 gives the correction factors for a Poisson's Ratio of 0.2, and varying values of the ratio of footing depth to footing width (Z/B). A rigidity factor of 0.8 over and above the Fox correction can be used for rigid pile-caps. FIGURE 21.8.7 Correction Factors after Fox (1948) 332 3-D FINITE ELEMENT METHOD A piled raft can be analysed using 3-Dimensional Finite Element methods incorporating the soil as an elastic continuum and the piles and reinforced concrete raft as structural elements. Careful modelling of the pile/soil interface is required. A typical Plaxis model is shown in FIGURE 21.8.8. MODEL SPACE - PILE GEOMETRY PILE DISPLACEMENT AND DEFORMATION FIGURE 21.8.8 3-D Finite Element Analysis (Plaxis) 333 21.9 STRUCTURAL DESIGN OF PILE SHAFTS PILE SHAFT DESIGN In South Africa there is only one code of practice which specifically relates to piling, although the implementation of EUROCODE 7 which comprehensively covers limit state design for piled foundations is summarised in Section 21.6. SABS 1200 F-1983: SABS Standardised Specification for Civil Engineering Construction F:PILING; which does not include pile design is the only applicable Code of Practice. SABS 088-1972 (as amended 1975, 1977 and 1980): South African Standard Code of Practice for Pile Foundations has been withdrawn and is obsolete. The structural design of pile shafts should be based on SABS 0100, Part 1, 1980. This code, which adopts the Limit State Design philosophy, does not specifically cover the design of piles and so a certain degree of interpretation is required. The design of pile shafts for both driven and bored piles must consider forces and stresses developed during handling and installation, as well as those imposed on the pile shaft by the supported structure. Care should be taken that bending moments and stresses induced on a precast concrete pile shaft during lifting and pitching should be checked, and the lifting points placed to minimise these moments and stresses. Care should also be exercised during driving, that excessive energy is not applied to the pile, causing cracking and spalling. Piles should be analysed as columns with varying degrees of fixity at the pile-head and toe. For normal soil conditions where the pile shaft is fully embedded and will remain so during its working life, the pile can be analysed as a stiff braced axially loaded column as given in Clause 3.5.3 of SABS 0100. Where the pile shaft is unsupported over a specified length below pile-cap level, as in marine structures or piles in riverbeds subject to scour conditions, consideration must be given to the slenderness and degree of fixity for the pile shaft. FIGURE 21.9.1 gives some guidance on the choice of fixity conditions that should be used for these structures. 334 - - FIGURE 21.9.1 Pile-head Fixity Conditions after Tomlinson (1977) 335 For normal pile groups where piles are fully embedded, the pile shaft should be designed as a short braced axially loaded column with allowance made for eccentricity due to construction tolerances and the ultimate axial load for a short column, which by nature of the structure cannot be subjected to significant moments, and shall not exceed the ultimate capacity N by: N = 0.4 fcu Ac + 0.67 Asc fy Where: fcu Ac A sc fy = = = = (21.9a) the characteristic strength of the concrete the area of the concrete the area of longitudinal reinforcement the characteristic strength of the compression reinforcement Where piles are in groups and are not subjected to significant moments, the ultimate axial load on the pile can be designed using EQUATION (21.9a). Where there is no moment in the pile shaft, the reinforcement can be nominal provided that the ultimate axial load ‘N’ does not exceed: N = 0.4 fcu Ac (21.9b) Nominal reinforcement can be regarded as 0.8% of the ‘required column area’ or 0.4% of the ‘actual column area’, whichever is the greater. Where there is no moment on the pile shaft but the axial load ‘N’ does exceed the above expression, the longitudinal reinforcement should not be less than 1.0% of the ‘actual column area’. Where piles are required to resist significant moments as well as axial loads, Clauses 3.5.5.1 and 3.5.5.2 of SABS 0100 cover this situation. These clauses are not quoted here because of their length and the fact that the formulae refer only to rectangular column sections. The design of circular column sections to resist combined moment and axial force is readily achieved using the tables published in Part 3 of the British Standard Code of Practice CP110. Alternatively, a detailed analysis can be carried out using the assumptions set out in Clause 3.3.5.1 of SABS 0100: Part I. FIGURES 21.9.1 to 21.9.6 give interaction curves for typical pile sizes and ultimate loads to determine the reinforcement required for ultimate applied forces and moments. A pile section that is designed for a bending moment combined with an axial tension force, should be checked for cracking and serviceability limits as per Clause 2.2.3.2 of SABS 0100, and applied where deemed necessary. Material Strength and Properties SABS 0100 states that the term ‘characteristic strength’ means the cube strength of the concrete or the yield or proof stress of reinforcement, unless otherwise indicated. The characteristic strength of concrete and reinforcing steel are given in TABLES 21.9.1 and 21.9.2. The design of the pile shaft should normally be designed on the concrete grade, and if required in exceptional circumstances the increased strength at time of load application can be used. 336 TABLE 21.9.1 Characteristic Strength of Concrete Cube Strength at an Age of 3 months 6 months 1 year (MPa) (MPa) (MPa) Grade Characteristic Strength fc u (MPa) 20 20.0 23 24 25 25 25.0 29 30 31 30 30.0 34 35 36 40 40.0 44 46 48 50 50.0 54 56 58 TABLE 21.9.2 Characteristic Strength of Reinforcement Designation of Reinforcement Nominal Sizes (mm) Characteristic Strength fy (MPa) Hot rolled mild steel (SABS 920) All sizes 250 Hot rolled high yield steel (SABS 920) All sizes 450 Cold worked high yield steel (SABS 920) Hard drawn steel wire All sizes 450 Up to and including 12 485 The design of the pile shaft should be based on the appropriate characteristic strength of reinforcement given in TABLE 21.9.2 or a lower value if necessary, to reduce deflection or control cracking. In the past it has been deemed prudent to reduce these stresses in certain cases for various reasons. With the pile shaft concrete for example there could well be a difference between the actual strength in the pile and that derived from test specimens due to: .. . Method of placement Contamination Limited, if any, mechanical compaction It is suggested that the characteristic concrete strength could be reduced by up to 10 MPa if such risks are present. A reduction of 5 MPa is recommended for cast-in-situ piles with a temporary casing, increasing to 10 MPa for deep cast-in-situ piles cast under water or bentonite. A reduction of up to 10 MPa should also be applied to CFA piles depending on the depth and pile diameter. It has not been common practice to reduce the characteristic strength of the concrete in precast piles. For similar reasons, the characteristic strength of the reinforcing steel should also be reviewed where such conditions are present. This is even more important if the piles are required to resist tension forces either in a temporary or permanent condition. It is suggested that a reduction of 30 percent be applied to the characteristic steel stress for piles that are subjected to temporary tension, and where the cover cannot be guaranteed, increasing to 45 percent if the tension is permanent. 337 With certain types of pile, a thick wall casing is a permanent feature of the final product. In southern Africa it is common practice that all or part of the permanent casing is used as reinforcement, provided that there is no aggressive groundwater and the level of the casing is below the oxygen replacement level. A reduction in thickness of 3.0 mm is normally allowed for corrosion. As a result of this, permanently cased piles often only have a top reinforcing cage. Where there is a risk of electrolytic action by stray electric currents, increased cover to the steel and high density concrete with high cement content are used to counter this problem. If this is not adequate then specialised techniques such as Cathodic Protection may have to be resorted to. Durability Due to the presence of aggressive groundwater conditions in many of our industrialised or mined areas, care should be exercised in the specification of the pile shaft materials and reference should be made to the Portland Cement Institute or papers covering this topic. In aggressive groundwater conditions crack widths in the pile shaft should be checked and minimised. Analysis of groundwater samples should always form part of the site investigation procedure where a piled foundation solution is envisaged. Applicable Norms As the subject of Pile Design is complex and there is ongoing development of theories and methodologies, not all aspects can be covered in one book. The reader can get more in-depth information on this topic at the following links / references: CIRIA R 144: Integrity Testing in Piling Practice (1997) CIRIA R 181: Piled Foundations in Weak Rock (1999) Frank et al: Designers Guide to EN 1997 – 1 Eurocode 2007: Geotechnical Design – General Rules (Thomas Telford 2006) Jardine, J. et al: ICP Design Methods for Driven Piles in Sands and Clays (Thomas Telford 2005) Tomlinson, M., Woodward, J.: Pile Design and Construction Practice, 5th Edition (Taylor and Francis 2008) 338 21.10 STRUCTURAL DESIGN OF PILE-CAPS The schedule of pile-cap geometries and reinforcement quantities given in TABLES 21.10.2 to 21.10.6 are intended for preliminary design and estimating purposes only and should not be used as a substitute for final design and detailing purposes. The design method used to obtain the dimensions and quantities is in accordance with SABS 0100 (1980). The methods used and the assumptions made in the design of the pile-caps are as follows: . . . . . . . .. Geometric Parameters Plan dimensions of the pile-cap were based on the pile spacing 'X', and an overhang of 150 mm; A tolerance of 75 mm in the plan position of any pile was used in determining forces and stresses within the pile-caps; In assessing pile-cap dimensions the pile diameters were calculated on the working loads tabulated and based on an average shaft stress of 6 MPa under these working loads; The column size was calculated from the pile loads using a characteristic concrete cube strength of 50 MPa with the contribution of the column reinforcement being ignored. A square column section was chosen. A high strength of concrete was chosen to minimise the column size and reduce the effect of column geometry on forces and stresses within the pile-cap; Cover to primary tension reinforcement assumed to be 75 mm Material Properties The design of the pile-cap reinforcement was based on the use of high tensile reinforcement throughout with the reinforcement having a characteristic strength of 450 MPa; The pile-cap design was based on a characteristic cube strength for the concrete of 30 MPa Loading .. All calculations were based on the pile working loads tabulated; Ultimate loads were calculated by multiplying the characteristic working loads by 1.45. This factor was chosen as an average of the factors of 1.4 and 1.6 for dead load and live load respectively. Dead loads generally form a higher percentage of the total working load than do live loads. This will not be applicable in many cases eg silos and will be incorrect if the live load acts in the opposite direction of the dead load; Only vertical axial loads have been used in calculating pile-cap forces and stresses; No applied horizontal loads or moments have been considered 339 . . Shear Stresses . Pile-cap depths were based on the critical shear force condition for no shear reinforcement calculated by considering all pile forces on one side of the centre line; Punching shear is not a critical condition for the pile spacings considered. A shear stress of υc = 0.35 N/mm2, increased by a factor 2d/a was used in determining the depth of the pile-caps. The use of the factor 2d/a is justified by the proximity of the load to the support. The dimension 'd' is the effective depth of the pile-cap and 'a' has been taken as the distance between the outside of the column face and a line 150 mm inside the pile centre. FIGURE 21.10.1 Shear Stress Calculation Criteria The depth of the cap was based on the following: The allowable stress υ = Where: . b d a = = = V bd and υc = 0.35 x 2d a width of cap effective depth of cap dimension shown in FIGURE 20.10.1 The pile-cap width 'b' was taken as the full width of the cap for all cases except the 3 pile group where 'b' was assumed to be the pile diameter + 450 mm. 340 Pile-cap Reinforcement The main pile-cap reinforcement resisting the bending moments and tie-forces was calculated using two methods: . Method 1 For 2, 3, 4 and 5 pile groups the truss theory was used which assumes the column load is transmitted to the pile by an inclined compressive thrust in the pile-cap with a corresponding tie-force in the reinforcement which maintains equilibrium as shown in FIGURE 21.10.2. FIGURE 21.10.2 Truss Theory The formulae outlined in FIGURE 21.10.4 were used for the 2, 3, 4 and 5 pile groups in determining the tensile force T between piles. 341 . . Method 2 For the 6 pile group the beam theory was used where the bending moments are calculated from the pile forces and the column load acting as an equivalent UDL acting at half the pile-cap depth; Even though no uplift, horizontal forces or moments have been considered in the calculation of pile-cap reinforcement, top reinforcement has been allowed for in the reinforcement quantities. Nominal reinforcement in both directions has been allowed so that a rigid pre-formed cage could be provided. Horizontal lacers on the vertical sides of the pile-cap have also been allowed. A typical detailed pile-cap is shown in FIGURE 21.10.3. TABLES 21.10.2 to 21.10.6 give details of pile-caps and required reinforcement. TABLE 21.10.1 gives reinforcement areas and masses for single bars and reinforcement matts. FIGURE 21.10.3 Typical Pile-cap Reinforcement 342 TENSILE FORCES ACROSS PILE-CAP BASED ON TRUSS THEORY N = ULTIMATE COLUMN LOAD FIGURE 21.10.4 Truss Theory Formulae 343 TABLE 21.10.1 Number 1 2 3 4 5 6 7 8 9 10 REINFORCING BAR CROSS-SECTIONAL AREAS IN mm2 6 mm 8 mm 10 mm 12 mm 16 mm 20 mm 25 mm 32 mm 28.3 50.3 78.5 113 201 314 491 804 56.6 101 157 226 402 628 982 1610 84.9 151 236 339 603 943 1470 2410 113 201 314 452 804 1260 1960 3220 142 252 393 566 1010 1570 2450 4020 170 302 471 679 1210 1890 2950 4830 198 352 550 792 1410 2200 3440 5630 226 402 628 905 1610 2510 3930 6430 255 453 707 1020 1810 2830 4420 7240 283 503 785 1130 2010 3140 4910 8040 40 mm 1260 2510 3770 5030 6280 7540 8800 10100 11300 12600 Spacing 50 75 100 125 150 175 200 250 300 REINFORCING MATT CROSS-SECTIONAL AREAS IN mm2 6mm 8mm 10mm 12mm 16mm 20mm 25mm 32mm 566 1010 1570 2260 4020 6280 9820 16100 377 671 1050 1510 2680 4190 6550 10700 283 503 785 1130 2010 3140 4910 8040 226 402 628 905 1610 2510 3930 6430 189 335 523 754 1340 2090 3270 5360 162 287 449 646 1150 1800 2810 4600 142 252 393 566 1010 1570 2450 4020 113 201 314 452 804 1260 1960 3220 94.3 168 262 377 670 1050 1640 2680 40mm 25100 16800 12600 10100 8380 7180 6280 5030 4190 Mesh Ref No 100 193 245 311 395 500 617 746 888 1042 289 341 433 517 655 772 943 1085 Wire Spacing (mm) Long Cross 200 200 200 200 200 200 200 200 200 200 200 200 200 200 200 200 200 200 200 200 100 200 100 200 100 200 100 200 100 200 100 200 100 200 100 200 STANDARD MESH FABRICS Wire Diam Area (mm) (mm/2m) Long Cross Long Cross 4.0 4.0 063 063 5.6 5.6 123 123 6.3 6.3 156 156 7.1 7.1 197 197 8.0 8.0 251 251 9.0 9.0 318 318 10.0 10.0 393 393 11.0 11.0 475 475 12.0 12.0 566 566 13.0 13.0 664 664 5.6 5.6 246 123 6.3 5.6 312 123 7.1 6.3 396 156 8.0 6.3 503 156 9.0 7.1 636 197 10.0 7.1 786 197 11 8 951 251 12 8 1131 251 Mesh Mass kg/m2 1.00 1.93 2.45 3.11 3.95 5.00 6.17 7.46 8.88 10.42 2.89 3.41 4.33 5.17 6.55 7.72 1.98 1.98 kg/m 6 mm 0.222 REINFORCING BAR MASS (kg) 10 mm 12 mm 16 mm 20 mm 25 mm 32 mm 40 mm 0.616 0.888 1.579 2.466 3.854 6.313 9.864 8 mm 0.395 344 Wire Mass kg/m2 Long Cross 0.50 0.50 0.96 0.96 1.22 1.22 1.55 1.55 1.97 1.97 2.50 2.50 3.08 3.08 3.73 3.73 4.44 4.44 5.21 5.21 1.93 0.96 2.45 0.96 3.11 1.22 3.95 1.22 5.00 1.55 6.17 1.55 9.43 7.45 10.85 8.87 345 Pile Load (kN) Column Load (kN) Column Size H (mm) x Pile Centres (mm) Cap Dimensions (mm) A B Cap Depth D (mm) Cap Volume (m3) Soffit Area (m2) Shutter Area (m2) Steel Mass (kg) Steel Mass / m3 (kg) Reinforcing B1 Layer B2 Layer TI Layer T2 Layer Steel Mass (kg) 300 600 200 750 1300 550 600 0.43 0.71 2.22 36 85 7Y12 6Y12 3Y12 6Y12 26 200 400 175 600 1150 150 500 0.32 0.63 1.70 30 106 5Y12 5Y12 3Y12 5Y12 21 1500 600 700 0.63 0.90 2.94 74 118 6Y16 6Y16 3Y16 6Y16 52 400 800 250 900 - 1550 650 700 0.71 1.01 3.08 79 112 7Y16 6Y16 3Y16 6Y16 55 500 1000 275 900 1550 650 750 0.76 1.01 3.30 91 121 8Y16 6Y16 4Y16 6Y16 64 600 1200 300 900 1850 700 900 1.17 1.29 4.59 125 107 9Y16 7Y16 4Y16 7Y16 87 700 1400 325 1100 1850 750 900 1.25 1.39 4.68 136 109 7Y20 7Y16 4Y16 7Y16 95 800 1600 350 1100 1900 750 950 1.35 1.42 5.03 142 105 7Y20 7Y16 4Y16 7Y16 106 900 1800 375 1100 1900 800 950 1.44 1.52 5.13 197 137 8Y20 7Y16 4Y16 7Y16 159 1000 2000 400 1100 2200 800 1050 1.85 1.76 6.30 171 93 9Y20 8Y16 4Y16 9Y16 128 1000 2000 400 1400 2200 800 1150 2.02 1.76 6.90 239 124 6Y25 6Y20 4Y20 8Y20 168 1200 2400 425 1400 TABLE 21.10.2 Pile-cap Details for 2 Pile Group 2300 900 1200 2.48 2.07 7.68 277 111 7Y25 8Y20 4Y20 8Y20 193 1400 2800 450 1400 2300 900 1250 2.59 2.07 8.00 274 106 7Y25 8Y20 4Y20 8Y20 206 1500 3000 475 1400 2500 900 1350 3.04 2.25 9.18 327 108 8Y25 9Y20 4Y20 8Y20 229 1500 3000 475 1600 2300 900 1250 2.59 2.07 8.00 291 112 8Y25 8Y20 4Y20 8Y20 203 1600 3200 500 1400 2500 900 1400 3.15 2.25 9.52 327 104 8Y25 9Y20 4Y20 9Y20 245 1600 3200 500 1600 2650 1050 1400 3.90 2.78 10.36 380 98 9Y25 10Y20 5Y20 10Y20 266 1800 3600 525 1600 2650 1050 1500 4.17 2.78 11.10 400 96 9Y25 10Y20 5Y20 10Y20 269 2000 4000 550 1600 346 Pile Load (kN) Column Load (kN) Column Size H (mm) x Pile Centres (mm) Cap Dimensions (mm) A B C Cap Depth D (mm) Cap Volume (m3) Soffit Area (m2) Shutter Area (m2) Steel Mass (kg) Steel Mass /m3 (kg/m3) Reinforcing B1 Layer B2 Layer B3 Layer TI Layer T2 Layer Lacers Steel Mass (kg) 300 900 250 750 1300 1200 400 500 0.60 1.20 2.17 50 64 3Y16 6Y12 10Y12 6Y12 6Y12 2Y12 38 200 600 200 600 1150 1075 400 450 0.44 0.98 1.76 41 74 4Y12 4Y12 8Y12 5Y12 5Y12 2Y12 31 1500 1375 450 550 0.87 1.58 2.74 93 82 5Y16 5Y16 10Y16 6Y16 6Y16 2Y16 70 400 1200 300 900 - 1550 1425 500 600 1.03 1.72 3.11 101 78 4Y20 5Y16 10Y16 6Y16 6Y16 2Y16 76 500 1500 325 900 1550 1425 500 650 1.12 1.72 3.37 126 88 4Y20 5Y16 10Y16 6Y16 6Y16 2Y16 95 600 1800 350 900 1800 1650 550 750 1.71 2.28 4.49 155 69 5Y20 6Y16 12Y16 7Y16 7Y16 3Y16 116 700 2100 400 1100 1850 1700 600 800 1.97 2.46 4.95 157 80 5Y20 6Y16 14Y16 7Y16 7Y16 3Y16 118 800 2400 425 1100 1850 1700 600 800 1.97 2.46 4.95 196 78 4Y25 5Y20 10Y20 7Y16 7Y16 3Y16 147 900 2700 450 1100 1900 1750 650 850 2.24 2.64 5.43 211 75 4Y25 5Y20 10Y20 7Y16 7Y16 3Y16 158 1000 3000 475 1100 2200 2000 650 1000 3.35 3.35 7.27 239 54 4Y25 6Y20 12Y20 8Y16 8Y16 3Y16 179 1000 3000 475 1400 2200 2000 650 1100 3.69 3.35 7.99 257 53 5Y25 8Y20 12Y20 8Y16 8Y16 3Y16 193 1200 3600 500 1400 TABLE 21.10.3 Pile-cap Details for 3 Pile Group 2300 2100 750 1100 4.16 3.78 8.43 290 55 5Y25 10Y20 14Y20 9Y16 9Y16 3Y16 218 1400 4200 550 1400 2300 2100 750 1150 4.35 3.78 8.82 388 70 6Y25 10Y20 14Y20 9Y16 9Y16 3Y16 291 1500 4500 575 1400 2500 2275 750 1250 5.44 4.35 9.52 378 53 6Y25 5Y25 8Y25 8Y20 8Y20 4Y16 284 1500 4500 575 1600 2300 2100 750 1150 4.35 3.78 8.82 390 70 6Y25 6Y25 10Y25 7Y20 7Y20 3Y16 292 1600 4800 600 1400 2500 2275 750 1300 5.66 4.35 10.75 423 57 6Y25 6Y25 10Y25 8Y20 8Y20 4Y16 318 1600 4800 600 1600 2650 2425 900 1300 6.62 5.09 11.53 512 61 6Y25 8Y25 12Y25 8Y20 8Y20 4Y16 384 1800 5400 625 1600 2650 2425 900 1300 6.62 5.09 11.53 526 63 7Y25 8Y25 12Y25 8Y20 8Y20 4Y16 395 2000 6000 650 1600 347 Pile Load (kN) 200 Column Load (kN) 800 Column Size H (mm) 250 x Pile Centres (mm) 600 Cap Dimensions (mm) A 1150 Cap Depth D (mm) 500 Cap Volume (m3) 0.66 Soffit Area (m2) 1.32 Shutter Area (m2) 2.30 Steel Mass (kg) 80.5 Steel Mass /m3 (kg/m3) 76.4 Reinforcing B1 Layer 10Y12 B2 Layer 10Y12 TI Layer 5Y12 T2 Layer 5Y12 Lacers 2Y12 Steel Mass (kg) 40.4 400 1600 350 900 1500 700 1.57 2.25 4.20 126 80.0 12Y16 12Y16 4Y16 4Y16 2Y16 101 300 1200 300 750 1300 600 1.01 1.69 3.12 89.5 88.2 8Y16 8Y16 8Y12 8Y12 2Y12 71.6 1550 700 1.68 2.40 4.34 139 82.8 14Y16 14Y16 4Y16 4Y16 2Y16 111 500 2000 375 900 - 1550 700 1.68 2.40 4.34 156 92.9 16Y16 16Y16 5Y16 5Y16 2Y16 125 600 2400 425 900 1850 800 2.74 3.42 5.92 225 92.1 20Y16 20Y16 5Y16 5Y16 3Y16 180 700 2800 450 1100 1850 800 2.74 3.42 5.92 275 100 14Y20 14Y20 7Y16 7Y16 3Y16 220 800 3200 500 1100 1900 850 3.07 3.61 6.46 324 105 16Y20 16Y20 8Y16 8Y16 3Y16 259 900 3600 525 1100 1900 900 3.25 3.61 6.84 326 100 16Y20 16Y20 8Y16 8Y16 3Y16 261 1000 4000 550 1100 2200 950 4.60 4.84 8.36 447 97 20Y20 20Y20 10Y16 10Y16 3Y16 357 1000 4000 550 1400 2200 950 4.60 4.84 8.36 462 100 14Y25 14Y25 7Y20 7Y20 3Y16 369 1200 4800 600 1400 TABLE 21.10.4 Pile-cap Details for 4 Pile Group 2300 950 5.03 5.29 8.74 591 118 16Y25 16Y25 8Y20 8Y20 3Y16 473 1400 5600 650 1400 2300 950 5.03 5.29 8.74 660 131 18Y25 18Y25 9Y20 9Y20 3Y16 528 1500 6000 650 1400 2500 1050 6.56 6.25 10.50 709 108 18Y25 18Y25 9Y20 9Y20 3Y16 567 1500 6000 650 1600 2300 1000 5.29 5.29 9.20 645 122 18Y25 18Y25 9Y20 9Y20 3Y16 516 1600 6400 700 1400 2500 1100 6.87 6.25 11.00 766 111 20Y25 20Y25 10Y20 10Y20 3Y16 613 1600 6400 700 1600 2650 1100 7.72 7.02 11.66 877 114 22Y25 22Y25 11Y20 11Y20 3Y16 702 1800 7200 750 1600 2650 1100 7.72 7.02 11.66 910 118 24Y25 24Y25 12Y20 12Y20 3Y16 728 2000 8000 800 1600 348 Pile Load (kN) Column Load (kN) Column Size H (mm) x Pile Centres (mm) Cap Dimensions (mm) A Cap Depth D (mm) Cap Volume (m3) Soffit Area (m2) Shutter Area (m2) Steel Mass (kg) Steel Mass /m3 (kg) Reinforcing B1 Layer B2 Layer TI Layer T2 Layer Lacers Steel Mass (kg) 300 1500 350 750 1600 600 1.54 2.56 3.84 89 57.9 20Y12 20Y12 10Y12 10Y12 2Y12 66.7 200 1000 275 600 1400 500 0.98 1.96 2.80 75 76.9 14Y12 14Y12 7Y12 7Y12 2Y12 56.6 1900 650 2.35 3.61 4.94 236 100 18Y16 18Y16 9Y16 9Y16 2Y16 176 400 2000 400 900 2000 700 2.80 4.00 5.60 271 96.9 20Y16 20Y16 10Y16 10Y16 2Y16 210 500 2500 425 900 - 2000 700 2.80 4.00 5.60 295 106 16Y20 16Y20 8Y16 8Y16 2Y16 229 600 3000 475 900 2250 850 4.30 5.06 7.65 404 93.9 18Y20 18Y20 9Y16 9Y16 3Y16 313 700 3500 500 1100 2300 900 4.76 5.29 8.28 414 86.9 18Y20 18Y20 9Y16 9Y16 3Y16 321 800 4000 550 1100 2300 900 4.76 5.29 8.28 455 90.5 20Y20 20Y20 10Y16 10Y16 3Y16 352 900 4500 575 1100 2350 950 5.25 5.52 8.93 500 95.3 14Y25 14Y25 7Y20 7Y20 3Y16 387 1000 5000 600 1100 2800 1050 8.23 7.84 11.8 651 79.1 16Y25 16Y25 8Y20 8Y20 3Y16 504 1000 5000 600 1400 2800 1100 8.62 7.84 12.3 727 84.3 18Y25 18Y25 9Y20 9Y20 3Y16 564 1200 6000 650 1400 TABLE 21.10.5 Pile-cap Details for 5 Pile Group 2900 1150 9.67 8.41 13.3 937 96.9 20Y25 20Y25 10Y20 10Y20 3Y16 726 1400 7000 725 1400 2900 1150 9.67 8.41 13.3 1025 106 22Y25 22Y25 11Y20 11Y20 3Y16 796 1500 7500 750 1400 3200 1250 12.8 10.24 16.0 995 77.8 22Y25 22Y25 11Y20 11Y20 4Y16 771 1500 7500 750 1600 2900 1200 10.1 8.41 13.9 925 91.7 22Y25 22Y25 11Y20 11Y20 4Y16 717 1600 8000 775 1400 3200 1250 12.8 10.2 16.0 1078 84.2 24Y25 24Y25 11Y20 11Y20 4Y16 836 1600 8000 775 1600 3350 1300 14.6 11.2 17.4 1235 81.5 16Y32 16Y32 16Y20 16Y20 4Y16 957 1800 9000 800 1600 3350 1350 15.2 11.2 18.1 1385 91.4 18Y32 18Y32 18Y20 18Y20 4Y16 1073 2000 10000 850 1600 349 Pile Load (kN) Column Load (kN) Column Size H (mm) x Pile Centres (mm) Cap Dimensions (mm) A B Cap Depth D (mm) Cap Volume (m3) Soffit Area (m2) Shutter Area (m2) Steel Mass (kg) Steel Mass / m3 (kg) Reinforcing B1 Layer B2 Layer TI Layer T2 Layer Lacers Steel Mass (kg) 300 1800 350 750 2050 1300 750 2.00 2.66 5.02 138 68.9 8Y20 9Y16 4Y16 5Y12 2Y16 107 200 1200 300 600 1750 1150 650 1.31 2.01 3.77 87.3 66.8 8Y16 10Y12 4Y16 5Y12 2Y12 67.7 2400 1500 900 3.24 3.60 7.02 237 73.3 7Y25 8Y20 4Y20 5Y16 2Y16 184 400 2400 400 900 2450 1550 1000 3.80 3.80 8.00 272 71.6 8Y25 9Y20 4Y20 5Y16 2Y16 211 500 3000 450 900 - 2450 1550 1050 3.99 3.80 8.40 294 73.3 9Y25 9Y20 5Y20 5Y16 3Y16 228 600 3600 500 900 2900 1800 1150 6.00 5.22 10.8 390 65.0 11Y25 12Y20 6Y20 6Y16 3Y16 302 700 4200 550 1100 2950 1850 1200 6.55 5.46 11.5 449 68.6 8Y32 9Y25 8Y20 6Y20 3Y16 348 800 4800 600 1100 2950 1850 1250 6.82 5.46 12.0 477 69.9 8Y32 9Y25 8Y20 6Y20 3Y16 369 900 5400 625 1100 3000 1900 1300 7.41 5.70 12.7 586 79.0 9Y32 10Y25 9Y20 8Y20 3Y16 454 1000 6000 650 1100 3600 2200 1450 11.5 7.92 16.8 782 68.1 10Y32 11Y25 10Y20 8Y20 4Y16 606 1000 6000 650 1400 3600 2200 1550 12.3 7.92 17.9 842 68.6 11Y32 13Y25 11Y20 8Y20 4Y16 652 1200 7200 725 1400 TABLE 21.10.6 Pile-cap Details for 6 Pile Group 3700 2300 1600 13.6 8.51 19.2 926 68.0 12Y32 14Y25 12Y20 8Y20 4Y16 717 1400 8400 775 1400 3700 2300 1650 14.0 8.51 19.8 1064 75.8 14Y32 14Y25 12Y20 8Y20 4Y16 824 1500 9000 800 1400 4100 2500 1750 17.9 10.2 23.1 1175 65.5 13Y32 16Y25 14Y20 8Y20 5Y16 910 1500 9000 800 1600 3700 2300 1700 14.5 8.51 20.4 1004 69.4 14Y32 15Y25 13Y20 9Y20 5Y16 778 1600 9600 850 1400 4100 2500 1800 18.4 10.2 23.7 1149 62.3 14Y32 16Y25 14Y20 9Y20 5Y16 890 1600 9600 850 1600 4250 2650 1850 20.8 11.3 25.5 1171 56.2 15Y32 17Y25 15Y20 9Y20 5Y16 908 1800 1080 900 1600 4250 2650 1900 21.4 11.3 26.2 1339 626 16Y32 18Y25 16Y20 9Y20 5Y16 1038 2000 12000 950 1600 350 351 352 353 354 355 22.0 .. .. DESIGN AIDS: SOIL IMPROVEMENT Improvement of soils is usually required for one of the following reasons: Reduce compressibility Increase shear strength Reduce permeability Prevent or reduce the risk of liquefaction The most common soil improvement technique is the compaction of soil using conventional compaction equipment. This type of compaction is usually carried out by earthworks contractors and does not form part of the services offered by Franki Africa. In-situ soil improvement using deep compaction methods is often an economical solution for sites with poor soil conditions. Details regarding the various deep compaction techniques that are offered by Franki Africa are given in SECTION 11.0: CLASSIFICATION OF SOIL IMPROVEMENT SYSTEMS, SECTION 12.0: SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS AND SECTION 13.0: TECHNICAL DETAILS OF SOIL IMPROVEMENT SYSTEMS. Design aids for the various systems will be given in this section. For all sites a detailed geotechnical investigation is a fundamental requirement for the evaluation and the design of a suitable soil improvement system. The requirements for such an investigation are summarised in SECTION 10.0: FACTORS INFLUENCING THE SELECTION OF A SOIL IMPROVEMENT SYSTEM. Reference should also be made to SECTION 2.0: GEOTECHNICAL INVESTIGATION and SECTION 3.0: SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS. 22.1 SOIL COMPACTION 22.1.1 VIBRATORY COMPACTION Vibratory Compaction using a vibrating immersion probe is usually only suitable for sands with a low silt and clay content. Details regarding acceptable soil profiles for this technique are given in SECTION 13.1. Vibratory Compaction of sands is usually carried out to increase the in-situ relative density. This in turn results in an improvement in shear strength and compressibility characteristics. The degree of improvement achieved is usually checked by carrying out post compaction tests using the CPT and SPT. In most instances it is possible to achieve post compaction relative densities of the order of 60% to 70%. Post compaction relative densities of up to 90% have however been reported by D'Appolonia (1953). The degree of improvement that can be achieved is dependent on the grading of the material that is being compacted and the spacing of compaction points. The actual spacing is best decided upon by carrying out test compaction patterns and monitoring the results using CPT and SPT tests. Correlations between CPT and SPT values and relative density are given in TABLE 22.1.1. These correlations can be used to measure the degree of improvement that has been achieved. 356 In foundation design a maximum allowable bearing pressure of 250 kPa is usually applicable. The correlations given in TABLE 22.1.1 between CPT and SPT values, modulus of compressibility, and relative density, can be used as a guide to determine the required degree of improvement for foundation design. A further consideration in foundation design is the depth of improvement that is required. For most structural and civil developments, improvement should be carried out to a depth of at least twice the breadth of the foundations or loaded areas associated with the development. If improvement to the required depth is not achieved, the compressibility characteristics of the soils below the improved zone need to be taken into consideration in the evaluation of the performance of foundations or other loaded areas. TABLE 22.1.1. Correlation between SPT and CPT and Sand Properties — Some Values Taken from Michell and Katti (1981) Sand Density Parameter Very Loose Loose Medium Dense Dense Very Dense SPT ‘N’ Value <4 4 - 10 10 - 30 30 - 50 > 50 CPT Cone Point Resistance qc (MPa) < 1.5 1.5 - 3 3 - 10 10 - 15 > 15 Equivalent Relative Density (%) < 15 15 - 35 35 - 65 65 - 85 85 - 100 φ’ degrees < 30 30 - 32 32 - 25 35 - 38 > 38 Modulus of Compressibility (MPa) < 10 10 - 15 15 - 30 30 - 45 > 45 D'Appolonia (1970) indicates that under most circumstances the risk of liquefaction reduces substantially when the relative density of sands exceeds 50% to 60%. The correlations given in TABLE 22.1.1 can be used as a guide to determine the degree of improvement required in instances when vibratory compaction is being used to reduce liquefaction potential. 357 22.1.2 DYNAMIC COMPACTION Most soil types, with the exception of soft silts, clays and peats can be compacted using dynamic compaction. The process is being increasingly used to improve sites which have been backfilled with general rubble which often includes large boulders and inorganic waste materials. In many instances it would not have been possible to economically develop these sites without the benefit of the dynamic compaction process. The general methodology for determining the depth to which treatment can be carried out is described in SECTION 13.2. As indicated in SECTION 13.2, the depth of compaction that can be achieved, is a function of the mass and diameter of the pounder, and the height of the drop. The current resources available to Franki Africa enable treatment of soils to maximum depths of 12 metres. Up to now this has been found to be suitable for most applications where dynamic compaction is required. The relationship, of applied energy to the depth and compaction, is given in Figure 22.1.1. The following aspects need to be taken into consideration in deciding on the required depth of treatment: . . In naturally deposited or residual soils, compaction is either carried out to improve the total thickness of the layer requiring treatment, or the depth of compaction is limited to 1.5 to 2.0 times the breadth of any foundation system or loaded area. Under certain circumstances it may not be possible to achieve either of these requirements. It is then necessary to take the compressibility, shear strength and permeability of the soils below the improved zone into consideration in the evaluation of the behaviour of the civil or structural development constructed on the dynamically compacted area; Due to the highly compressible and collapsible characteristics that usually occur within loose unconsolidated fills, it is usually necessary to treat the full depth of fill. With deep fills (greater than 8.0 metres) consideration can be given to forming a raft of compacted material, to limit differential settlements, due to consolidation or collapse of the underlying unimproved fill. This procedure is usually carried out below access roads, parking areas and industrial floors, but can also be considered for lightly loaded, settlement tolerant structures. It is important to emphasise that any decisions with regard to the required depth of compaction can only be made in association with a detailed geotechnical investigation. Reference should be made to SECTION 2.0: GEOTECHNICAL INVESTIGATION in this regard. Although the majority of soil types can be treated with dynamic consolidation, the compressibility characteristics that are achieved with the process vary considerably. The values given in TABLE 22.1.2 should be used as general guidelines in this regard. 358 FIGURE 22.1.1. Typical Depth of Influence / Energy Requirements for Dynamic Compaction TABLE 22.1.2. Typical Characteristics of Material Improved by Dynamic Compaction Type of Material Allowable Bearing Pressure (kPa) Modulus of Compressibility (MPa) Anticipated Total Settlement of Typical Foundations (mm) Well Graded Gravel and Rockfill 250 60 - 100 5 - 10 Sandy Gravels 200 30 - 50 5 - 15 Silty Sands 150 20 - 40 15 - 20 Clayey Sands and Silts 100 - 150 10 - 25 20 - 25 Waste Materials: Tailings, Builders Rubble. Inorganic Waste 75 - 150 10 - 30 20 - 30 359 The variation of allowable bearing capacity with cone resistance or SPT ‘N’ value, after compaction, can be estimated from the formula based on Schmertmann’s work (1978): qallowable = 0.09 Where: qc B = = qc B (22.1a) the cone resistance in MPa and the foundation width in metres Noting that N = 2.5qc gives the allowable bearing pressure in terms of SPT ‘N’ as: qallowable = 0.036 N B Where: (22.1b) qallowable is in MPa, for a foundation settlement of 25 mm Bearing pressures are reduced pro rata for settlements less than 25 mm. Dynamic compaction has been used to increase the density of sands and soils in order to reduce the risk of liquefaction. The guidelines given in SECTION 22.1 in this regard, can also be used for dynamic compaction. The nature of the process is such that the generation of vibration during compaction is inevitable. This is an important factor to take into consideration when working in developed areas. Advisable maximum levels for peak particle velocity due to ground vibrations are given in TABLE 22.1.3. Experience from many dynamic compaction projects has shown that peak particle velocities greater than 25 mm / sec are only exceeded under unusual circumstances. Using the correct techniques it is therefore possible to carry out dynamic compaction as close as 3 metres from underground services and 5 metres from sound structures. From TABLE 22.1.3 it is apparent that very low vibrations can cause annoyance to humans. This is an important consideration in developed areas since the reaction of people to vibration is often unpredictable. Quality control measures must be carried out to check the degree and depth of improvement that is being achieved. Control testing may be divided into three types: .. . Production Environmental Specification Typical patterns of improvement that can be expected are shown in FIGURE 22.1.2. 360 FIGURE 22.1.2 Typical Compaction Improvement after Berry et al (2004) 361 Production control includes quality assurance aspects such as pounder penetration tests, keeping detailed records of energy levels, and elevation surveys of the working surface. Dynamic Probe Super Heavy Tests (DPSH) are also used to obtain a qualitative assessment of the effectiveness of the process and to monitor possible changes in characteristics of the material being treated. Environmental control consists of measuring ground vibration levels and carrying out procedures to limit the effects that the process may have on adjacent properties. Specification or verification controls are carried out both during and after treatment is completed to certify that the objectives of the treatment have been achieved. Procedures used in this regard are described in SECTION 2.0: GEOTECHNICAL INVESTIGATIONS and include penetration tests (DPSH, SPT and CPT), pressuremeter tests and plate load tests. TABLE 22.1.3 Maximum Peak Particle Velocities (Vibration) — from SAICE Code of Practice (1989) Lateral Support in Surface Excavations Maximum Peak Particle Velocity (mm/sec) Effect on People and Buildings 0.5 Threshold of human perception 5 Historical monuments 25 Limit for private dwellings in order to reduce disturbance to residents to a minimum 50 Limit for residential structure on good foundations 80 - 90 Level at which minor cracking can be expected 120 Maximum level for sturdy reinforced concrete structures 22.1.3 COMPACTION GROUTING Compaction grouting is a highly specialised process, which is usually carried out as a unique solution, for a specific problem. Only a few compaction grouting contracts have been carried out in southern Africa. The process has mainly been used to consolidate poorly compacted fills below surface beds, or adjacent to retaining walls. There are no clearly defined design guides for compaction grouting. The design phase generally forms part of the grouting process and comprises monitoring the degree of improvement being achieved, and adapting the grouting process as required. The degree of improvement that is required is dependent on the nature and function of the fill that is being improved. With compaction grouting it is generally only possible to obtain an average overall density of about 90% Mod AASHTO. In most instances compaction grouting is carried out in areas with difficult access and the monitoring procedures have therefore to be adapted to take this into consideration. The use of dynamic penetration tests have proven to be very useful in this regard, DPSH and DPL tests, as described in SECTION 2.0: GEOTECHNICAL INVESTIGATION. The DPL test is particularly useful because of the light versatile equipment required and available correlations between the test results and compaction characteristics, such as the California Bearing Ratio (CBR). Reference should be made to SECTION 3.0: SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS, in this regard. 362 22.2 SOIL REPLACEMENT Soil replacement is a technique whereby columns of gravel, rock or even builders’ rubble are installed into soft soils. These types of columns are collectively referred to as stone columns. Three different systems for the installation of stone columns, VIBRATORY REPLACEMENT, DYNAMIC REPLACEMENT AND DRIVEN STONE COLUMNS are described in SECTIONS 13.4, 13.5 and 13.6 respectively. Although there are some fundamental differences in the types of equipment used, installation techniques, and nature of the stone columns that are installed by these three systems, they essentially fulfill the same function. Although stone columns are suitable for most soft soils their effectiveness and suitability in highly sensitive clays is open to question. This is due to possible effects of the installation process on the strength of the sensitive clays. In certain soil profiles excessive pore water pressures can result in liquefaction of the entire soil mass. Under these conditions the formation of stone columns will not be possible. Stone columns are often used to provide vertical support for structures or embankments. They can also be designed to resist shear and improve slope stability. 22.2.1 VERTICAL SUPPORT Stone columns to provide vertical support are in many ways similar to piled foundations except that pile-caps, structural connections and deep penetration into underlying competent strata is not necessarily required. On the other hand, stone columns do not provide the same rigidity of support as piled foundations. When used for vertical support of structures, or embankments, the load capacity and load-settlement behaviour of the stone columns is therefore of primary concern. The load capacity of a stone column is controlled by side shear and end-bearing capacity between the column and the surrounding soil, and also the passive resistance of the surrounding soil that can be mobilised to withstand radial bulging of the column. An evaluation of side shear and end-bearing capacities is carried out using conventional pile design procedures. These are described in SECTION 21.0: DESIGN AIDS: PILING. An analysis procedure based on cylindrical cavity expansion theory has been developed by Vesic (1972) to determine the ultimate capacity of stone columns in relation to the passive resistance provided by the surrounding soil. The procedures to be used in the analyses are given in FIGURE 22.2.1. An alternative method is given by Hughes and Withers (1974) and Thorburn (1975) in which the allowable vertical stress, σv , on a single column in a cohesive soil can be expressed by: σv = Where: cu F = = 25cu F the undrained shear strength of the surrounding soil a Factor of Safety, for which a value of 3 is recommended 363 (22.2a) In the calculation of the required load capacity of the stone columns the assumption is usually made that all applied vertical loads are carried entirely by the stone columns. This is a conservative assumption since vertical loads will in fact be shared between the stone columns and the surrounding ground in proportion to the relative stiffness of the two materials. Load sharing behaviour is usually taken into consideration in the evaluation of settlement of a composite stone column/soft soil foundation system. A test on a single stone column will usually give settlements of the order of 5 to 10 mm, which implies a modulus of compressibility of about 40 to 70 MPa. Experience and analyses, Mitchell and Katti (1980), indicates that settlement of a large loaded area supported by stone columns will be between 5 and 10 times greater than the settlement of a single stone column. The actual value will be dependent on the relative stiffness between the stone column and the surrounding soil, as well as the crosssectional area, spacing and depth of the stone columns. Where: [ 1 + sin φ’ ] [ ] 1 – sin φ’ss q = φ’ = ultimate shaft stress on the stone column effective angle of friction of surrounding soil effective angle of friction of the stone column which is usually 40° to 45° q = c’.Fc’ + q.Fq’ φ’s = 1 Ir = E 2(1 + υ)(c’ + q tan φ’) E = Modulus of compressibility of the surrounding soil c’ = effective cohesion of surrounding soil υ = Poisson’s ratio of surrounding soil F’c , F’q are cavity expansion factors FIGURE 22.2.1 Ultimate Capacity of Stone Columns in Relation to Passive Resistance of the Surrounding Soil, after Vesic (1972) 364 In order to estimate the settlement of a composite stone column/soft soil foundation Mitchell and Katti (1980) have defined a settlement reduction ratio β as: β= Where: ρ ρ‘ n = = = as As Ac = = = ρ 11 = ρ’ [1 + (n – 1)a s] (22.2b) the settlement of the soft soils without stone columns the settlement of the composite foundation system the ratio of vertical stress in the stone columns to that in the soft ground and falls in the range of 2 to 6, with values of 3 to 4 being usual a replacement ratio equal to A s /(A s + A c ) the cross-sectional area of the stone column the plan area of soft soil per stone column Greenwood (1970) has shown that β does not usually exceed 0.75 even for widely spaced stone columns, and can be as low as 0.1 for very closely spaced stone columns. In carrying out a load-settlement analysis using the procedures given above, it is necessary to add any anticipated settlement from strata underlying the stone columns to ρ’, the estimated settlement of the composite stone column/soft soil foundation. Drainage through stone columns can also accelerate settlement and this may be an important aspect in certain applications, see SECTION 22.3: ACCELERATED CONSOLIDATION. 365 22.2.2 SHEAR RESISTANCE The shear resistance provided by stone columns can be used in slope stability applications. This is illustrated in FIGURE 22.2.2. τ = (1- as ) τc + as τs cos α Where: τ Pz = = = = = φ’s = as τc τs (22.2c) Composite shearing resistance along the failure surface Replacement ratio as defined in SECTION 22.2.1 Shear strength of soft in-situ soils Shear strength of stone column = Pz tan φ’s cos α Average vertical stress on the stone columns along the sliding surface Angle of friction of the stone column usually 40° to 45° FIGURE 22.2.2 Stone Columns for Use in Slope Stability Applications after Mitchell and Katti (1980) In order to evaluate the stabilising effect of the stone columns, it is necessary to make some assumptions with regard to the composite shear strength along any potential failure surface. Mitchell and Katti (1980) describe a number of procedures in this regard. One of these procedures is given in FIGURE 22.2.2. 366 22.3 ACCELERATED CONSOLIDATION Consolidation and strengthening of soils under an applied static load is one of the oldest and widely used methods for soil improvement. In many cases, the time required for consolidation is excessive and vertical drains are used to accelerate the rate of consolidation. A further acceleration can also usually be achieved by combining vertical drains with surcharge loadings. Consolidation times vary in accordance with the square of the drainage path length. The reduction in drainage path due to installation of vertical drains is therefore the most significant aspect in accelerating consolidation. A further important aspect is that many soft and compressible soils have a greater permeability in the horizontal direction than in the vertical direction. Vertical drains are generally ineffective in organic clays, peats and other soils whose settlement behaviour is dominated by secondary compression. The theory for consolidation by radial drainage is well developed, Barron (1948). This theory allows an analysis to be carried out to estimate the vertical drain spacing required to achieve the desired degree of consolidation within the time available. The following expression is normally used in this regard. t= Where: t Th Ch D D D = = = = = = Th 2 D Ch (22.3a) elapsed time time factor for radial drainage coefficient of horizontal consolidation zone of influence of the drain 1.13 x drain spacing for a square grid 1.05 x drain spacing for a triangular grid The value of Th varies with the ratio of the zone of influence of the drain, D, to the equivalent drain diameter Dd relationship between Th , percentage consolidation and the ratio, D/Dd is given in FIGURE 22.3.1. In the case of sand or sandwick drains, the equivalent drain diameter is normally taken as equal to the nominal diameter of the drain. Although there is no general agreement on the subject, the equivalent diameter of band drains can be taken as the diameter of a circle having the same free surface as the band drain. The coefficient of horizontal consolidation, C h , is of critical importance to the design of a vertical drainage system. It is, however, an extremely difficult parameter to measure with any degree of accuracy using conventional laboratory testing techniques. It is generally accepted that laboratory measurements of the coefficient of consolidation using oedometer equipment will overestimate C h by at least an order of magnitude. 367 In recent years Rust and Jones (1990) have shown that the determination of pore water pressure dissipation time in the execution of piezocone (CPTU) tests, is the best available procedure for estimating C h. Even with this testing procedure, considerable judgement, coupled with a detailed knowledge of the soil profile, is required to evaluate Ch with any degree of accuracy. A further complicating factor in this regard is that the value of Ch immediately adjacent to the drain can be adversely effected by remoulding during the installation procedure. Experience has shown that this is usually more significant with sand or sandwick drains, than with band drains. FIGURE 22.3.1 Variation of Time Factor Th with Percentage Consolidation U for Various Ratios of n = D/Dd Applicable Norms As the topic of Ground Improvement is broad, not all aspects can be adequately covered in one book. The reader can get further details on this important topic at the following links / references: BR 458: Specifying Dynamic Compaction (2003) CIRIA C 572: Treated Ground Engineering Properties and Performance (2002) 368 23.0 DESIGN AIDS: LATERAL SUPPORT An evaluation of worldwide practice as presented as part of the International Symposium on Underground Construction in Soft Ground (1994) shows that semiempirical and limit equilibrium methods are the most commonly used for the design of lateral support systems. The use of finite element methods have only recently gained acceptance and were previously confined to research and very sensitive projects. Only the most commonly used semi-empirical and limit equilibrium methods will be considered in this section. The following three main aspects need to be considered with regard to support of excavations and slopes: .. . The type of soil or rock supported The type of lateral support system adopted The method used for design of the retention structure The type of soil or rock which is supported is related to the collection and interpretation of the geotechnical investigation data. Although recommendations in this regard are given in SECTION 2.0: GEOTECHNICAL INVESTIGATION and SECTION 3.0: SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS, certain specific recommendations with regard to design parameters for lateral support are also given in this section. The type of lateral support system and the method to be used for design are essentially interrelated. Design aids will therefore be given in this section for the various lateral support systems dealt with in SECTION 17.0: TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS. 23.1 GEOTECHNICAL DESIGN PARAMETERS It is recommended that the design of lateral support systems be carried out in terms of effective stress. It is therefore necessary to determine the effective cohesion (c’) and effective angle of friction (φ’) of the soil or rock horizons to be supported. These parameters should be determined from a suitably designed and executed laboratory or in-situ testing programme, which forms part of the overall geotechnical investigation process. Recommendations are given in SECTION 2.0: GEOTECHNICAL INVESTIGATION. For guidelines to evaluate effective strength parameters using methods other than the relevant specific laboratory tests, reference should be made to SECTION 3.0: SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS. It is important to emphasise that the effective cohesion (c’), often associated with cohesive soils, is an unreliable design parameter and its measurement in laboratory testing is often a function of the test and the non-representative nature of the sample being tested. It is therefore recommended that, for design purposes, the effective cohesion (c’) be taken as zero, unless it can be established that the material being supported is intact and not fissured. A non-zero value of effective cohesion should therefore be confirmed by an appropriate laboratory testing programme and on-site evaluation, taking due account of jointing, fissuring or slicken-siding of the material to be retained. In rock, the geometry of the failure surface and the most likely failure mechanism are usually controlled by discontinuities within the rock mass. 369 The shear strength properties of the intact rock material are usually of lesser importance than the properties of the discontinuities. In the design of a support system for rock, it is therefore important that sufficient information be obtained with regard to the orientation, spacing, continuity, roughness and shear strength of the discontinuities. Certain recommendations with regard to obtaining and evaluating these parameters are given in SECTION 3.0: SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS. Reference should also be made to Hoek and Bray (1977) and Barton and Chouby (1977). Usually limited information to obtain design parameters is retrieved from rotary cored boreholes, and any design assumptions made from borehole data must be checked by mapping of discontinuities during the excavation phase. Closely jointed rocks, having a rock quality designation (RQD) close to zero, are mostly conventionally treated as being composed of interlocking granular fragments with an effective angle of friction. Typical effective angles of friction for closely jointed rocks are given in TABLE 23.1.1. Where there is sufficient available information, usually in the form of back analysis of slopes and excavations, it may be appropriate to use a low value of effective cohesion in combination with the values given in TABLE 23.1.1. TABLE 23.1.1. Typical Effective Angles of Friction for Closely Jointed Rock Rock Type φ’ (degrees) Sandstones, quartzite or granite 40 Diabase/dolerite 38 Siltstone 35 Shale 30 Mudstone 25 A further important parameter is the density of the material to be supported. Guidelines given in TABLE 23.1.2 may be used in the absence of reliable test results. Refer to SECTION 3.3.1 for details of how to estimate the rock friction angles from UCS testing, if tri-axial testing of the rock mass is not available/feasible. TABLE 23.1.2 Typical Density Values Soil or Rock Type Moist Bulk Density (kN/m3) Saturated Bulk density (kN/m3) Well graded gravel and sand Loose: 17 Dense: 18.5 19 20 Silty sand Loose: 17 Dense: 18.5 19 20 Clays, silty clays or sandy clays Soft: 17 Stiff: 19 18 20 Very soft rock or soft rock 21 22 Medium hard rock or hard rock 22 23 370 23.2 EARTH PRESSURES The magnitude and distribution of earth pressure is fundamental to the design of a lateral support system. The theory of lateral earth pressures has been dealt with in many textbooks on soil mechanics, Winterkorn and Fang (1975), Lambe and Whitman (1979), Perloff and Baron (1976), and it is not the intention to cover this aspect in detail. For the purposes of this section consideration will be given to the following three lateral earth pressure conditions. 23.2.1 EARTH PRESSURE AT REST The horizontal effective stress that exists in a natural soil, in its undisturbed state, is defined as the earth pressure at rest. In terms of lateral support systems, at rest conditions are only realised in practice in the case of rigid retaining structures. For normally consolidated soils the coefficient of earth at rest Ko , is given by: Ko = 1 -- sin φ Alternatively: Where: υ = (23.2a) Ko = υ /(1 -- υ) can be adopted Poisson’s ratio after Tschebotarioff (1976) Ko is known to increase with the over-consolidation of the soil. In over-consolidated soils the following expression of Ko may be used: Ko = (1 -- sin φ’) (OCR) 0.5 Where: OCR = (23.2b) Over-consolidation ratio of the soil Typical values of Ko for various soil types are given in TABLE 23.2.1. TABLE 23.2.1 Typical Values for K o after Whitlow (1990) Type of Soil Ko Loose Sand 0.45 - 0.6 Dense Sand 0.3 - 0.5 Normally Consolidated Clay 0.5 - 0.7 Over-consolidated Clay 1.0 - 4.0 Compacted Fill 0.7 - 2.0 371 23.2.2 ACTIVE EARTH PRESSURE The active earth pressure is the minimum value of lateral earth pressure that a soil mass can exert against a yielding retaining structure. It represents the failure condition at which the shear strength of the soil is fully mobilised in resisting gravity forces. For most retaining structures the design of the support system will normally be satisfactory if the system is capable of resisting the active pressure with a suitable margin of safety. The following expression may be used to determine the active earth pressure coefficient Ka: Ka = 1 -- sinφ’ (23.2c) 1 + sinφ’ Typical values for Ka for various soil types are given in TABLE 23.2.2. TABLE 23.2.2 Typical Values for K a and Kp (ignoring wall friction) Type of Soil Ka Kp Loose Sands 0.33 - 0.4 2.5 - 3.0 Dense Sands 0.25 - 0.33 3.0 - 4.0 Clays with Low Plasticity 0.3 - 0.4 2.5 - 3.3 Clays with Moderate Plasticity 0.37 - 0.5 2.0 - 2.7 Clays with High Plasticity 0.4 - 0.6 1.7 - 2.5 The effects of wall friction may be used in the design of an earth retaining structure and reference should be made to publications such as FHWA Circular No. 4 (June 1999) in this regard. Extreme care should be exercised if seismic conditions may be present. 23.2.3 PASSIVE EARTH PRESSURE The passive earth pressure is the maximum earth pressure that can be mobilised by the relative movement of a structure against a soil mass. It represents failure conditions at which the shear strength of the soil is fully mobilised in resisting the lateral forces. The lateral strain required to mobilise the shear strength can be large and this needs to be taken into consideration in the evaluation of available passive resistance. The following expression may be used to calculate the passive earth pressure coefficient Kp: Ka = 1 + sinφ’ 1 -- sinφ’ Typical values of Kp for various soil types are given in TABLE 23.2.2. 372 (23.2d) Active and passive earth pressure coefficients are influenced by wall friction. The effect of wall friction on passive pressure can be large, but definite movement is required for mobilisation. The theoretical predictions regarding passive resistance as a function of wall friction are not well confirmed in practice, and the influences due to wall friction can be significantly overestimated in certain circumstances. In lateral support design, wall friction associated with passive resistance is often ignored as a stabilising force. Reference should be made to Tschebotarioff (1973) for detailed analysis procedures to take wall friction into consideration in the evaluation of passive resistance. Many design problems involve sloping excavation faces, with or without a sloping backfill. Formulae to determine the horizontal components (Kah and Kph) of the active and passive earth pressure coefficients under these conditions are given in FIGURE 23.2.1. Once the earth pressure coefficients have been determined, the magnitude of the earth pressure is obtained by multiplying the earth pressure coefficient by the effective overburden pressure. The effective overburden pressure is obtained by the determination of total vertical pressure, σv , at the depth being considered, and then deducting the pore water pressure to determine the effective vertical pressure. Kah = cos2 (φ’ + α) [ cos2 α 1 + Kph = sinφ’.sin(φ’ -- β) cosα.cos(α + β) 2 ] cos2 (φ’ -- α) [ cos2 α 1 + sinφ’.sin(φ’ + β) cosα.cos(α + β) 2 ] FIGURE 23.2.1 Horizontal Components of Active and Passive Earth Pressure Coefficients for a Sloping Excavation Face and Sloping Backfill 373 23.3 WATER PRESSURES AND SURCHARGE LOADS In the calculation of the total lateral pressures that need to be supported, it is necessary to take water pressures and surcharge loads into consideration. 23.3.1 WATER PRESSURES Net, unbalanced, water pressure must be included in the calculation of total lateral pressures. In most cases, this is equivalent to the hydrostatic pressure below the watertable. 23.3.2 SURCHARGE LOADS Surcharge loads must be included with earth and water pressures. Surcharge loads can be associated with traffic, construction, fills and buildings. These usually take the form of uniform loads, line loads, strip loads or point loads. When a uniform surcharge, q, is applied, it may be assumed that the vertical effective stress is increased by the value of the surcharge. The lateral earth pressures are then increased by the relevant earth pressure coefficient (Ko , Ka or Kp ) multiplied by q, the surcharge loading. Solutions for line loads and point loads have been obtained by modification of the theory of elasticity. These solutions are shown in FIGURE 23.3.1. Simple addition of surcharge pressures to earth and water pressures is normally sufficient where these pressures are small compared to the earth pressures (less than about 30% of earth pressures). With large surcharge loads, which may occur some distance away from the slope or excavation being supported, it is also necessary to check the required support forces using single wedge, multiple wedge, or circular arc failure surfaces, also see SECTION 23.5. 374 Line Load Point Load m < 0.4 m < 0.4 2 σh H = 0.20m 2 and Ph = 0.55QL 2 2 p 2 3 σh H = 0.28n [Q ] (0.16 +n ) [Q ] 0.16 + n L m > 0.4 2 n and P = σh H = 1.28m h 2 2 [Q ] (m + n ) L m > 0.4 0.64QL (m2 + n2) Section A - A 2 σh H = 2 2 n [Q ] (m1.77m +n ) p 2 2 σ ’h = σ h cos2 (1.1θ ) FIGURE 23.3.1 Horizontal Pressures Due to Point and Line Surcharges 375 23.4 EMBEDDED WALLS 23.4.1 CANTILEVER RETAINING WALLS A Rankine earth pressure distribution is recommended for cantilever retaining walls. A typical pressure distribution diagram is shown in FIGURE 23.4.1. It is assumed that the wall will fail by rotating about a point just above the toe of the wall, and that the active pressure is balanced by the passive pressure. Design guidelines are given in FIGURE 23.4.1. Only pressures induced by active and passive states within the soil are illustrated in FIGURE 23.4.1. Where water occurs or surcharge loads are present, the relevant water pressure and imposed surcharge pressures must be included, see SECTION 23.3. In the design procedure for the overall stability of a cantilever wall, it is recommended that appropriate safety factors be applied to the soil strength parameters, rather than to the passive forces only. A more detailed discussion on factors of safety is given in SECTION 23.6. Finite Element Programs which allow the wall to be modelled as beam elements and the soil as a continuum are the most suitable to determine bending moments and shear forces. For a simple cantilever wall, a first estimate of the maximum bending moment can also be obtained by the method given in FIGURE 23.4.1. For final designs, the values obtained from this procedure need to be checked by a more rigorous analysis. In terms of the structural design, it is currently recommended that a cantilever wall system be analysed with a factor of safety of 1.0, and that the bending moments and shears derived from this calculation are adopted as ‘working’ moments and shears. These bending moments and shears should then be multiplied by an appropriate load factor for ultimate limit state design to structural codes. Reference can be made to Eurocode 7 for limit state design methodology. 23.4.2 BRACED WALLS Semi-empirical pressure distributions such as those proposed by Peck (1969), are used to determine the magnitude and distribution of lateral support pressures for braced excavations. These distributions are given in FIGURE 23.4.2. Strut loads for specific spacings can be estimated directly from the pressure distributions given in FIGURE 23.4.2. It is important to emphasise that the distributions given are for end of construction conditions, and support forces at each excavation level should be considered in sequence. The wall elements are usually designed as continuous members supported at strut levels. 376 (a) (b) (c) (a) Deflected Shape with Rotation About Point C (b) Theoretical Pressure Distribution (c) Simplified Pressure Distribution FIGURE 23.4.1 Pressure Distribution for a Cantilever Retaining Wall Pp is the required passive resistance for stability and is obtained by taking moments about point C. The available passive resistance for the assumed depth, d, is then checked and compared with the required Pp. If necessary d is adjusted and the calculation repeated. ds is the required penetration depth and is equal to 1.2d to 1.5d. For the simplified pressure distribution shown and ignoring wall friction: d= H Kp 0.67 -- 1 (23.4a) L = span of the wall for the calculation of the maximum bending moment. L = K x H where K is obtained as follows: φ’ 20° 30° 35° 40° K 2.0 1.5 1.4 1.3 377 φ’ φ’ FIGURE 23.4.2 Pressure Distributions for Braced Excavations after Peck (1969) 23.4.3 TIED-BACK WALLS Design procedures for Tied-back walls vary from the relatively simple procedures that are used for walls with a single tie-back to complicated soil/structure interaction analyses for multi-tied walls. Either the free-earth or fixed-earth support methods can be used for the design of single tied-back walls. A Rankine pressure distribution is generally used in the analysis. Typical pressure distributions and design details are given in FIGURES 23.4.5 and 23.4.6. The design of multi-tied walls is a complex soil/structure interaction problem in which the earth pressure distribution which governs the design depends on the wall stiffness, the method of construction, the tie-back spacing and pre-stress load. Sophisticated programs such as PLAXIS or FREW can be used to carry out rigorous numerical analyses of multi-tied walls, as shown in FIGURE 23.4.3. The procedures given below are recommended in the absence of sophisticated software. These procedures should be carried out for all stages as the excavation and tie-back installation proceeds: . Evaluate the magnitude of support force required using limit equilibrium techniques involving the analysis of single, or multiple wedges, or circular arc failures. A typical single wedge analysis is illustrated in FIGURE 23.4.4. The single wedge analysis can be used with confidence for routine designs with uniform geometry and simple surcharge loads. Software is available for the analyses of complex geometric problems. In carrying out this analysis suitable safety factors should be applied to the soil shear strength parameters, also see SECTION 23.6. 378 . . . Check the magnitude of the support forces using a suitable earth pressure distribution. The earth pressure coefficient used in this calculation will vary between Ka and Ko depending on anticipated and allowable wall movements and surcharge loading. These coefficients should be determined using an appropriate factor of safety applied to the soil shear strength parameters, see SECTION 23.6. In general terms a triangular distribution of earth pressure will be the most appropriate. Once the procedures given above have been used to determine the magnitude of support force required then the distribution of the anchor forces needs to be decided upon. In less complex cases this can often be decided upon using judgement decisions based on geometric considerations and taking due cognizance of the soil/rock profile and surcharge loads. A simple evaluation of anchor forces can also be obtained by considering the contributory area of the earth pressure diagram applicable to each anchor. More rigorous analysis, such as that given by Littlejohn, Jack and Sliwinski (1971) can also be used under appropriate conditions. Consideration also needs to be given to the overall anchor arrangement in terms of other factors, such as minimum free lengths, and minimum depth below ground surface. For guidelines in this regard reference should be made to the SAICE Code of Practice for Lateral Support in Surface Excavations (1989). The support system as a whole should be checked for overall stability. This will comprise a series of analyses in which the system containing the wall and anchors is checked for stability along selected circular and non-circular potential failure surfaces. The wall elements for both single and multi-tied walls should be designed to resist the induced bending moments and shear forces from the overall anchor and earth pressure system during all stages of excavation and construction. Rowe (1952) has shown that the flexibility of the wall is an important consideration for single tied-back walls in which the support forces mobilised are a combination of passive resistance from the earth below excavation level, and the force provided by the brace or tie-back system. Depending on the flexibility of the wall there is a reduction in bending moment due to redistribution of soil pressures associated with arching of the soil being supported. Typical bending moment reduction factors are given in TABLE 23.4.1. For multi-tied walls the wall elements can be designed as continuous members supported at tie-back positions with a load distribution in accordance with the assumed earth pressure distribution. The most suitable procedure for multi-tied walls is to use a subgrade reaction model, where the wall is modelled as beam elements, and the soil as a system of springs. In the structural design of the wall elements, the overall lateral support system should be analysed with a factor of safety of 1.0 applied to the lateral pressures, and the bending moments and shear forces derived should be adopted as ‘working’ moments and shears. These moments and shears should then be multiplied by an appropriate load factor for ultimate limit state design to structural codes. 379 380 FIGURE 23.4.3 Deformations from a Plaxis Finite Element Analysis of a Tie-Back Wall TABLE 23.4.1 Bending Moments Reduction Factors for Single Tied-Back Walls after Reynolds (1981) Bending Moment Reduction Factor Ratio of Wall Thickness to Span φ = 20o φ = 30o φ = 35o 0.02 0.70 0.56 0.48 0.10 0.80 0.69 0.62 0.20 0.86 0.78 0.72 0.30 0.90 0.83 0.78 0.40 0.91 0.87 0.83 FIGURE 23.4.4 Forces Acting in Single Wedge Mechanism of Failure after BS 8081 (1989) Known Unknown P = surcharge W = weight of sliding wedge = 1/2 γ H2 cot B γ = density of wedge H = depth of excavation S = shear resistance of retained material along plane of rupture. S = (cH/sinB) x N tan φ’ A = angle of inclination of the anchor φ’ = effective angle of friction c’ = effective cohesion B = angle of inclination of potential plane of rupture (B should be varied and plotted against values of T) T = anchor force N = normal force on the wedge N = (P + γ H/2)HcosB.cotB + T sin (A + B) P + γ H H cos β (F -- cot β tan φ ’) -- c’ H ( ( sin β) 2 ) T= sin (A + β ) tan φ ’ + F cos (A +β ) Where: F = the Factor of Safety required 381 (23.4a) (b) (a) (c) (a) Deflected Shape (b) Assumed Pressure Distribution (c) Bending Moment Distribution FIGURE 23.4.5 Pressure Distribution for Free-Earth Support Free-earth support assumes that the passive resistance in front of the wall is sufficient to resist forward movement at the toe, but not sufficient to prevent rotation. Free-earth support is recommended for loose cohesionless soils, silts and clays. Where: T Pp = = the required tie-back force the theoretical passive resistance required for stability of the wall and can be determined by taking moments about E. The available passive resistance for the assumed depth, d, is then checked and compared with the required Pp . If necessary d is adjusted and the calculation repeated. For equilibrium Σ horizontal forces = 0. This allows T to be determined for the pressure distribution shown: T = PA -- Pp (23.4b) The maximum bending moment is calculated using the assumed pressure distribution and assuming that the wall is simply supported at E and D. 382 (a) (b) (c) (a) Deflected Shape (b) Assumed Pressure Distribution (c) Bending Moment Distribution FIGURE 23.4.6 Pressure Distribution for Fixed-Earth Support Fixed-earth support assumes that the passive pressure in front of the wall is sufficient to prevent both forward movement and rotation at the toe. Fixed-earth support is recommended for dense sands and gravels. For analysis purposes the wall is considered to be two equivalent beams EF and FC connected by a pin joint at F. The following relationship between x/H and φ’ is used to determine the position of F. φ’ 20° 25° 30° 35° 40° x/H 0.25 0.15 0.08 0.033 – 0.01 The tie-back force, T, is calculated by equating moments about E, to calculate the reaction at F, and then determining T, for horizontal equilibrium of beam EF. The required depth, d, is calculated by equating moments about C, and solving for (d -- x). An increase of 20% to 50% is made to the calculated value of d, to allow for the length CD. The maximum bending moment is determined by considering the wall as a beam simply supported at E and F. 383 23.5 REINFORCED SOILS The design procedures for reinforced soils are no different to any other lateral support system. To guarantee a sufficient margin of safety, equilibrium should be checked for all possible failure mechanisms using safe assumptions for the properties of the soil and reinforcement and taking due consideration of water pressures and surcharge loadings. 23.5.1 GEONAILS A GeoNail system should be checked for both internal and overall stability in accordance with the procedures given below. These procedures should be applied to all stages of construction. Internal Stability This is the most important design check for a GeoNail reinforced soil mass. Limit equilibrium procedures are generally adopted in which the equilibrium of all possible failure surfaces through the GeoNail reinforced zone are checked. It is essential to demonstrate that equilibrium can be maintained on all possible failure surfaces, and it is not sufficient to focus on a single ‘critical surface’. It is questionable whether a stability analysis using a general method of slices is appropriate to check internal stability of a GeoNail reinforced soil mass. This is due to the unknown influence of the GeoNails on the forces between individual slices. Limit equilibrium analyses using rigid body mechanisms would appear to be the most suitable method for checking equilibrium, since there is no need for assumptions on internal slices. The simplest method in this regard is a single wedge analysis. This type of analysis is illustrated in FIGURE 23.5.1, and can be used with confidence for designs with uniform geometry and soil profiles and simple surcharge loads. A two-part wedge analysis may be appropriate with complex geometric conditions and high surcharge loads, particularly if these loads occur some distance beyond the crest of the excavation. For more specific details with regard to design procedures, reference should be made to Gassler and Gudehus (1981), Shen et al (1982), Gassler (1988), and Long et al (1990). It is necessary to take into consideration that the internal stability of a GeoNail structure depends on the geometry, surcharge loadings, shear resistance of the soil, and the tensile strength and pull-out resistance of the GeoNails. Taking these factors into consideration, it is apparent that the concept of a single factor of safety, as applied to conventional slope stability analysis, is not satisfactory for a GeoNail reinforced system. The most appropriate design procedure is to carry out the analysis using the appropriate geometry, water pressures and surcharge loads, and to select safe values for the soil shear strength, and tensile and pull-out resistance of the GeoNails. These safe values are obtained by applying the appropriate factors of safety to the relevant parameters, see SECTION 23.6. The design process should then demonstrate satisfactory equilibrium, by determining the required forces to maintain equilibrium, and then designing a GeoNail layout so that the available forces provided by the GeoNail system exceeds the required forces on all the potential failure mechanisms. 384 Where: Q W R = = = θ = φ = FNreq = FNact = surcharge on the wedge weight of the sliding wedge reaction force on the failure plane angle of inclination of the failure plane angle of shearing resistance of soil required force for stability force provided by the GeoNails Tm Sh li mean force per metre length of GeoNail horizontal spacing of the GeoNail length of nails beyond the failure plane = = = FIGURE 23.5.1 Single Wedge Stability Analysis of a GeoNail Structure, after SAICE Code of Practice Lateral Support in Surface Excavations (1989) The value of pull-out resistance of the GeoNails is probably the most significant factor in determining available forces. The assumption that ultimate bond (τult) is related to the effective cohesion, and the normal effective stress multiplied by the effective angle of friction (τult = c’ + kσ v’ tan φ ’), appears to give excessively conservative values in most instances. Work carried out by Heymann et al (1992) has shown that for insensitive cohesive soils and soft rocks the ultimate bond for GeoNails can be obtained using the same procedures as those used to predict the ultimate skin friction for piles. For recommendations in this regard reference should be made to SECTION 21.0: DESIGN AIDS: PILING. Heymann et al (1992) has also shown that for sandy residual soils of low to moderate plasticity, such as residual granites and sandstones, a lower bound value for the ultimate bond τult can be estimated from: τult in kPa = 4 φ’ for dilatant material τult in kPa = 3 φ’ for silty materials τult in kPa = 2 φ’ for clayey material NOTE: For permanent works, creep tests should be undertaken to confirm the ultimate capacity, with measurement of deflections at different load magnitudes over time. Creep deflections are not covered by the local code of practice, and reference should be made to prEN 14490:2002, and CIRIA C637:2005 (section 11.3), for further details. 385 For many soil types there does however not appear to be a satisfactory method of accurately predicting ultimate bond values. For most projects the most satisfactory procedure is to make the best possible design assumptions, carry out suitable in-situ pull-out tests as soon as possible after the commencement of the project, and then revise the design. Overall Stability This comprises a check on the stability of the overall GeoNail structure. This is firstly carried out using the analogy that the GeoNail structure acts as a homogenous and resistant unit to support the soil behind in a manner similar to a gravity retaining wall. A second design check should then be carried out comprising a series of analyses in which the overall GeoNail structure is checked for stability along selected external circular and non-circular failure surfaces. These concepts are illustrated in FIGURE 23.5.2. FIGURE 23.5.2 Design Check for Overall Stability of a Geonail Structure 386 23.5.2 RETICULATED MICROPILES Design procedures for reticulated micropiles are described by Lizzi (1983) and (1989). In these publications the author stresses that the design of reticulated micropiles is not amenable to rigorous theoretical analysis and that the design approach is empirical and based on engineering judgement obtained from back-analysis of real cases. To quote from Lizzi (1989): ‘In spite of a very large quantity of studies and calculations, some of them with the help of sophisticated computers, we cannot state that the matter is theoretically under control’. The following general design guidelines, which are based mainly on Lizzi (1983) and (1989), should be applied to a reticulated micropile support system: . . . The reticulated micropile structure should be considered to act as a gravity retaining structure. The lateral thrust on the structure due to earth pressure, water pressure and surcharge loads can be calculated from the recommendations given previously in SECTIONS 23.1, 23.2 and 23.3. The earth pressure coefficient used in the evaluation of lateral thrust will fall between Ka and Ko but probably closer to Ko. In the determination of the earth pressure coefficient a suitable factor of safety should be applied to the soil shear strength parameters, see SECTION 23.6; The next step in the design process is to choose a suitable geometric layout. The layout usually consists of closely spaced vertical and raked piles, see SECTION 17.10. Lizzi (1983) recommends that the sliding stability of the reticulated micropile structure should be resisted by the soil only through friction and cohesion, and that shear along the piles be ignored. This would seem to be a conservative approach, but can be used as a basis in deciding on the overall geometry of the pile layout system at any critical sliding surface; An evaluation can be made of the compressive, tensile and shear forces within individual piles. Lizzi (1983) recommends a procedure, which he considers to be analogous to reinforced concrete, whereby the compressive loads due to the lateral earth pressures are carried jointly by the piles and the surrounding soil in accordance with an amplification factor. This amplification factor is defined as the ratio of the elastic modulus of the pile to that of the soil. The tensile loads induced by the lateral earth pressures are carried only by the piles. These tensile and compressive loads are to be carried only by the length of pile below any critical failure surface. Reference should be made to SECTION 21.0: DESIGN AIDS: PILING for procedures to evaluate compressive and tensile capacity of the piles. Lizzi (1989) indicates that the pile loads calculated are approximate, but that this is not particularly significant, since the loads are usually small due to the high density of piles. If a pile does become overstressed there will also be load transfer to the remaining piles. A similar design approach to evaluate individual pile loads is described by Dash and Jovino (1980), except that the concept of an amplification factor is not used. 387 . It is necessary to ensure that there is sufficient stability against sliding on any critical surface. As indicated previously, Lizzi (1983) recommends that sliding be resisted by the soil only through friction and cohesion. This approach is probably conservative. Dash and Jovino (1980) take the opposite viewpoint and assume that sliding is resisted entirely by shear on the piles. It would seem to be sensible to include shear resistance from the piles in analysing sliding stability. This is particularly the case if one considers that the piles are very much stiffer than the surrounding soil and will tend to attract load even if only small movements takes place along potential sliding surfaces. 23.5.3 SOIL DOWELLING With soil dowelling, heavily reinforced large diameter piles are installed into marginally stable slopes, to provide stability through shear and bending resistance. The forces that need to be resisted by a soil dowelling system are usually of large magnitude. The failure surface in this instance is usually well defined and the magnitude of the forces to be resisted can be evaluated using conventional slope stability analysis. The recommendations given in SECTIONS 23.1 and 23.6 should be used in carrying out the slope stability analysis. Once the forces to be supported have been determined, the design of the soil dowels should be carried using conventional procedures for laterally loaded piled foundations. Reference should be made to SECTION 21.0: DESIGN AIDS: PILING in this regard. 388 23.6 FACTORS OF SAFETY In deciding on a minimum Factor of Safety for the design of a lateral support system the design engineer needs to take the following into consideration: . .. .. . The reliability of the measured, or assumed values, of the relevant soil or rock parameters involved in the analysis; The reliability and accuracy of the mathematical model used in the analysis; The magnitude of surcharge loads and the confidence level in the prediction of these loads; Previous experience in similar geotechnical conditions; Previous experience with the lateral support system to be utilised; The consequences of failure. It is convenient to consider failure in this context in terms of limit state theory. The extreme event in this case would be the ultimate limit state where complete collapse of the lateral support system occurs. For most lateral support systems the serviceability limit state will however be the most important criteria. With the serviceability limit state failure can be considered to occur when movement of the ground being supported exceeds an allowable value. This allowable value will obviously be much higher for the support of virgin ground, or a street face, compared to the support of a heavily loaded, settlement sensitive structure. The above factors have been taken into consideration in compiling a set of recommended minimum Factors of Safety for various lateral support applications. These recommended minimum values are given in TABLE 23.6.1. It is important to emphasise that these minimum values should be used as guideline values by the design engineer and each project should be treated on its merits. TABLE 23.6.1. Guidelines for Minimum Factors of Safety Recommended Minimum Factor of Safety Type of Support System Temporary Support Permanent Support Low Surcharge High Surcharge Low Surcharge High Surcharge and/or and/or and/or and/or movement movement movement movement not critical critical not critical critical EMBEDDED WALLS Soil and rock effective shear strength parameters Tie-back capacity REINFORCED SOILS Soil and rock effective shear strength parameters 1.25 1.5 1.5 1.5 - 2.0 1.25 1.25 1.5 1.5 - 2.0 1.25 Only applicable under certain circumstances 1.5 Generally not applicable GeoNail bond 2.0 Tensile or compressive capacity of Micropiles 2.0 2.0 - 2.5 2.0 - 2.5 389 23.7 MOVEMENTS ASSOCIATED WITH EXCAVATIONS An essential point in considering movements associated with excavation is that no excavation, however well supported, can be made without causing some ground movement. The ground movements are generally due to lateral yield of the soil/rock towards the excavation, with an associated component of vertical movement. The magnitude of movement that occurs is related to the height of the excavation to be retained, the nature of the soil/rock strata being retained, and the magnitude of surcharge loads. It is possible to exercise some control over the magnitude of the movement by the selection of the correct lateral support system and construction techniques. For example, if a support system is required for an excavation adjacent to a settlement sensitive structure with relatively high surcharge loads, then the designer would not choose a passive support system such as GeoNails, but rather an active system with post-stressed anchors. The potential for movement can be even further reduced by using closely spaced large diameter concrete soldier piles as a wall element, in conjunction with the post-stressed anchors, and by adopting carefully phased excavation procedures. Prediction of movements cannot be made with the classical theories used in lateral support design. Experience in recent years indicates that under certain conditions it may be possible to arrive at reasonably close estimates of movements using Finite Element techniques. At present the conventional approach in the prediction of movements is to use empirical methods based on observational data of previous excavations. The chart presented as FIGURE 23.7.1 has been used in certain applications. This chart was derived, by Peck (1969), from data obtained from a number of instrumented excavations, in a variety of soil conditions. Movements much smaller than those that would be predicted from FIGURE 23.7.1 have been observed in the monitoring of numerous basement excavations in southern Africa. This may be due to the fact that the majority of the basement excavations that have been monitored are in residual soils above the water-table. A further reason may be due to the advances in lateral support design and construction procedures that have occurred subsequent to the work carried out by Peck (1969). Empirical values, based on local experience, for the prediction of horizontal movement at the crest of excavations, are given in TABLE 23.7.1. Local experience has also shown that vertical settlement at the crest of an excavation is usually smaller than the horizontal movement. It is important to emphasise that the values given in TABLE 23.7.1 should be used by the designer as a guide to obtain an indication of the magnitude of movement that can be anticipated. In instances where it is necessary to exercise some control over the magnitude of movement, the designer should consult the nearest Franki office to decide on the most suitable lateral support system and construction techniques. 390 ZONE I: Sand and soft to hard clay, average workmanship ZONE II: Very soft to soft clay ZONE III: Very soft to soft clay to a significant depth below bottom of excavation FIGURE 23.7.1 Settlement Adjacent to Open Cuts in Various Soil Conditions after Peck (1969) TABLE 23.7.1 Empirical Values to Predict Horizontal Movements at the Crest of an Excavation Type of Support System Horizontal Movement as a Percentage of Excavation Height EMBEDDED WALLS Cantilever retaining walls Walls with prop supports Tied-back walls 0.5% 0.2 to 0.5% 0.05 to 0.15% REINFORCED SOILS GeoNail Systems 0.1 to 0.3% Applicable Norms As the subject of Lateral Support Design and Slope Stabilisation is broad, not all aspects can be covered in one book. The subject has been adequately covered by the South African Code of Practice outlined below: CIRIA C 515: Groundwater Control – Design and Practice (2000) CIRIA PR 77: Prop Loads in Large Braced Excavations (2000) CIRIA SP 201: Response of Building to Excavation – Induced Ground Movements (2002) CIRIA C 580: Embedded Retaining Wall – Guidance for Economic Design (2003) SAICE: Geotechnical Division: Code of Practice for Lateral Support in Surface Excavation (1989) 391 24.0 QUALITY ASSURANCE AND SAFETY To ensure that the Franki Africa slogan ‘Quality is our Foundation’ is more than a mere statement, a rigorous quality assurance programme has been developed by the company for use in the design and installation of its products, as well as the services that it offers. This programme is flexible and can also accommodate any additional quality assurance requirements that a client might impose on a project. The programme governs all the company's activities by means of standardised procedures in accordance with three levels of control. The three levels are referred to as Level 1, Level 2 and Level 3. The first level is a basic level quality assurance, and is applied to all the Company's activities, other than where one of the other levels is applied. Level 2 is used when there are special circumstances, such as difficult ground conditions, or a particularly difficult construction method, and where a higher level of control is desirable. If there are Quality Assurance requirements in the contract document, then Level 3, the highest level, is applied. The following is a more detailed account of these three levels. 24.1 LEVEL 1 This is the basic level and covers the company's everyday operations which are anticipated to have no undue complications. It is the minimum level of quality assurance that the company applies to all its operations. It covers the following: (a) Contract Review Prior to the submission of a tender for a contract the specified conditions are reviewed to ensure they can be met. This review is recorded formally. (b) Document Control At this level the documentation of the foundation is formally controlled, ensuring that all staff are working with the latest drawings and specifications. All official documentation such as correspondence with the Engineer/ Architect/Employer any transmittal of contract drawings, material and product test results and the like are formally recorded and sent under cover of a receipted transmittal note. This ensures that work is executed to the latest instructions and recorded properly for traceability. (c) Concrete Testing In addition all concrete and grouts are tested regularly to ensure that, the strength, slump and any other requirement is met. All testing of concrete and grout will be executed as per specifications and recorded and transmitted formally. Results will be recorded as per instructions form the engineer. (d) Training All personnel will have received the requisite training in the correct installation of the company's products according to the company's product manuals. This training is recorded on the personal files of the staff concerned. 392 24.2 LEVEL 2 This level is used when special systems are to be installed, or where complications might be expected in the execution, due to uncertain ground conditions, or where any other aspect of the contract suggests a higher level of control is desirable. It introduces the concept of executing the work in accordance with written procedures. This means that all operations are reduced to written standard procedures with the recording of different steps in the process to ensure compliance with the specifications and /or design criteria. These procedures are all set out on a standard format. 24.3 LEVEL 3 This is the highest level of quality assurance and is applied to contracts where such a level is a requirement of the contract documents. At present this is the case with large complicated contracts such as the construction of power stations, refineries and metallurgical plants. On these contracts everything related to a construction activity is formally documented. A particular feature would be the traceability of materials back to their source and obliges suppliers and sub-contractors to fulfill all the requirements of the specifications and the quality assurance programme. It is the responsibility of the main contractor to audit suppliers and sub-contractors in this regard. The following are some of the quality assurance activities covered by this level: (a) Purchasing The purchase of built-in materials are controlled to ensure compliance with the specifications. (b) Process Control This requires that all construction and manufacturing operations be covered by detailed written procedures. The process control requirements are also applied at Level 2. (c) Inspection, Measuring and Test Equipment This ensures that such equipment is identified, registered, inspected, calibrated, and maintained at regular intervals. (d) Non-conformance This deals with the identification, documentation, evaluation and reworking of non-conforming materials and system elements. (e) Internal Quality Control It is also a requirement of the quality assurance programme that all operations are audited on a regular basis to ensure compliance by all concerned. 393 It should be noted in particular that the quality assurance system, at whatever level, requires that any non-conformity, or anomaly, is to be fully recorded. Such recording is done on standardised forms for transmittal to the relevant parties for analysis and disposition. This disposition needs to be approved by the designer. By using its Quality Assurance programme Franki Africa is able to ensure that the quality of its products and services is constantly monitored. In addition, the standardised recording of installation procedures allows the company to monitor the efficacy of the work procedures, and to modify them accordingly. When errors, or incorrect procedures, are identified, it is then able to ensure that corrective action is taken to avoid further non-conformance. Corrective action may involve the adaptation of the procedure to eliminate the nonconforming condition. It may also involve changing to another foundation system if it is found that the non-conformance cannot be eliminated through adaptation. Retraining of staff may also form part of the corrective action. Because most of the work is carried out below ground level, Franki Africa considers the use of its Quality Assurance programme essential if it is to ensure the quality of the products and services described in this book. By its use the company is able to guarantee its products with confidence. 394 24.4 ISO 9001 CERTIFICATION Franki achieved an SABS ISO 9001 certification in 1996, the first geotechnical engineering contractor to be awarded this certification in the South African construction industry. To ensure that the Franki slogan ‘Quality is our Foundation’ is more than a mere statement, a rigorous quality assurance programme has been developed by the company for use in the design and installation of its products, as well as the services that it offers. This programme can also accommodate any additional quality assurance requirements that a client might need on a project. I S O 9 0 0 1 395 25.0 REFERENCE INFORMATION 25.1 NORMAL PLANT CLEARANCE REQUIREMENTS NOTES 1. Clearances given are the absolute minimum for vertical piles. For raking piles additional clearance may be required and this should be discussed with your local Franki office. 2. Headroom dimensions refer to standard systems and may be reduced under certain circumstances and using special or modified equipment. 3. Working in close proximity to other structures will slow down the production rate and thus increase the cost. Avoid working to the minimum clearances if at all possible. 4. Special attention must be given in the case of the working area being in close proximity to an overhead powerline. The requirements of the appropriate authority must be obtained and adhered to. 396 397 A 300 250 380 350 200 300 250 1200 900 1050 800 800 800 550 1750 1500 1125 800 1200 1180 625 1125 800 2150 2000 2350 1950 1850 2400 2050 2250 2150 B 6190 5320 5875 6550 2650 6190 5320 8510 6210 9420 7850 6800 7800 7000 10525 11650 C 7022 6027 5735 8742 1830 7022 6027 20905 18000 23135 19600 18046 19200 18500 24000 24000 D CLEARANCE DIMENSION (mm) NOTE Dimension A includes a clearance of 100 mm Dimension B includes a clearance of 150 mm Dimension C includes a clearance of 200 mm Dimension D includes normal height clearance Dimension E is the maximum height clearance requirement for rig with mast extensions CASSAGRANDE C6 CASSAGRANDE C4 I.R. ECM 350 TEI ROCK TD-75 CASSAGRANDE C6 CASSAGRANDE C4 MICROPILES LATERAL SUPPORT ANCHORS / NAILS BAUER MBG 24 / 48 BAUER BG 15H SOILMEC SR-80 SOILMEC SR-50 SOILMEC SR-30 BORED PILES CASSAGRANDE C8 FRANKI SA 81 CRAWLER FRANKI SA 83 CRAWLER LIEBHERR 845 CRAWLER AJAX C60 / C75 CRAWLER DRIVEN PILES JET GROUTING RIG TYPE PRODUCT TYPE E 9922 10819 5735 14742 1830 9922 10819 23405 23100 28702 25470 21900 19200 18500 52000 27400 25.1 NORMAL PLANT CLEARANCE REQUIREMENTS 3450 2250 FRANKI SA 81 CRAWLER PILING RIG 398 4500 (TRACK EXTENDED) 3750 2250 TRACK CENTRES 8750 18450 4750 8100 2600 APPROXIMATE MASS: MACHINE ONLY: 62 tonne TUBE and HAMMER: 10 tonne 72 tonne 4950 3320 25.2 RIG DIMENSIONS 25.2.1 RIG DIMENSIONS: FRANKI RIGS 25.2.2 RIG DIMENSIONS: CRANES m 5.59 4.17 4.15 1.16 4.41 A = Length of Crawlers C = Overall Width E = Tail Radius over Counterweight F = Clearance under Counterweight H = Height of ‘A’ Frame H F 45 o m et re s 48 .8 51 .8 42.7 45 .7 39.6 42 .7 36.6 o 40 39 .6 33.5 33 .5 36 .6 30.5 30 .5 27.4 27 .4 24.4 o 30 o 25 21 .3 24 .4 21.3 o 35 18 .3 18.3 15.2 o 20 15 .2 HEIGHT ABOVE GROUND IN METRES 50 o 55 o 60 o 45.7 70 o 75 o A 65 o C E o 15 12.2 9.1 RADIUS DIAGRAM 4.6 6.1 9.1 12.2 15.2 18.3 21.3 24.4 27.4 30.5 33.5 36.6 39.6 DISTANCE FROM CL OF ROTATION IN METRES NCK AJAX C75 / C60 CRAWLER CRANE 399 25.2.2 RIG DIMENSIONS: CRANES 1145 57 3380 1370 1870 400 11860 2850 1200 5000 5960 7380 2500 1545 R4 4500 0 700 3000 3800 345 LIEBHERR HS 845HD CRAWLER CRANE 25.2.2 RIG DIMENSIONS: CRANES m 60 56 52 48 44 40 36 32 28 24 20 16 12 8 4 0 48 44 40 36 32 28 24 20 16 12 8 4 0m Heavy Main Boom Configuration Length (metres) Configuration for Boom Lengths (11 – 50 metres) Boom Foot 5.5 1 Boom Insert 3.0 Boom Insert 6.0 Boom Insert 12.0 Boom Head 5.5 1 Boom Length 11 (metres) Amount of Boom Extensions 1 1 1 14 1 1 1 1 1 1 2 1 1 2 1 17 1 20 1 23 1 26 1 1 1 1 29 1 1 1 1 1 32 1 2 1 1 35 1 1 2 1 1 38 1 1 2 1 41 LIEBHERR HS 845HD CRAWLER CRANE 401 1 1 1 2 1 44 1 2 2 1 47 1 1 2 2 1 50 25.2.3 RIG DIMENSIONS: PILING (LARGE DIAMETER BORED) 20306 Kelly BK 25/394/3/21 (A = 9.71 m) 16921 7206 5369 1000 3453 1230 Kellyhub / kelly stroke 5600 mm 9053 0 226 4685 5600 3860 3650 BAUER MBG 24 / 48 HYDRAULIC DRILLING RIG 402 25.2.3 RIG DIMENSIONS: PILING (CFA) 6000 22900 13720 13685 13400 15905 13400 15905 900 900 320 0 285 0 4717 3010 4717 3200 3010 BAUER BG 15H HYDRAULIC DRILLING RIG 403 3200 9405 25.2.3 RIG DIMENSIONS: PILING (CFA) 23135 22765 17580 1030 346 1213 1356 Rotary 481 STROKE 15790 19298 28702 Rotary 5681 4800 900 4620 4700 MIN 4200 / MAX 4800 SOILMEC SR-80 HYDRAULIC DRILLING RIG 404 2105 19558 PULL-DOWN IN HIGH POSITION 12700 11555 1650 25.2.3 RIG DIMENSIONS: PILING (AUGER) 1070 1188 3310 3749 2135 4291 563 250 PULL-DOWN IN LOW POSITION 750 539 5203 1455 3902 700 700 3900 SOILMEC SR-50 HYDRAULIC DRILLING RIG 405 8800 10490 299 1000 1198 3268 4511 1531 2205 1596 3800 939 1302 3500 (MIN 3000) 874 939 750 3318 1693 395 2479 3500 PULL-DOWN STROKE 894 3206 2299 H 4709 18046 25.2.3 RIG DIMENSIONS: PILING (AUGER) 4646 12424 SOILMEC SR-30 HYDRAULIC DRILLING RIG 406 4101 312 25.2.4 RIG DIMENSIONS: LATERAL SUPPORT / JET GROUTING 1400 90O 8742 7600 5600 1322 4202 90O 2110 830 500 2555 /3355 2345 6700 3270 3040 2360 3235 500 500 2350 8742 CASAGRANDE C8 HYDRAULIC DRILLING RIG 407 4000 2600 1298 710 580 7022 6000 1605 9922 6520 25.2.4 RIG DIMENSIONS: LATERAL SUPPORT / MICROPILES 932 1238 7322 500 370 o 18 2767 2518 500 1898 3364 2280 400 2990 7712 2250 CASAGRANDE C6 HYDRAULIC DRILLING RIG 408 2600 10819 25.2.4 RIG DIMENSIONS: LATERAL SUPPORT / MICROPILES 2234 5478 373 5580 1514 2370 3933 1000 1543 2234 1496 1600 470 6027 2800 277 CASAGRANDE C4 HYDRAULIC DRILLING RIG 409 TEI-ROCK TD75 DRILLING RIG 410 1890 2750 DRIVE TRAVEL 1180 TRAM CONTROL 750 1250 38KW ELECTRIC MOTOR PIVOTING DRILL CONTROLS 2032 TRACKS RETRACTED TRACKS EXTENDED FEED EXTENDED 520 TRAMMING CONFIGURATION 2450 1829 10O 90O PIVOTING STABILIZERS 90O O 30 30 O 25.2.5 RIG DIMENSIONS: MICROPILES REFERENCES Baguelin, F., Jezequel, J.F. and Shields, D.H. 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(1977), The Use of a Self-boring Pressuremeter to Determine the Undrained Properties of Clays, Ground Engineering. Winterkorn, H.F. and Fang, H. (Ed) (1975), Foundation Engineering Handbook, Van Nostrand Reinhold, New York, U.S.A. 416 INDEX A Consistency, 7, 8, 21, 31, 32, 49, 51, 63 Consolidation, 167 Consolidation Tests, 28 Contiguous Pile Walls, 204, 208, 209, 224 - 226 Continuous Flight Auger (CFA) Piles, 65, 68, 69, 114 - 118 Core Barrel, 12, 13 Core Orientation, 4, 13 Core Recovery, 42 Corrosion Protection: Ground Anchors, 234 - 237 Steel Sheet-piles, 215 Cut-off Walls, 201, 266, 267 Cutter Soil Mixing, 201 Accelerated Consolidation, 168, 192, 367, 368 Active Earth Pressure, 372, 373, 376 Allowable Bearing Pressure, 165, 357, 359, 360 Anchored Walls, 204, 216 - 219, 221, 227 Anchors: Corrosion Protection, 234, 235, 237 Deadman, 234 Soft Ground, 240 Anchor Piles, 158 Angle of Friction φ, 7, 8, 18, 32, 49, 51, 52 Artesian Conditions, 78, 129 At Rest Pressure, 371 Atterberg Limits, 7, 8, 29, 46, 51 Auger Piles, 2, 67, 68, 97 -104, 258, 271 Auger Trial Holes, 2, 4, 6, 7, 9 - 11, 19, 262 D B Deadman Anchors, 234, 236 Deep Soil Mixing, 199, 200 Deformation Modulus, 59 - 63 Density Tests, 26, 29 Design: Dynamic Compaction, 358 - 362 GeoNails, 384 - 386 Lateral Support, 369 - 391 Piles, 288 - 318 Pile-caps, 339 - 349 Pile Groups, 322 - 324 Pile-shafts, 334 - 338, 350 - 355 Soil Improvement, 356 - 362 Soil Replacement, 363 - 365 Diaphragm Walls, 204, 208, 209, 227 - 231 Dolomites, 261, 262 Dolphins, 287 Downdrag, 215 Drainage Blanket, 194 Drilled Micropile, 67, 138 - 142 Driven Cast-in-situ Piles, 67 - 69, 70 - 85 Driven Pre-formed Piles, 67 - 69, 86 - 96 Dry Density, 51, 52 Dry Docks, 272, 277, 278 Dynamic Compaction, 175 - 181, 356 - 358 Dynamic Cone Penetration Test (DPSH), 4, 7, 8, 21 - 25 Dynamic Probe Light (DPL), 4, 7, 25 Dynamic Replacement, 187 - 190, 363, 364 Band Drains, 193, 194, 367 Barettes, 106, 110, 113 Bentonite, 106, 107, 109, 111, 228, 229 Bored Cast-in-situ Piles, 67, 97 - 137 Boreholes, 6, 12, 13, 16, 18, 26, 43 Boulders, 33, 65, 69, 88, 97, 103, 117, 123, 129, 134 Braced Walls, 376 - 383 Bulk Density, 7, 29, 370 C California Bearing Ratio (CBR), 25 Cantilever Walls, 204, 205, 218, 376 Cavities, 261 Clay: Dispersive, 263 Expansive, 251 - 255 Cofferdams, 210, 280, 281 Cohesion, 7, 8, 369, 381 Cohesionless Soils, 49, 50 Cohesive Soils, 49, 50, 290 - 293, 303, 385 Collapse Potential Index (CPI), 28 Collapse Tests, 7, 8, 28 Collapsible Soils, 177, 256 - 258 Compaction, 168, 170 - 183, 356 - 362 Compaction Grouting, 182, 183 Concrete Soldier Piles, 204, 208, 209, 221 - 223 Cone Penetration Test (CPT), 4, 7, 8, 24, 25, 47, 48, 51 Cone Penetrometer, 24 Cone Penetrometer, Electric, 24 417 E I Earth Pressures: 371 - 373 Active, 372 At Rest, 371 Passive, 372, 373 End-bearing Capacity, 291, 294, 295, 298, 300, 301, 303 Environmental Investigations, 264, 265 Expansive Soils, 48, 250 - 255 Extensometers, 15 Inclinometers, 15 Index Property Tests, 7, 8, 29, 49, 51 In-situ Density Tests, 26 Integrity Testing of Piles, 110, 131, 161 - 164 ICP Pile Design, 300 - 303 J Jet Grouting, 168, 169, 195 -198, 249 Jetting, 75, 83, 89, 95 Jetties, 268, 270, 284, 285, 287 Joints, in Rock, 40 F Factor of Safety, 317, 389 Field Investigation, 4, 9 - 27 Fixed-earth Support, 383 Forum Bored Piles, 65 - 67, 124 - 128 Fox Correction Factors, 332 Franki Piles, 65 - 67, 70 - 79 Franki Precast Composite Pile, 76 Free-earth Support, 382 Friction Angle, 7, 8, 18 Friction Capacity, 292, 293, 296, 297, 299, 300 Friction Ratio, 24 Full Displacement Screwpile, 65 - 67, 119 - 123 K Kelly, 98 Karst, 134, 137, 138 L Laboratory Testing, 2, 4, 28, 29, 56 Lateral Load Test, 159 Lateral Support: Classification, 204 - 209 Design Aids, 369 - 394 Selection, 202, 203 Technical Details, 210 - 249 Liquefaction, 24, 263 Liquid Limit, 29 Load Testing of Piles: 156 - 160 Compression, 156 - 158 Tension, 159 Lateral, 159 Lugeon Test, 26 G GeoNails, 241 - 245 Geotechnical Investigation, 2, 4 - 27 Geophysical Techniques, 27 Grading Analysis, 7, 29 Ground Anchors, 234 - 240 Groundwater Monitoring, 264, 265 Groundwater Table, 7, 8, 35, 36 Group Index, 44 Grouting: Anchor, 238 Compaction, 182, 183 Jet, 155, 195 - 198, 249 Gunite, 207, 208, 219, 222, 241, 242 M Marine: Piling, 268 - 271 Ground Improvement, 282, 283 Construction, 284 - 286 Mod AASHTO Test, 29 Modulus of: Compressibility, 7, 8, 24, 28, 29, 59 - 62 Sub-grade Reaction, 63 Moisture Content, 7, 8, 29 Moisture Density Relationship, 2, 29 Movement Associated with Excavation, 390, 391 H Heave: Pile, 77, 84, 90, 96 Soil, 48, 74, 101, 314 - 316 Heave Tests, 7, 8 Hydrometer, 29 418 N Pile Types and Classification, 67 - 69 Piling Equipment: Normal Plant Clearance Requirements, 396, 397 Rig Dimensions, 398 - 410 Piston Samples, 13 Plastic Limit, 29, 51 Plasticity Index, 29, 51 Plate Load Test, 7, 18, 19 Point Load Index Test, 57 Precast Piles, 67 - 69, 86 - 92 Pre-drilling, 75, 77, 89 Pressuremeter Test, 7, 8, 18 Problem Soils, 250 - 263 Procter Test, 29 Prop Supports, 232, 233 Post-stressed Anchors, 234 - 240 Negative Friction, 89, 260 Noise Pollution, 69, 70, 80, 86, 93, 97, 105, 114, 119, 129, 134 O Oedometer Test, 7, 8, 28 Oscillator Piles, 67 - 69, 129 - 133 P Particle Size, 33 Passive Earth Pressure, 372, 373 Pedogenic Material, 34 Permeability, 7, 8, 29 Piezocone Penetration Test (CPTU), 25 Piezometers, 12, 15 Pile-caps: Design and Detailing, 339 - 343 Typical Details, 344 - 349 Pile: Bearing Capacity, 290 - 307 Behaviour, 288, 289 Capacity: Compression, 288 - 307 In Heaving Subsoil, 314 - 316 Lateral, 310 - 313 Tension, 308, 309 Classification, 67 Design Aids, 288 - 338 Driveability, 24, 65 Driving Formulae, 306, 307 End-bearing, 291, 294, 295, 298, 300, 303, 305 Factors of Safety, 317 Friction, 292, 293, 296, 297, 299, 300, 302, 305 Heave, 77, 84, 90, 96 Integrity Testing, 110, 131, 161 - 164 Load Testing, 156 - 160 Load Transfer Functions, 325 - 327 Penetrating Ability, 69 Selection, 65, 66 Settlement, 325 - 333 Shaft Capacity, 290, 292, 293, 296, 297, 299, 300, 302, 303, 305 Technical Details, 70 - 142 Tension, 308, 309 Pile-shaft: Bending Moment Curves, 350 - 355 Design, 334 - 338 Design Curves, 350 - 355 Reinforcement, 337 Q Quality Assurance, 392 - 395 Quay Walls, 268, 269, 272 - 280, 284 R Raymond Spoon, 20, 21 Reinforcement Areas, 344 Relative Density, 49, 263 Residual Soils, 34 Reticulated Micropiles, 246, 247, 387, 388 Rock, 13, 33 Rock: Classification, 38 Colour, 38 Cores, 12, 13, 38, 42, 43, 56 Discontinuities, 40 - 43 Fabric, 39, 40 Hardness, 40 Stratigraphy, 40 Strength Classification, 56 - 58 Type, 42 Weathering, 39 Rock Mass Description, 38 - 43 Rock Quality Designation (RQD), 42 Rock Shoes, 81, 88 Rock-socket Design, 299, 300 Rotary Core Drilling, 12, 13 Rotary Drilling, 12, 13 Rotary Percussion Drilling, 15 Rotapiles, 67, 68, 134 - 137 419 S Soil Dowelling, 248, 388 Soil Improvement: 165 - 201 Classification, 167 Design Aids, 356 - 368 Selection, 165, 166 Technical Details, 170 - 201 Soil Nails, 241 - 245, 384 - 386 Soiltech, 4, 9, 27, 264 Soldier Piles, 218 - 223 Steel H-piles, 67 - 69, 93 - 96 Steel Sheet Piles, 201 - 216, 272 - 275 Standard Penetration Test: 12, 20 SPT ‘N’ number, 50, 51, 54 - 56, 59 - 61 304, 305 Stone Columns, 167 - 169, 190, 191, 363 - 366 Sub-grade Modulus, 63 Surcharge Loads, 374, 375 Swelling Potential, 48 Swell Under Load Test, 7, 8, 28 Samples: Block, 7, 8, 28, 29 Disturbed, 7, 8, 28, 29 Piston, 13, 28, 29 Raymond Spoon, 20, 21 Shelby Tube, 13, 28, 29 Undisturbed, 6, 7, 8, 28, 29 Sand Drains, 192, 193 Sandwick Drains, 193 Secant Pile Wall, 204, 208, 209, 224 - 226 Shear Box Test, 28 Shear Strength, 16, 20 - 25 Sheet Piling, 204, 208 - 217 Shelby Tube Samples, 4, 13, 28, 29 Site Investigation, 2, 4 - 27 Slipways, 287 Slip Coating of Piles, 89 Slurry WaIls, 266, 267 Soft Clays, 259, 260 Soil: Classification: 32 - 37, 44 - 48 From Cone Penetrometer, 47 Standard, 32 - 37 Cohesionless, 51, 54 Cohesive, 50, 55 Colour, 31 Compaction, 170 - 183 Compressibility, 59 - 61 Consistency, 49 - 56 Consolidation, 168, 192 - 194 Densification, 170 - 182 Description, 7, 31 - 36 Expansive, 48, 251 - 255, 314 - 316 Grading, 29 Liquefaction, 263 Moisture, 31 Origin, 34 Pedogenic, 34 Pressures, 371 - 373 Profiling, 31 - 36 Profile, Typical, 36 Profiling, Symbols, 37 Replacement, 184 - 189 Residual, 34 Sampling, 13 Strength Classification, 50, 51 Structure, 32 Transported, 34 Type, 33 T Tension Test Load, 159 Test Pits, 2, 4, 10 Tie-backs, 204, 208, 234 - 240 Tied-back Walls, 378 - 383 Titan Anchors, 240 Titan Micropiles, 139, 142 Transported Soil, 34, 35 Tremie Concrete, 107, 108, 131 Trial Piles, 157 Triaxial Compression Test, 28, 29 Tube-a-manchette, 238 U Unconfined Compression Strength (UCS), 53, 56, 57, 298 - 300 Underpinning, 143 - 155 Underreams, 101 Underslurry Piles, 67, 68, 105 - 113 Undisturbed Samples, 7, 8, 13 Undrained Cohesion, 50, 52, 55 Undrained Shear Strength, 28, 29, 52, 55 Unified Soil Classification System, 44, 46 420 V Vane Shear Test, 7, 8, 16, 17 Vibration, 69, 70, 78, 80, 84, 86, 90, 168 Vibratory Compaction, 167 - 169, 170 - 174, 356, 357 Vibratory Replacement, 167 - 169, 184 - 186, 363 - 365 Vibrocompaction, 167 - 169, 170 - 174, 356, 357 W Walers, 219, 221, 332 Washboring, 12 Water Pressure, 374 Water-table, 7, 8, 35 Wave Equation, 307 Wedge Analysis, 381, 385 Wells, 264 Y Y-probe Compaction, 172, 173 421 LIST OF SYMBOLS A B b c cu D d E Ep e f G Go H I K k l M N n P Q q R r s V W z α β δ φ γ ν π θ ρ σ τ υ Area of pile (base or shaft) Breadth of pile group Breadth of pile Recoverable pile movement (during driving) Undrained shear strength Diameter Pile diameter Modulus Young’s modulus of an equivalent solid pile of the same radius Exponent in calculation of group stiffness efficiency, voids ratio Unit friction Shear modulus of the soil Shear modulus at ground surface Lateral load on piles Influence factor Earth pressure coefficient Efficiency of pile driving hammer; stiffness of pile, cap or foundation; coefficient of sub-grade reaction Embedded length of the pile Applied moment Bearing capacity factor Number of piles; rate of increase with depth of coefficient of sub-grade reaction Axial force, usually at working load, vertical pressure Axial force, usually ultimate capacity Pressure (eg end-bearing) or resistance (eg cone) Resistance (pile driving); correction factor, settlement or deflection ratio Ratio, radial co-ordinate Centre to centre pile spacing; set of pile during driving Shear force Soil moisture content Depth, co-ordinate perpendicular to ground surface Ratio of skin friction to undrained shear strength; interaction factor Ratio of skin friction to effective overburden pressure; departure angle between direction of loading and line joining pile centres; angle of retained soil surface to horizontal Interface angle of friction Angle of friction for soil Unit weight of soil (dry, bulk, saturated) Poisson’s ratio for the soil Mathematical constant Rotation of pile-head under lateral loading Settlement of a foundation Normal stress Shear stress Shear stress in concrete 422 PRINCIPAL SUBSCRIPTS b f h o p r s t v c’ cw H N R α β denoting base of pile foundation horizontal denoting value at the pile shaft pile radial denoting shaft of pile denoting top of pile vertical Effective cohesion Wall adhesion Height Standard Penetration Test result (blows/ 300mm) Resultant force Inclination of wall to horizontal Angle of retained soil surface to horizontal 423

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