Tomasz Siwowski
Tomasz Siwowski Bridge Engineering. Selected Issues.
Tomasz Siwowski – dr hab. inż., profesor PRz jest Kierownikiem
Zakładu Dróg i Mostów na Wydziale Budownictwa, Inżynierii
Środowiska i Architektury Politechniki Rzeszowskiej. Jego
działalność naukowa skupia się wokół problemów zastosowań
nowych materiałów w mostownictwie (np. kompozyty FRP, stopy
aluminium, betony niekonwencjonalne,) współczesnych
technologii budowy i utrzymania mostów oraz diagnostyki i badań
obiektów mostowych. Jest autorem ponad 150 publikacji z tego
zakresu. Zajmuje się także działalnością konsul ngową w zakresie
projektowania i zarządzania inwestycjami infrastrukturalnymi. Jest
projektantem kilkudziesięciu obiektów mostowych.
Bridge Engineering
Selected Issues
Rzeszów 2015
Wydano za zgodą Rektora
Podręcznik opracowano w ramach projektu pt.: „Kształcenie innowacyjnych kadr GOW
w Politechnice Rzeszowskiej” – umowa nr UDA-POKL.04.03.00-00-036/12-00,
realizowanego w ramach Programu Operacyjnego Kapitał Ludzki, Priorytet IV.
Szkolnictwo wyższe i nauka, Działanie 4.3. Wzmocnienie potencjału dydaktycznego
uczelni w obszarach kluczowych w kontekście celów Strategii Europa 2020.
Opiniodawca
prof. dr hab. inż. Wojciech RADOMSKI, dr h.c.
prof. dr hab. inż. Jan BIEŃ
W procesie wydawniczym pominięto etap opracowania językowego
w Oficynie Wydawniczej Politechniki Rzeszowskiej.
Podręcznik wydrukowano z matryc dostarczonych przez Autorów.
Korekta językowa
Katarzyna KOWALSKA
bridge construction
bridge maintenance
advanced materials
Publikacja przekazywana nieodpłatnie
© Copyright by Oficyna Wydawnicza Politechniki Rzeszowskiej
Rzeszów 2015
ISBN 978-83-7934-036-1
Oficyna Wydawnicza Politechniki Rzeszowskiej
al. Powstańców Warszawy 12, 35-959 Rzeszów
e-mail: oficyna1@prz.rzeszow.pl
Nakład 60 egz. Ark. wyd. 14,71. Ark. druk. 9,75. Papier offset. 80g B1.
Oddano do druku w grudniu 2015 r. Wydrukowano w grudniu 2015 r.
Drukarnia Oficyny Wydawniczej, al. Powstańców Warszawy 12, 35-959 Rzeszów
Zam. nr 164/15
TABLE OF CONTENTS
Abbreviations ....................................................................................................... 5
PREFACE ............................................................................................................ 6
1.
Concrete bridges construction ................................................................... 8
1.1. CONSTRUCTION ON SCAFFOLDS OR FALSEWORK .................................................................................. 10
1.2. SUPERSTRUCTURE MADE BY PRECAST BEAMS AND SEGMENTS............................................................. 12
1.3. SPAN-BY-SPAN METHOD ...................................................................................................................... 20
1.4. CANTILEVER METHOD.......................................................................................................................... 25
1.5. INCREMENTAL LAUNCHING .................................................................................................................. 33
1.6. CASE STUDY NO.1. CONSTRUCTION OF A BRIDGE ON SCAFFOLDS ........................................................ 38
1.7. CASE STUDY NO.2. CONSTRUCTION OF A CONCRETE BRIDGE USING BALANCED CANTILEVERING........ 42
2.
Steel bridges construction......................................................................... 46
2.1. CONSTRUCTION WITH A CRANE FROM THE GROUND............................................................................. 47
2.2. INCREMENTAL LAUNCHING .................................................................................................................. 49
2.3. CANTILEVER ERECTION ....................................................................................................................... 52
2.4. HEAVY LIFTING.................................................................................................................................... 56
2.5. PLACEMENT OF THE COMPLETE SPAN OR LARGE BRIDGE ELEMENTS .................................................... 58
2.6. CASE STUDY NO.1. ERECTION OF THE STEEL SUPERSTRUCTURE BY CRANE ......................................... 61
2.7. CASE STUDY NO.2. ROTATION OF THE LONG ARCH SPAN ..................................................................... 64
2.8. CASE STUDY NO.3. INCREMENTAL LAUNCHING OF CABLE-STAYED SUPERSTRUCTURE ........................ 70
3.
Maintenance techniques for concrete bridges ........................................ 75
3.1. CONCRETE REPAIR ............................................................................................................................... 76
3.2. STRENGTHENING & REHABILITATION .................................................................................................. 80
3.3. STRENGTHENING WITH EXTERNALLY BONDED FRP COMPOSITES ........................................................ 87
3.4. CASE STUDY NO.1. STRENGTHENING BY SECTION ENLARGEMENT ....................................................... 92
3.5. CASE STUDY NO.2. REHABILITATION OF PRECAST CONCRETE BRIDGE ................................................. 94
3.6. CASE STUDY NO.3. TWO CONCRETE BRIDGE STRENGTHENING METHODS – COMPARISON .................... 98
3.7. CASE STUDIES NO.4. STRENGTHENING CONCRETE BRIDGES WITH PRESTRESSED CFRP STRIPS ......... 102
4.
Maintenance techniques for steel bridges ............................................. 108
4.1. REPAIRS & REPLACEMENT ................................................................................................................. 109
4.2. STRENGTHENING & REHABILITATION ................................................................................................ 114
4.3. BRIDGE REDECKING WITH LIGHTWEIGHT DECK.................................................................................. 122
4.4. CASE STUDY NO.1. BRIDGE STRENGTHENING WITH PRESTRESSED CFRP STRIPS ............................... 126
4.5. CASE STUDY NO.2. TRUSS BRIDGE REHABILITATION ......................................................................... 128
4.6. CASE STUDY NO.3. REHABILITATION METHOD SELECTION BASED ON LCCA .................................... 134
5.
Maintenance techniques for masonry bridges ..................................... 143
5.1. REPAIRING TECHNIQUES .................................................................................................................... 144
5.2. STRENGTHENING TECHNIQUES ........................................................................................................... 146
5.3. CASE STUDY NO.1. REHABILITATION OF 19-CENTURY MASONRY VIADUCT ....................................... 150
5.4. CASE STUDY NO.2. RECONSTRUCTION OF THE BRICK ARCH VIADUCT ................................................ 158
3
6.
Advanced materials in bridge engineering ........................................... 161
6.1. ALUMINIUM ALLOYS .......................................................................................................................... 162
6.2. FRP COMPOSITES ............................................................................................................................... 169
6.3. GLULAM TIMBER................................................................................................................................ 179
6.4. SELF-COMPACTING CONCRETE ........................................................................................................... 182
7.
4
References ................................................................................................ 186
Abbreviations
ABC - accelerated bridge construction
BCC - bridge cantilevered construction
CFA - continuous flight auger (piling)
CFRP - carbon fibre reinforced polymer (composite)
CFST - concrete filled steel tube
DRA - discontinuously reinforced aluminium
EMPA - the Swiss Federal Laboratories for Materials Science and Technology
FEM - finite element method
FRP - fibre reinforced polymer (composite)
GDDKiA – Generalna Dyrekcja Dróg Krajowych i Autostrad (ang. General Directorate of
Polish Roads and Highways)
GFRP - glass fibre reinforced polymer (composite)
HDPE - high-density polyethylene
HL - heavy lifting (construction method)
HPC - high-performance concrete
HPS - high-performance steel
HSFG - high strength friction grip (bolt)
LCC - life-cycle cost
LCCA - life cycle costing analysis
MMC - metal-matrix composite
MSS - movable scaffolding system
NPV - net present value
OM&R - operation, maintenance & repair (period)
PCC - polymer cement concrete
PTFE - poli(tetrafluoroetylen), Teflon
RC - reinforced concrete
RUT - Rzeszów University of Technology, (Politechnika Rzeszowska)
SCC - self-compacting concrete
TIG - tungsten inert gas (welding)
VARTM - vacuum assisted resin transfer moulding (process)
5
PREFACE
The construction and maintenance of bridges is a highly specialized field in engineering.
Bridges are not a new concept, they have existed even in ancient times. But the sheer variety,
design, and construction methods of bridges have seen an exponential growth in the last century.
Bridges may vary widely by the method of construction, structural design, material, the purpose
for which they are built, etc. The location of the bridge and the purpose it serves could also
influence the construction methods and applied materials as well as maintenance techniques.
The bridge engineering industry creates an ever-increasing variation of challenges. Bridge
engineering has changed as the new millennium began. New construction and maintenance
techniques and new materials have been emerging. There are also new issues facing the bridge
building industry relative to the research needs associated with these new techniques and
materials.
Bridge engineering. Selected issues handbook is a comprehensive and state-of-the-art
reference work and resource book covering domestic practices in selected three major areas of
bridge engineering. The areas of bridge engineering include construction, maintenance actions
as repair, strengthening or rehabilitation, and advanced materials in bridges. They have been
selected as the supplement for bridge design issues which are the subject of the author’s former
script Basis of bridge design, where most of the background information referring to bridge
engineering can be found, f.e. basic definitions, classification of bridges, main bridge
components, traffic loads on bridges and bridge planning issues. Both books cover almost all
of the aspects of bridge engineering and have been written with civil engineering students in
mind. The ideal readers will be the third year civil and/or structural engineering students with
the need for a single reference source to keep abreast of the state-of-the-practice and new
developments, as well as to review standard practices in the field of bridge engineering.
To provide students with a well-organized, user-friendly, and easy-to-follow resource,
the handbook is divided into six main chapters. Chapter 1, Concrete bridges construction,
presents five concrete bridge construction methods, and two case studies covering most
conventional concreting operations for bridge construction. Chapter 2, Steel bridges
construction, reviews how to build various bridges of steel, and presents three case studies of a
typical projects. Chapter 3, Maintenance techniques for concrete bridges, addresses the various
methods of concrete bridge upgrading, from small material repairs to comprehensive
rehabilitation. Three case studies complete the chapter with the comparison of interesting
methods as the final conclusion. Chapter 4, Maintenance techniques for steel bridges, provides
the same scope as the preceding chapter, and describes three case studies supplemented with
the life cycle costing analysis (LCCA) in one case to juxtapose the methods. Chapter 5,
6
Maintenance techniques for masonry bridges, presents repairing and strengthening techniques
for old brick or stone bridges with two interesting case studies, showing the comprehensive
revitalization of the 19-century masonry bridge. Finally, Chapter 6, Advanced materials in
bridge engineering, addresses four advanced materials, namely aluminium alloys, fiber
reinforced polymer (FRP) composites glulam timber and self-compacting concrete, and their
recent developments in bridge engineering.
Bridge engineering. Selected issues ties the material to the Polish context, discussing
the methods and practices of bridge engineering applied in Poland. The author has also drawn
on his own experience to elucidate various concepts. The handbook stresses professional
applications and practical solutions, mainly based on the author’s own experience. Real
domestic case studies and dozens of helpful illustrations are also included in this practical
resource. Emphasis has been placed on construction technologies and maintenance techniques,
and special attention is given to advanced materials in bridge engineering. The handbook
describes the basic concepts and assumptions, omitting the derivations of formulas and theories,
and covers both traditional and new innovative practice. References at the end of the handbook
can be consulted for more detailed studies.
In general, the handbook targets the needs of undergraduate students, but the material
may be reorganized to accommodate graduate bridge courses and practising engineers. The
book along with previously issued script Basis of bridge design can serve as the material for a
two-semester course in bridge engineering at undergraduate courses (bachelor degree). It can
also be helpful as a reference material for the professionals who have just started their careers.
And last but not least, the book may also be used as the review of the recent Polish bridge
engineering achievements.
The author acknowledges with thanks the invaluable assistance of PhD student
Agnieszka Wiater in the development of the handbook as well as Katarzyna Kowalska for her
support in English language verification.
7
1. Concrete bridges construction
The technological aspects of construction influence the modern bridge industry from the
very first steps of the design. Entire families of concrete bridges, such as launched bridges,
span-by-span bridges, and balanced-cantilever bridges, take their names straight from the
construction method. Generally these methods can be divided in two main groups: cast-in-situ
construction and precast construction (prefabrication). In basic terms, cast-in-situ construction
describes a process whereby segments are progressively cast on site in their final positions
within the structure. By comparison, for precast construction, the segments are prefabricated at
a casting plant – either on site or at a remote facility – then transported to the project site and
erected as completed units in their final positions.
Each method of constructing concrete bridges has its own advantages and challenges.
In the absence of particular requirements that make one solution immediately preferable to the
others, the evaluation of the possible variants is a difficult task. Comparisons based on the
quantities of materials consider only one of the components of the construction cost of a bridge.
In industrialized countries, the cost of a bridge is more and more influenced by the processing
costs of the materials, such as labour, investments for specialty equipment, delivery and
assembly costs for the equipment, and energy. The main components of the entire cost per unit
m2 of a concrete bridge superstructure is shown in Table 1.1.
Table 1.1. Cost in US dollars per unit m2 of a concrete bridge superstructure
Construction method
Conventional
scaffolding/falsework
Precast beams erection
Cast-in-situ
span-by-span
Cast-in-situ
balanced cantilever
Precast
balanced cantilever
Incremental launching
Typical span
length (m)
Material cost
Labour cost
Facilities /
machines cost
Total cost
25
110
110
50
260
25
110
80
50
240
40
110
60
70
240
120
150
90
80
320
80
130
80
90
300
40
140
50
40
230
Safety, serviceability, cost-effectiveness, aesthetics and particular technical issues are
typically the controlling factors in the selection of the proper concrete bridge construction
method. The selection is further complicated by other considerations such as the deflection
limit, the life-cycle cost, including traffic disruptions during construction stages, scheduling
8
and feasibility of falsework layout. In many cases, a prestressed concrete bridge is a costeffective choice. Typically, segmental concrete bridge construction is utilized, which is the
most common method of bridge construction.
Segmental construction method typically introduces: (a) the conventional cast-in-situ
bridge construction, (b) the precast prestressed beam deck construction with continuous castin-situ deck slabs, (c) the balanced cantilever bridge construction, which either utilises
scaffolding (traveller) or precast deck segments and (d) the progressive and span-by-span
incrementally launched bridge construction. Segmental cast-in-situ bridge construction is
preferable in case of straight and curved in plan bridges with relatively small bent heights and
when prestressing is applied in the longitudinal direction of the superstructure. The formworks
are typically supported directly to the ground or to a well compacted temporary embankment.
Hundreds of bridges have been built by segmental precasting even though the need for avoiding
joint decompression increases the cost of prestressing. However, the investments in specialty
equipment are also high, so segmental precasting is typically used for long bridges that allow
amortization of precasting facilities and erection machines. Construction of precast concrete
bridges with spans ranging from 30 m to more than 180 m is mostly based on the use of selflaunching machines. On shorter bridges, prefabrication is limited to the concrete beams and the
deck slab is cast in place. Typical bridge span lengths in relation to construction method are
provided in Table 1.2.
Table 1.2. Typical concrete bridge span ranges for particular construction method
Construction method of concrete bridge
Conventional scaffolding/falsework
Cast-in-situ span-by-span
Precast segmental span-by-span
Cast-in-situ balanced cantilever
(girder bridges / arch bridges)
Precast balanced cantilever
(girder bridges / arch bridges)
Incremental launching
Range (m)
up to 80
20 - 45
25 - 50
70 – 250 / 50 - 200
45 -135 / 50 - 200
30 - 60
Precasting or prefabrication presents many advantages over traditional construction for
bridges. They have resulted from the industrial approach and favourable working environment,
protected from adverse weather conditions. For many years, the profession has been familiar
with the principles and methods aiming at mastering the quality of the precast products. Each
plant has a self-control system, defining the working procedures and modules for internal
inspection and control. Among the many advantages, the following are of particular interest to
bridge construction:
 quality and regularity of the concrete strength
 elements with elaborated shapes designed to get a maximum benefit from the materials and
the prefabrication; they require the use of relatively complex moulds, but at the same time
enable high quality surfaces with respect to shape, texture, dimensional tolerances, etc., to
be achieved
 absence of laborious and hindering scaffolding
9
shorter construction time as that precast elements are made at the plant independent from the
foundations and other preparatory works on site.
The basic concrete bridge construction methods in both cast-in-situ and precast variety
are described in this chapter with supplement of two case studies, presenting the practical
application the two of them.

1.1. Construction on scaffolds or falsework
The most basic way of erection is the cast-in-situ method and the use of scaffolds or
falsework along with the formwork to support the construction of a bridge (Fig.1.1). Cast-insitu method of bridge construction is a flexible method in which the demands of more unusual
geometrical shapes can be easily met. This method is commonly used for short span bridges for
the cost effective construction of solid, voided (cellular) or ribbed reinforced concrete slab
bridges. Each bridge type is designed to allow each span to be cast in one continuous pour.
Construction is simplistic in form consisting of a birdcage scaffold with plywood formwork.
Figure 1.1. Bridge constructed with scaffolding (top) or with falsework (bottom).
A bridge is built on-site using formwork supported by temporary scaffolds or falsework.
This technique gives high flexibility to the design of the bridge, since the succeeding segment
of the superstructure will be supported until the construction is completed. Formwork creates
the shape of the concrete section and any internal voids or diaphragms. This means that the
most slender and effective bridge designs are often constructed using scaffolding or falsework.
When a bridge is constructed using scaffolds closely spaced cross-braced struts are used. These
are designed to support the load directly from the formwork down to the ground. The scaffolds
are most often combined of standardized reusable cross-braced modules of steel that can be
used over and over again (Fig.1.2). However, scaffolds of timber are often used because they
are flexible and cheap.
10
Figure 1.2. Standardized reusable scaffold combined of cross-braced modules of steel
Figure 1.3. Scaffolding (left) and falsework (right)
The main difference between scaffolding and falsework is that falsework requires larger
and heavier temporary supports for the superstructure (Fig.1.3). Instead of supporting the load
via a large number of cross-braced struts, steel beams between temporarily established supports
are used. Spans can have a range of 10 to 20 m. Falsework can therefore be used when obstacles
like small rivers or roads need to be crossed over. On the other hand, scaffolds can only be used
if it is possible to place the temporary steel posts all the way under the superstructure. Both
methods mentioned above are suitable to use if the free height under the bridge is not greater
than up to 6 m.
Formwork is the temporary or permanent moulds into which concrete is poured
(Fig.1.4). In the context of concrete bridge construction, the scaffolding or falsework support
11
the shuttering moulds. The contemporary engineered formwork system is made up of
prefabricated modules with a metal frame (steel or aluminium) and covered on the concrete side
with material having the wanted surface structure (steel, aluminium, timber, plywood, etc.). The
two major advantages of formwork systems, compared to traditional timber formwork, are
speed of construction (modular systems pin, clip or screw together quickly) and lower life-cycle
costs (bearing major force, the frame is almost indestructible, while the covering, if made of
wood, may have to be replaced after a few - or a few dozen - uses, but if the covering is made
of steel or aluminium the form can achieve up to two thousand uses depending on care and the
applications). Different formwork systems provide a wide range of concrete construction
solutions that can be chosen to suit the needs of a particular development. The main types of
formwork systems in use now are: table form/flying form, system column formwork, horizontal
panel and slip form. The modern formwork systems listed above are mostly modular, which are
designed for speed and efficiency. They are designed to provide increased accuracy and
minimize waste in construction and most have enhanced health and safety features built-in.
Figure 1.4. Typical steel (left) and timber (right) formwork
1.2. Superstructure made by precast beams and segments
In precast bridge construction, the individual components (beams or segments) are
manufactured off-site and assembled on-site. This method usually increases the component
durability, reduces on-site work and construction time, minimises traffic disruption, and lowers
life-cycle costs. Bridge superstructures made of precast beams are widely used. Precasting gives
the benefit of good control of the quality of construction, while the deck construction is simple
and repetitive. Such decks are usually statically determinate, although techniques are available
to make them partially or totally continuous. Statically determinate beams are frequently
economically competitive with continuous beams, due to the lack of secondary sources of
bending moment and locked-in stresses. Modern techniques which allow several spans to be
linked to minimise the number of expansion joints have eliminated one of the main
disadvantages to the use of statically determinate spans.
12
There are many forms of precast beams, including various forms of I-beams, boxes and
troughs. One may divide precast beams for bridge decks into two categories: standard precast
beams and custom-designed precast beams. Standard pre- and post-tensioned precast beams
such as the T, inverted T, I and U beams are widely used worldwide for bridges with spans that
are generally below 30 m, but which may attain 40 m for the most powerful units. They are
very useful when it is impossible to use falsework, such as for bridging railways or motorways
under traffic. They are also useful for small bridges where they may be built by contractors with
little bridge expertise.
Figure 1.5. Small bridge decks constructed with precast units
Figure 1.6. The precast Kujan beams for precast bridge deck system
Small bridge spans can be constructed with precast beams and cast-in-situ deck slabs,
acting together as a composite structure. In a more advanced solution, the span is composed of
I-shaped, inverted T-profiles or heavy double-tee units, placed side by side, and connected with
a cast in-situ topping and infill concrete. The additional reinforcement of the cast in-situ part
comprises a transversal reinforcement through openings in the webs of the beams and a top
reinforcement above the beams. The system is suited for bridges with a span length between
13
approximately 6 and 20 m. The edge of the bridge can be realised with a precast side profile or
a cast in-situ cantilevering slab. Figure 1.5 shows systems, which are used in some countries of
Western Europe and figure 1.6 shows the most popular Polish system (Kujan beam). The
solution results in heavy, but very durable bridge spans.
Large beam bridges constitute the main solution for precast bridges built from the
1960th. The bridge deck is composed of several T, inverted T or I-shaped beams positioned at
a certain distance. The beams are connected by a transversal diaphragm beam at each support
and sometimes also in the middle of the span, depending on the length. After erection of the
beams and casting of the diaphragm beams, a deck slab is cast on site. The top of the beams has
protruding reinforcement for the connection with the deck slab. The system is used both for
simply supported and continuous bridge structures. The system with T-beams is suitable for
span lengths between approximately 15 to 27 m (Fig.1.7). The bridge system with I-shaped
beams is suitable for span lengths between approximately 18 to 55 m (Fig.1.8). The distance
between the units varies and is the function of the needed span/load capacity.
Figure 1.7. The precast T-beams for precast bridge deck system
Figure 1.8. Cross section of a bridge deck with I-shaped precast beams
The great majority of customised precast beams are T-beams although there have been
a few very significant bridges built with customised U-beams. Customised precast T-beams are
one of the most traditional forms of prestressed concrete span and they remain versatile and
economical (Fig.1.9). These essentially statically determinate beams are between 1,5 m and 3
m deep, and are typically used for spans between 25 m and 45 m, although longer spans also
have been built. In general, the most economical span-depth ratio of such superstructures is
14
about 15, although they are still viable at a ratio as high as 20. They are usually post-tensioned,
with a roughly parabolic prestress centroid that balances a large part of the shear force as well
as the bending moments (Benaim, 2007).
Precast segmental superstructures are regularly used in traditional cantilever
construction of large span bridges. The segments comprise the full width of the bridge and the
length of the units is related to their weight and to the means of transportation and lifting. Large
segments are usually precast on site. The length and weight of the precast concrete segments
for box-girder bridges are usually governed by handling and transportation requirements.
Lengths up to 3,5 m are often transportable on public roads without excessive restrictions. If
the precasting plant is close to the construction site and no transportation restrictions exist, the
segments are made as long as practical, but they rarely exceed lengths of about 4,5 m.
Figure 1.9. Customised “U” beams for railway viaducts (dimension in cm)
The erection is mostly done on a temporary scaffolding or – in a span-by-span method
– on launching gantries. After filling the transversal joints, the units are post-tensioned in the
longitudinal direction. With the span-by-span method, all of the segments for a span are
positioned before the prestressing tendons are installed, and the complete span is lowered onto
the bearings (see p.1.3).
The ultimate form of prefabrication for the construction of bridge superstructures is
building a complete span in a casting yard and then transporting it to the construction head and
launching into place (macrosegmental erection). The bridge is composed of a large trapezoidal
beam (mono-box) with cantilevering or braced cast in-situ deck slab (Fig.1.10). The bridges
can be designed as continuous structures, with spans up to 90 m and more. For reasons of
handling and transport, the size of the single box beam is limited to about 45 m. When longer
spans are needed the bridge is constructed with several beams, made continuous by posttensioning. The segment weight is excessive for ground cranes and also for most gantries, so
special twin-upper-beam units are used for macrosegmental erection (Fig.1.11). The segments
are transported along the completed deck. The length and weight of the segments are so big that
the gantry cannot rotate them, so the segments are delivered with their final alignment. The
15
complexity of the operations and the cost of gantry are really huge so macrosegmental
construction is typically used for long parallel bridges low to the ground. This type of bridges
is a more complex way of building precast bridges but the system enables building longer spans
than with normal beams (either simply supported or continuous). Up to now several bridges
have been constructed with spans ranging from 50 m to 90 m.
Figure 1.10. Cross-section of mono-box bridge
Figure 1.11. Twin-upper-beam launching gantry used for macrosegmental erection:
scheme (top) and gantry at work (bottom, source: bbesols.com, 12.12.2015)
16
In the early days of precast bridge construction, it was considered logical to design the
bridge decks as simply supported structures with transversal joints at intermediate spans and
between end spans and abutments. The beams were normally positioned on individual bearings
– one at each beam end. The most economical way of supporting the beams is to rest each of
them on a laminated rubber bearing carried by a crosshead of a pier, located beneath the deck
(Fig.1.12a). It does not look good, and a variation is to design a crosshead of inverted T-cross
section, that is mainly within the thickness of the deck, with a bearing ledge protruding below
(Fig.1.12b).
Figure 1.12. Supporting the precast beams: (a) crosshead located beneath the deck or
(b) crosshead of inverted T-cross section
Many thousands of bridges and viaducts have been built in this way, but they do not
behave very well. Although the beams themselves have proved to be very successful, there are
disadvantages inherent in simply supported span systems. The main problems with simply
supported spans originate from the presence of expansion joints: their long-term durability
versus de-icing salt and discomfort to traffic (see p.3.5). It is quite obvious that the best solution
to prevent these problems consists in eliminating the transversal joints within the bridge deck,
either by continuous deck systems or by integral bridges.
Partial continuity is a method to provide only continuity of the deck slab, the beams
being designed as simply supported. This means that no distribution of vertical load effects
between the intermediate bridge spans can occur. It applies to all vertical loads, including selfweight and variable loading. The solution shown in Figure 1.13 is a simple measure to provide
simply supported continuous decks, with a minimum of extra design and construction effort.
However, the Polish experience has revealed short-term durability resulting from the use of deicing salt and the lack of proper maintenance. Now this solution is not allowed in Poland.
Multi-span bridges with mechanical continuity between adjacent spans are realised by
integration of the bridge beams into a reinforced concrete crosshead on top of the piers. The
construction is done in two steps:
17


in the first step the beams are simply supported and carry their own weight plus the load
from the formwork and the wet cast concrete of the slab
in the second step, after hardening of the in-situ concrete, the structure becomes continuous,
but only for the additional dead load and the life loading.
Figure 1.13. Typical solution for partial continuity of precast beam spans
Different solutions are applied to realise the continuity. It is possible to make such decks
fully continuous by supporting the beams on falsework and casting in situ a section of a severalmetres-long deck over the piers (Figure 1.14). The continuity is usually created in reinforced
concrete. The cast-in-situ section may incorporate a crossbeam, usually within the thickness of
the deck, which eliminates the need for a pier crosshead and allows the deck to be supported
directly by the pier. It solves the problem with joints, reduces the problem of camber and
improves the appearance.
Industrialised precast bridge elements are manufactured in permanent facilities under an
officially agreed system of quality surveillance. The elements are generally cast on long line
prestressing beds, although for very large and very heavy units, such as mono-box girders and
some special units, other prestressing techniques may be used (fib, 2004). The moulds are
normally made of steel. If steam curing is adopted, it is feasible to achieve a concrete strength
of 20–25 MPa in the morning after casting. The transport of long and heavy bridge units may
require a special truck and selected itineraries. The transport is often done at night, in order to
limit the disturbance of the normal traffic. In some countries, transport is also done by ship and
by rail.
Figure 1.14. Cast-in-situ crosshead to realize the span continuity
18
Precast concrete beams or segments can be erected with ground cranes if the piers are
not too tall and the area under the bridge is accessible (Fig.1.15). In most cases, the erection of
bridge beams is done with mobile cranes, standing and operating on the ground below or on the
approaching road or span (Fig.1.16). The crane capacities now reach up to 600 ton. Very often
two cranes are needed to hoist the long bridge elements up to their final position. Additional
site work includes the preparation and casting of additional structural components such as
diaphragm beams, infill concrete and deck slabs, plus finishing works. Sensitive environments,
valleys with steep slopes, tall piers, and inhabited areas often require assembly with a launching
gantry, and in such a case the technological costs increase significantly.
Figure 1.15. Erection of precast beams with mobile crane: standing and operating on the
ground below (left) or on the approaching road (right)
Figure 1.16. Erection with mobile crane standing and operating: (a) on the approaching span
or (b) on the ground below
19
1.3. Span-by-span method
Cast-in-situ span-by-span construction is applicable to decks with spans that lie
generally between 20 and 45 m. Due to the difficulties of casting boxes in-situ, spans are most
commonly solid or voided slab or twin rib, and are usually continuous. However, some
contractors have special skills in casting boxes in situ. Bridges built with this method may be
only three spans long, or may extend for many kilometres. Consequently, the falsework varies
from the simplest to the most highly mechanised self-launching rigs. The falsework may be a
simple scaffold, a self-launching falsework rig called movable scaffolding system (MSS), or
something in between (Fig.1.17).
The span-by-span construction of precast concrete segmental box girders is well adapted
to long bridge structures with spans that generally do not exceed 50 m. The decks may be simply
supported or continuous. The segment joints may be glued with internal prestress or dry with
external tendons. Upper- or lower-beam gantries are used in the span-by-span erection to
support a complete span of segments, which are pulled together by prestressing bars during
gluing of the joints and then by the permanent tendons. The gantry then releases the span onto
the bearings and launches itself forward to erect the next span.
Movable scaffolding systems
The movable scaffolding systems (MSSs) were developed after the introduction of
prestressing technology for concrete bridges to raise efficiency of bridge construction using
mechanisation in some steps of the construction cycle. These systems help to build multiplespan cast-in-situ bridges with a variable degree of independence from the ground works
regarding their dismantling and resetting in successive spans. The working principle of this fullspan construction method is based on the idea of moving formwork and its supporting structure
from under the previously cast bridge span to the next pier and aims at minimising labour and
lifting equipment costs in all stages between casting one span and the following one.
Figure 1.17. Span-by-span construction: a self-launching falsework rig for cast-in-situ
construction (left, source: strukturas.no) and a launching gantry for precast concrete
segmental box girders (right, source: zzhz.com, 12.12.2015)
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The construction joint for span-by-span construction is normally situated at about 0,25
of the span. The MSSs can be applied for variable cross-sections, bearing systems, curvature
radii, span lengths and regardless of almost all topographic restraints. The equipment requires
a comparatively large quantity of material and time to be erected before it can start operating.
Therefore with this construction method, economic advantage can mainly be gained with
multiple span bridges. MSS is used for span-by-span construction of continuous beams.
Economically effective use is for bridge structures with length of up to 250 m. The MSSs can
also easily be adapted to any bridge cross-section (single box, double box, double T, etc.) and
span configuration. This flexibility provides the contractor the opportunity to re-use the
equipment from one project to another. For MSS the great speed of the construction process is
characteristic since it usually takes 11 days for completing each span.
In general, MSSs can be classified from the position of the main gantry girders related
to the superstructure (Fig.1.18). Underslung and alongside systems provide full clearance for
material feeding from above. Both MSS girders are supported directly on bridge piers or on
pier foundations. It makes MSS independent of geological and terrain conditions and allows for
surmounting a variety of natural and man-made obstacles under the bridge. This is the lightest
and most common type of MSS. With alongside systems, more vertical clearance under the
bridge can also be realized. Overhead systems – supported on the concrete span - need the
smallest space under the bridge when the formwork is closed. Feeding of materials
(reinforcement, etc.) cannot be executed with tower cranes because the formwork supporting
the construction obstructs deliveries from above. MSSs consist of five main components that
should be chosen according to the requirements resulting from bridge design, regional
characteristics and economic considerations: formwork, formwork supporting structure, maingirders, moving devices and supports (Fig.1.19).
Figure 1.18. MSS types: underslung (left), alongside (middle) and overhead (right)
On underslung systems, formwork construction can be adapted to changes of crosssection or span length of the bridge without interfering with other MSS construction parts. MSS
girder is situated under concreted bridge structure (Fig.1.20). Underslung MSS is especially
suitable for single beam bridge structures with either a linear or cranked bottom line. MSS
superstructure consists of two steel box girders, placed under the bridge superstructure and
21
supported by triangle brackets, fixed to both sides of the bridge pier column. On the brackets
are launching wagons with hydraulic equipment for vertical and horizontal MSS axis
adjustment. Formwork for bridge beam and for side cantilevers is fixed to MSS beams and real
adjustment of bridge beam geometry is realized by mechanical screw jacks. No other formwork
operations are necessary before, during and after launching MSS into the next span which
speeds up the work procedures (Fig.1.21).
Figure 1.19. Main components of MSSs: underslung (left) and overhead (right)
Figure 1.20. Underslung MSS system scheme
22
Figure 1.21. Underslung MSS system examples (source: strukturas.no, 12.12.2015)
Overhead MSS is suited especially for double beam bridge structures with either a
linear or cranked bottom line. MSS superstructure consists of one steel box girder, placed above
the bridge deck and supported by a steel frame, standing on the pier head and penetrating the
bridge slab. Transversally to MSS girder there are steel brackets, on which the bridge formwork
is suspended by thread rods. Formwork geometry is horizontally adjusted by the transverse
position of the rods and vertically by their length. At the front end of MSS is suspended the
prestressing platform, from which the strands are pulled into cable ducts and cables are
tensioned by jacks after concrete has hardened.
Figure 1.22. Overhead MSS system examples (right, source: strukturas.no, 12.12.2015)
Launching gantries for erection of precast segments
Alike in MSSs there are two types of launching gantries for erection of precast segments:
overhead gantries and underslung gantries. A typical overhead gantry (or twin-upper-beam)
is made up of two parallel trusses or box girders supported on crossbeams (Fig.1.23). The twinupper-beam units are easily adaptable to different span lengths, and they are able to cope with
variations in deck geometry. As the main girders are located above the deck, these units are less
affected by ground constraints; however, they are more complex to design, assemble, and
operate and the units are slower in erecting the segments than an underslung gantry. The total
length of the unit thus becomes about 2,3 times bigger than the standard span of the bridge
23
(Rosignoli, 2010). Crossbeams anchored to the piers support the main girders with saddles that
permit longitudinal launching and lateral movements of the unit. The support legs of the
crossbeams are adjustable to ensure that the frame is levelled. Hydraulic cylinders are used to
adjust the elevation, and safety ring nuts lock the cylinders during operation and launching. The
crossbeams are anchored to the pier cap with prestressing bars that resist uplift forces. The
crossbeams have lateral overhangs to set the gantry with the appropriate eccentricity and to
launch the unit on curved spans, so significant uplift forces may arise in the anchor systems.
Figure 1.23. Overhead launching gantries for precast segments: twin-upper-beam gantry made
up of two parallel trusses (left, source: strukturas.no) and single-upper-beam made up of box
girder (right, source: launching-gantry-operator.com, 12.12.2015)
Single-upper-beam gantries may also be used for span-by-span precast concrete
segmental erection. In these units, the carrying structure is a longitudinal girder that is supported
at the front pier of the span to be erected and at the rear pier (in the case of simply supported
spans) or on the front overhang of the completed deck (in the case of a continuous bridge). The
main girder may comprise two braced trusses or plate girders, or it may be a triangular truss
with one upper chord and two bottom chords. A winch trolley runs along the unit and moves
the deck segments to the assembly locations. A light front extension typically controls
overturning during launching. The rear end of the unit is supported by the completed span. No
rear nose is necessary, so these units are shorter than a typical twin-upper-beam gantry and are
more adaptable to curvatures in the bridge. The main girder is stiffer than two parallel trusses,
and its support systems are also stiffer.
The overhead gantries operate in one of two ways depending on how the deck segments
are delivered. If the segments are delivered along the completed deck, a winch trolley picks
them up at the rear end of the gantry, moves them over the span until reaching the assembly
location, and lowers them down to the deck level. If the segments are delivered at the ground
level, the winch trolley raises them up to the deck level. Typically, all of the segments for the
span are suspended from the gantry before the joints are glued. Epoxy is applied to groups of
segments that are then pressed together with temporary clamping bars. The permanent tendons
are usually tensioned from a stressing platform attached to the front segment.
24
Many precast concrete segmental bridges have also been erected with underslung
gantries (Fig.1.24). These units are positioned beneath the deck with the two trusses or box
girders on opposite sides of the pier, and the gantry supports the segments under the lateral
overhangs. The unit is typically supported on pier brackets or props from foundations. With
front and rear extensions, the length of the unit is more than twice the typical span length
(Rosignoli, 2010). The segments are placed onto the gantry with a crane or lifting frame. When
the segments are delivered through the completed deck, the lifter is placed at the rear end of the
gantry. When the segments are delivered on the ground, the crane is placed at the front end of
the gantry. The segments are placed onto the gantry close to the lifter and are moved along the
gantry to the assembly position with rollers. Upon completion of assembly and application of
prestress, the gantry lowers the span onto the bearings.
Figure 1.24. Underslung launching gantry (source: strukturas.no, 12.12.2015)
1.4. Cantilever method
Free cantilevering is a method of bridge construction where a superstructure is built
outward from a fixed point to form a cantilever structure, without temporary support, using
staged cast-in-situ or precast construction. When two opposing free cantilever structures are
attached as a single structure and erected in the same step, it is known as “balanced cantilever”.
The cantilever construction method is a modern technique, in which a structure is built segment
by segment above the ground level. As far as prestressed concrete is concerned, construction
with the cantilever method mainly applies to bridges whose decks can be combined with straight
or horizontally curved beams and which are built out from their piers, with cast-in-situ or
prefabricated segments . It should be mentioned that the cantilever method is often used for the
construction of cable-stayed and arch bridges (Fig.1.25).
This construction method consists in erecting the majority of a bridge span without
falsework or scaffolding on the ground, by working in consecutive sections known as segments,
each of which is cantilevered out of the preceding segment. After a segment is built, the
prestressing tendons fixed to the extremities are tensioned, firmly attaching them to the
25
preceding segments and thus forming a self-supporting cantilever (hammerhead) which serves
as a support for the subsequent operations (Fig.1.26).
Figure 1.25. Balanced cantilevering method for: (a) beam bridges, (b) portal bridges, (c)
cable-stayed bridges and (d) arch bridges
Figure 1.26. Balanced cantilever method
26



Construction is carried out:
symmetrically in general, either side of a pier in order to minimize the moments transmitted
to this support during erection; the resulting double overhang is called a balanced cantilever
(Fig.1.27a)
asymmetrically on a single side of a balanced cantilever, when the other side is already
joined to the adjacent span (Fig.1.27b)
more rarely from an abutment; in this case, the overturning moment exerted by the span is
compensated for by an appropriately dimensioned counterweight which forms part of the
deck itself.
Figure 1.27. Balanced cantilevering carried out: (a) symmetrically and (b) asymmetrically
The depth of the deck for large bridges is inevitably a compromise between the economy
of materials, the appearance and the ease of construction. Now, prestressed concrete box-girder
bridges built with the cantilever method are designed for spans between 70 and 250 m in length
(up to 300 m when using lightweight concrete). For spans of less than 80 m, bridges built with
the cantilever method are most commonly of a constant depth. For spans between 70 and 120
m in length, bridges built with the cantilever method may be of constant or variable depth. For
spans between 100 m and 200 m in length, bridges built with the cantilever method are nearly
always of variable depth. For spans less than about 100 m, for economy the depth of the deck
at the support should ideally be approximately L/14. Large bridges generally adopt a support
depth of about L/20 to overcome the disadvantages of the great structural depths otherwise
attained (Benaim, 2007).
Cast-in-situ balanced cantilever construction
Cast-in-situ balanced cantilever construction is ideally suited to box section bridges of
medium-size or long spans, where precasting is not justified due to the insufficient repetition.
The method becomes economical for bridges with the main span of 70 m and more, and remains
viable up to the largest span that may be built, currently about 300 m. The travellers (type of
falsework) used to build cantilever bridges are usually simple steel trusses that suspend a
platform on which the formwork is placed (Fig.1.28). They are counterbalanced so that they are
27
stable under their own self-weight, and are tied down to the deck to resist the weight of concrete.
The main design criterion for the truss is to limit its deflection as the concrete is cast.
Figure 1.28. Travellers used to build cantilever bridges: scheme (top) and examples (bottom)
Figure 1.29. Travellers on hammerhead to build a first deck segments
Typically, a hammerhead at least 6 m long is cast off falsework on each pier. A traveller
is then lifted onto the hammerhead and builds a first deck segment. This traveller is then
launched forwards, clearing the hammerhead for the second traveller that will build the
balancing half-span (Fig.1.29). The pair of travellers then proceeds to build the two adjacent
half-spans in balanced pairs of segments. As each pair of segments is completed, prestressing
cables are threaded and stressed. When the balanced cantilevers are complete, a mid-span stitch,
28
usually 2–3 m long, is cast, using one of the travellers as falsework. Continuity cables are then
threaded through the stitch and stressed.
Figure 1.30. Span division for hammerheads (top) and hammerhead division for segments
(bottom)
Each deck segment is normally between 3,5 m and 5 m long (Fig.1.30). It is traditional
to cast a pair of segments each week. A reasonable approximation to a construction programme
is to assume that the hammerhead takes 6–8 weeks to build, the first pair of segments four
weeks, the second pair three weeks, the third and fourth pair two weeks, and one week per pair
thereafter. A weekly rhythm has the advantage that concrete curing may take place over a
weekend. However, this rate of construction is not the best that can be achieved. The economy
of this method of construction is dependent on the use of the minimum number of travellers to
complete a bridge within a given programme. In some circumstances, speeding up the rhythm
to a pair of segments every 4 or 5 days may allow the use of one less pair of travellers (Benaim,
2007).
Figure 1.31. Typical prestress scheme for cast-in-situ balanced cantilevering
Construction may be accelerated by prefabricating the reinforcement cage in whole or
in part. The prestress scheme is usually arranged so that cables are stressed for each pair of
29
segments cast (Fig.1.31). When respecting a weekly rhythm the cables must be stressed as soon
as the concrete has achieved a cube strength of 25 MPa. This is less than the design strength
required for standard anchors, but can be achieved by using oversized bearing plates.
The complete cantilever is likely to be deflected downwards, as the self-weight
downwards deflection is usually greater than the upwards prestress deflection, and always so
for spans in excess of 70 m. To obtain a correct alignment pre-cambers are built in during
construction. The pre-camber calculations are complicated by the variation of the Young’s
modulus of the concrete with age (creep and shrinkage). In reality, this causes a few problems
for spans shorter than about 100 m, but requires a sophisticated calculation for longer bridges
(Hewson, 2003).
Precast segmental balanced cantilever construction
The most widely used method of erection of precast segmental bridges is balanced
cantilever (Fig.1.32). It is adaptable to spans from 30 m up to about 150 m, and can cope with
virtually any succession of span lengths and deck alignments. The upper limit on span is
generally imposed by the weight of the deeper segments and the cost of the casting cells,
although if there is enough repetition, longer spans are viable. A typical deck consists of pier
segments, and a number of span segments that are usually symmetrically placed about each
pier, in balanced cantilever. The span is closed by a mid-span stitch, cast in-situ. The joints are
usually glued, although this is only essential when internal tendons are adopted. Multiple
dovetail is created in each segment to enhance shear resistance of the joint section (Fig.1.33).
One significant difference from cast-in-situ construction is that the segments are several weeks
old when they are erected, reducing the changes in bending moment due to creep. Whereas
typically the creep coefficient for the concrete of a cast-in-situ deck is between 2 and 3, for a
precast deck it is of the order of unity. Except for very long spans, the effects of creep on the
erection geometry are reduced to virtual insignificance by the speed of erection for precast
decks: six segments in a day is usual (ASBI, 2008).
Figure 1.32. Precast segmental balanced cantilever construction: assembling by segment
erector (left, source: bbesols.com, 12.12.2015) and by mobile cranes (right, source: djc.com,
12.12.2015)
30
Figure 1.33. Glued and prestressed joints with multiple dovetail between segments
Precast balanced cantilever bridges may be erected with a crane, with shear legs (or
beam and winch), or with an overhead gantry (Fig.1.34) and with special segment erector
(Fig.1.32, left). The choice of method depends on the scale of the bridge, on the weight of the
segments, on the height of the deck above the ground level and on the nature of the terrain
crossed. The erection and attachment of the pier segment, the stability of the balanced
cantilever, the erection of the end spans and the construction of the mid-span stitch are common
to cast-in-situ balanced cantilever method of erection.
The simplest method of erecting bridge decks in balanced cantilever is by using groundbased cranes (Fig.1.34, left). If the bridge is over navigable water, barge-mounted cranes may
be used. Crane erection is simple and safe, using equipment with well-established safety factors
and operating procedures rather than purpose-made falsework. Also, no temporary loads are
imposed on the deck during erection, simplifying the stability falsework. It offers the flexibility
of being able to work out of sequence in a long or complex viaduct. Cranes can erect decks very
quickly, four to six segments in a day being typical if the segments are assembled on temporary
bars without the delay of installing permanent prestress. The main limitation on crane erection
is clearly the suitability of the bridge site. On an urban site, the delivery of segments on the
ground and the operation of cranes may be too disruptive to traffic.
Figure 1.34. Erection of precast balanced cantilever bridges with a mobile crane (left), with an
overhead gantry (middle) or beam and winch (right)
31
Where it is possible to deliver segments on the ground to all points of a bridge, the
segments may be erected by a pair of shear legs (or beam and winch). These are non-slewing
derricks equipped with electric winches, resting on the deck (Fig.1.34, right). The shear legs
are generally counter-weighted for self-weight and stressed down to the deck when lifting a
segment. An access platform is carried beyond the segment being lifted, for the prestressing
gangs. Generally, shear legs erect only one segment per day at each end of the double cantilever
with permanent prestress installed in each pair of segments. After each pair of segments has
been erected, the shear legs are moved forwards.
The overhead gantries are employed to erect bridges on high piers, in cities or over
water (Fig.1.34, middle). Generally a gantry will first erect the pier segment, launch itself
forwards to rest on this segment, receive segments delivered along the deck of the erected
viaduct and erect them in balanced cantilever while stabilizing the deck. The size, complexity
and degree of mechanical sophistication of a gantry depend principally on the maximum span
of the deck, on the weight of the segments to be handled, on the radius of curvature to be
negotiated, on the rate of construction planned and on the combinations of these factors. In the
majority of projects, the gantry applies loads to the deck as it launches itself forwards. As
gantries weigh between 150 tons and 400 tons, the decks need to be designed specifically to
carry their weight during the launching cycle (fib, 2004). The principal advantages of gantry
erection are the minimal interference with activities on the ground on urban sites, the ability to
cross any terrain, and the rate of erection that is possible. The principal disadvantages are the
time required for the design, construction, erection, testing and commissioning, which is
generally of the order of 15 months (unless an existing gantry is available for hire), the high
first cost which requires a bridge of at least 20.000 m2 of deck area, and the essentially linear,
less flexible construction programme. There are very many individual designs of gantry
(Benaim, 2007).
The cantilever construction method offers many advantages. Firstly, the bridge decks
are mostly built without any contact with the ground, making it possible to build structures over
the rivers which are subject to severe flooding or above very deep and rugged valleys. This
method can also be used to erect structures with very different geometries. Thus, in elevation,
it is possible to design decks of a constant or variable depth. For the latter, parabolic, cubic or
linear variations are all possible. Finally, construction using elements of 3 to 4 m in length is
cost-effective in terms of the formwork tools required for the bridge deck, even if the spans are
few in number and of various lengths. In case of prefabricated segments, the small size of these
components also helps to limit the weight of the elements to be assembled, thus reducing the
cost of the installation equipment.
Balanced cantilever is one of the most popular concrete bridge construction methods
also in Poland (Wanecki, 2002). The proven and safe balanced cantilever method is often
appropriate and cost-effective for the construction of long span concrete bridges including
various landmark structures where height, topography or geotechnical conditions render the use
of conventional formwork uneconomical.
32
1.5. Incremental launching
The incremental launching method is particularly suited to the construction of
continuous post-tensioned multi-span bridges. It involves casting 15-30 m long segments of the
bridge superstructure in stationary formwork behind an abutment (casting yard) and pushing a
completed segment forward with hydraulic jacks or friction launching system along the bridge
axis (Fig.1.35). The bridge segments are constructed in lengths that are determined so that the
construction sequence can be approximately one week. The segments are cast contiguously and
then stressed together. Low friction bearings on the piers make it possible to slide the heavy
superstructure forward. To keep the bending moment low in the superstructure during
construction, a launching nose (avanbeck) is attached to the front of the bridge span. This
construction method is rather industrialised since the formwork can be reused and therefore less
labour cost is needed. Also temporary piers, bearings and the launching nose can, after a
finished job, be reused in a new projects later on. One major advantage of incremental launching
is that no scaffolding or falsework is used when the superstructure is constructed.
It is used almost exclusively to build box girder bridges and is best adapted to that deck
form. It is normally used for bridges with spans between 30 m and 60 m, with a plan area in
excess of 3.000 m2 and the total length more than 150 m (VSL, 1977). Longer spans may be
launched, but they require intermediate support towers to cut down the launching span. Even
bridges over 1.200 m in length have been built with this method. Rates of construction are
typically one 15–30 m segment per week. However, there are large variations in the methods
and rate of construction (Benaim, 2007).
Figure 1.35. Incremental launching method in two general steps
33
The major falsework required for the construction of a launched bridge span is made up
only of the casting yard and the launching nose (Fig.1.36). Also required are the launching
jacks, the launch bearings and the jacking arrangements at the head of each pier. The casting
yard is normally set back some 30 m behind the abutment, so that it is not affected by the
deflections of the deck as it spans from the abutment to the first pier. If it is impossible to extend
the construction site by this amount, one or two temporary support towers may be placed within
the first span to reduce the deck deflections. The construction yard can be covered in order to
enable more protected environment against different weather conditions. The production of the
bridge in the construction yard makes it easier to overview and control the production and leads
to higher quality. Separate working steps recur in continuous cycles. The whole construction of
the superstructure is located in the casting yard.
Figure 1.36. Casting yard (left) and launching nose (right)
Figure 1.37. Typical structure of the launching nose
34
The launching nose usually consists of two tapering stiffened plate girders, typically
65% of the length of the launching span, wind-braced and cross-braced together (Fig.1.37). The
girders are usually 1,5 m deep at the front, and of the same height as the deck at the rear. Due
to the torsional stiffness of the deck, the nose should be designed so that one girder may take
the entire reaction without permanent damage. When the nose of the launching girder
approaches a pier, it is likely to have a deflection of at least 100 mm and it is common practice
to equip it with landing jacks that lift it to the level of the slide bearings. The launching nose is
usually attached to the deck with short prestressing bars. The joint with the deck must resist
principally sagging bending moments and the shear force that arises just before the concrete
span lands on the pier.
Figure 1.38. Typical sliding bearing for incremental launching
The sliding bearings consist of concrete blocks covered with polished stainless steel
sheet (Fig.1.38). They have lead-in and exit tapers at each end. To allow the span to move over
them, pads consisting of two layers of rubber laminated with a central steel sheet, bonded to a
Teflon (PTFE) lower surface are fed between the span and the rear end of the bearings. The
pads are collected as they are expelled from the front of the bearing and re-fed in at the rear.
The bearings should be as short as possible, as small deck rotations have to be accommodated,
particularly at the front and rear of the launch. Making the bearings too wide, however, may
increase the bending moments on the webs.
The deck is usually moved forwards by one of two systems. In the first case, two hollow
ram jacks reacting against the abutment pull on cables that pass beneath the casting area and
are anchored to the newly cast deck. Alternatively, the thrust force is transferred by friction
using one or more pairs of “launchers” placed under the webs of the superstructure (Fig. 1.39).
In the simplest version, a launcher is composed of a vertical jack pushed along a lubricated
surface by a horizontal piston. The piston acts against a foundation block or an abutment that
supports the deck in the pauses between two subsequent launches. It is important that the
35
abutment can resist high horizontal forces from the hydraulic jacks during launching, especially
at the late stage of construction when the jack has to move a very long bridge forward. In the
most advanced version, a friction launcher is a monolithic device composed of a sledge
containing the vertical jacks. The sledge slides between longitudinal guides along a steel-Teflon
base under the thrust of one or two pistons pivoted to a rear steel block (Rosignoli, 2002).
Figure 1.39. Working cycle of a friction launcher
In a conventional launched bridge deck, where 15–30 m segments are being built, the
following programme should be achieved. The first segment to be built is non-typical, as it
includes the arrangements for attaching the launching nose. Building this segment is likely to
take about 6 weeks. The first “typical” segment may be expected to take 4 weeks, the second 3
weeks, the third and fourth 2 weeks each and from the fifth segment onwards the weekly cycle
should be achieved. The construction sequence in the casting yard is carried out according to
the following steps (for concrete box section): casting of the bottom slab, casting of the webs,
casting of the deck slab, further hardening of the concrete, tensioning of the tendons and
launching of the bridge segment.
In contrast to all other construction methods, a central prestress is required during the
construction stage in the incremental launching method (Fig.1.40). This is due to the alternating
bending moments during launching. Central prestressing cables are arranged so that the
resultant compressive stresses at all points of the cross-section are equal and therefore it makes
no difference whether the tensile stresses produced during launching occur in the upper or lower
parts of the section. This type of prestress is, of course, quite incorrect for the pattern of
moments in the completed state and moreover cannot be subsequently adapted to that pattern.
By adopting a relatively low span-depth ratio, however, it is possible to keep the central
prestressing low and economical (Rosignoli, 2002). The arrangement of the central prestressing
cables is so that, in conjunction with the reinforcement, they provide the necessary factor of
safety against rupture during construction. When the bridge superstructure has been completely
launched, the continuity tendons are pulled or pushed through and stressed. Their lay-out is
designed according to the bending moments in the completed state in which they supplement
36
the central prestressing, which, of course, remains active. In planning the stressing programme,
careful consideration is given to the changes in forces and stresses which will be produced.
Figure 1.40. Prestressing cables arrangement during incremental launching and in the final
stage: (1) external cables with stiffeners and (2) cables in the web
The main advantages for using this construction method, rather than other traditional
methods, are: minimal disturbance to environmentally sensitive areas, smaller assembly zone
required, greater safety during construction which is mainly carried out on the ground, economy
of transportation and general reduction in construction elements, higher quality finish and
performance derived from easier working conditions and repeatability of tasks and finally ease
of access to restricted or limited sites – such as over rivers, deep valleys, road or train lines, in
poor soil conditions or environmentally protected areas (Göhler & Pearson, 2000). Although its
significant advantages make using this technique a highly attractive option, certain aspects
require a high level of expertise – both in terms of people and equipment. The main
disadvantages of the method are the additional costs for launching jacks, launching nose,
additional prestressing, the construction yard and extra amount of concrete needed to increase
the cross-section due to extra stresses caused by the launching.
Due to the limited extent of the falsework, and to the repetitive nature of the process,
launching is frequently the most economical method of construction a prestressed concrete
bridge within the span lengths and overall length defined above and with a suitable alignment,
despite the greater depth of the deck and the greater weight of prestress. It may still be
37
competitive for longer spans, but the need for intermediate temporary towers compromises its
economy to some degree.
1.6. Case study No.1. Construction of a bridge on scaffolds
The bypass of Dobczyce is the transportation project bringing road and pedestrian traffic
over the floodplain of the Raba River, over 600 m in width. The total length of the bypass is
1.230 m. Its basic part is two bridges, 645 m in length, i.e. the 180-metre-long bridge and the
access viaduct , 465 m in length. The main bridge is a three-span continuous beam with the
following span lengths: 51,0+76,0+52,0 m (Fig.1.41) and total width of the deck of 15,9 m.
Figure 1.41. Side view of the bridge
Figure 1.42. Vertical cross-section of the bridge (left) along with superstructure details (right)
38
The superstructure, which is a beam-and-slab system, is made up of two prestressed
beams of the constant depth 1,72 m, placed with the centreline spacing of 10,2 m. The main
beams are braced with reinforced concrete (RC) crossbeams with the cross-section of 1,3 x 0,8
m, spaced along the superstructure every 25 m. The grillage is covered with a monolithic RC
deck slab, whose thickness varies from 0,25 to 0,36 m. The bridge spans are prestressed with
internal and external cables, placed at big eccentricity out of the beams' cross-section and
supported with two RC pylons 15,0 m in height and with the cross-section of 0,70 x 2,0 m,
monolithically connected with the spans. The pylons were located in the axis of the main beams.
The external cables are attached to the pylons in a fan layout (Fig.1.42).
The main bridge beams have the same depth of 1,72 m and variable width, from 1,50 m
in the side spans and in the central part of the middle span, to 2,0 m in pylons cross-section, i.e.
over the intermediate supports. In order to obtain the suitable carrying capacity (without
increasing the beams depth), suitable prestressing with big eccentricity was required, which was
possible due to the external location of the main prestressing cables and supporting them with
the pylons. The concrete structure was prestressed with 12 internal multistrand tendons 19x15,7
made of Y1860 steel and external multistrand tendons 12x15,7 made of Y1860 steel with highdensity polyethylene (HDPE) sheathing. The external cables were anchored at the pylons with
passive anchorages. The active anchorages are located in the main beams beneath the deck. The
anchorages enable the access to the bearing plate (for a monostrand jack) and have regulation
nuts, which, in turn, enables the regulation of the cable length with no interference with the
anchorage.
Due to the big spacing between the main beams, the RC crossbeams were designed,
spaced at 25,0 m and as deep as the main beams. The crossbeams above the abutments and the
span crossbeams are 1,40 m wide, while the crossbeams above the intermediate (pylon)
supports are 2,80 m wide. The deck slab thickness varies from 0,25 m to 0,36 m and its width
varies from 14,1 to 15,7 m. The geometry of the deck slab fits the bypass gradeline, regarding
vertical arches and transition curves. The deck slab was installed with the 2% transverse
gradient at the road width and the 3% transverse gradient at the pavements. The whole
superstructure was made of C40/50 concrete.
Figure 1.43. Temporary supports of the falsework (left) and assembling of falsework (right)
39
Concreting of the bridge spans was performed with the help of steel falsework and
modular formwork. The main elements of falsework were columns made from steel pipes, caps
and girders made from I-profiles (Fig.1.41). Altogether ten supports along the bridge were
performed to carry all the formwork for superstructure. The elements of the formwork system
were supported on the falsework. The formwork included precast frames of soft reinforcement,
steel ducts for prestressing tendons and sheathing for external tendons anchorages (Fig.1.44).
The pylons were cast in segments with traditional scaffolding supported on the completed
segment of the main span.
Figure 1.44. Formwork (left) and reinforcement (middle) of the superstructure and the pylon
scaffolding (right)
Concreting the spans was performed in 3 steps. The first step consisted in performing
the whole river span (76,0 m) along with the segments of the side spans adjacent to the supports,
approximately 10 m in length. This stage also included concreting the pylons. During the second
and the third stages the other segments of the side spans were made (approximately 41 m).
Concreting operations were carried out continuously, covering the whole cross-section and
making several technological intervals (due to concrete shrinking).
The system of prestressing and tensioning of the individual groups of cables and tendons
was performed with regard to the stages of superstructure erection. The main phases were as
follow: concreting the spans and the pylons, prestressing of the spans (main tendons), assembly
and preliminary tensioning of the external cables (phase 1), placing the bridge equipment,
tensioning of the external cables (phase 2), prestressing of the spans (section tendons). The
two-phase prestressing operation for the superstructure and the two-phase tensioning operation
for external cables were envisaged. The effects of the operations in phase 1, i.e. prestressing of
the main tendons and preliminary tensioning of the external cables, were determined in the plan
of tensioning. Phase 2 was planned to be carried out after equipping the bridge, which resulted
from the regulatory character of the operations within that phase.
40
Figure 1.45. External cable numbers and the order of tensioning (details in the text)
Tensioning of the external cables was carried out by symmetrical simultaneous
prestressing of four cables mounted at one height of the pylon. That way to carry out the
operation enabled minimizing bending of the pylon and even tensioning of the external cables.
Two methods of tensioning of a single cable were applied: tensioning of the whole cable with
a large hydraulic jack (short cables) and strand-by-strand tensioning which applies the
technique balancing forces in strands (other cables). First to be tensioned were cable pairs W8W9 to W4-W13, next W24-W25 to W17-W32, and finally W3-W14 to W1-W16 (Fig.1.45).
Tensioning of the cables in phase 2 was performed analogically.
Figure 1.46. Falsework with modular formwork for approaching viaduct construction (left)
and the viaduct from beneath (right)
Simultaneously with the main bridge construction the approaching viaduct was built.
The viaduct is a 16-span concrete continuous beam structure, made up of two RC post-tensioned
beams of uniform depth, cast monolithically with crossbeams and deck slab. The viaduct spans
have the lengths of 21,5+14 x 30,0 + 21,5 m, and the total length of the approaching viaduct is
465,0 m. (Fig.1.46). In plan, the viaduct is a straight line going into a horizontal curve with the
radius R = 400 m, and then into another straight line. The cross-section and the usable widths
of the viaduct are the same as the ones of the main bridge. The superstructure of the viaduct
was performed in the technology of construction on falsework and formwork for several two41
span sections approximately 60 m in length each. Technological concreting brakes were located
at the distance 0,2L from the support. Individual sections were post-tensioned after being
manufactured with internal tendons, extended for the following section with the use of system
couplers. In the individual sections, tendons were anchoraged in specially shaped external
anchoring blocks. After prestressing one section of the viaduct, the falsework and the formwork
were moved to another section.
1.7. Case study No.2. Construction of a concrete bridge
using balanced cantilevering
The motorway bridge MA 78 over Dunajec River near Tarnow with an overall length
of 600 m and twelve pairs of piers supports the two carriageways13 m wide each (Fig.1.47).
The sections between the seventh and tenth piers with the total length of 220 m were built with
balanced cantilever method, whereas in two side sections: left one 286 m in length and right
one with 94 m in length, the bridge was constructed with the help of falsework. The main span
built by balanced cantilevering has the length of 110 m. The single­cell hollow box
cross­section of the superstructure varies in depth from around 2,50 m in the middle up to
6,10 m at supports although the width remains constant throughout (Fig.1.48).
A total of four cantilevered construction units were used for the building of the bridge
superstructure. The construction of the superstructure moved forward symmetrically in both
directions from the starting point of two piers in each case. Totally 4 x 11 sections with the
lengths of 3,5 m to 5,0 m were concreted alternatively until the gap was finally closed. For
balanced cantilever construction of the bridge the new PERI’s bridge cantilevered construction
(BCC) equipment with a high degree of flexibility for the forming of the bridge cross­sections
was used (peri.com).
Figure 1.47. The motorway bridge MA 78 over Dunajec River in Tarnow: under construction
(left, source: wydawnictwo.inzynieria.com, 12.12.2015) and after completion (right)
42
Figure 1.48. The single­cell hollow box cross­section of the superstructure
(source: peri.com, 12.12.2015)
The new system components for the longitudinal direction of the cantilevering
construction were optimized for 5­metre long concreting sections. This length corresponded to
current project requirements. The compatibility achieved with the PERI UP modular scaffold
ensured the provision of safe working areas with a minimum of effort for the construction team
as the scaffolding can be connected to the BCC components by means of simple connection
parts. For construction of the road bridge near Tarnow, a total of four cantilevered construction
units were used based on the BCC engineering construction equipment. The concreting section
lengths range from 3,5 m to 5,0 m and were completed in four to five working days. The
haunched cross­section of the bridge featured a variable web height and required constant
adjustment of the formwork for the hollow­box sections. Girder wall formwork elements
provided sufficient flexibility for accurate adjustment of the superstructure formwork. By
means of a formable steel plate connection between the girder formwork elements, the solution
had the required flexibility (Fig.1.49).
Figure 1.49. Two cantilevered construction units used for each bridge erection
(source: peri.com, 12.12.2015)
43
The construction using standard components from the BCC engineering construction kit
could be easily adapted in a transverse direction – without requiring any steelwork. This
solution allows the use of all components for future applications – even for tunnel and civil
engineering projects. For connecting the PERI UP modular scaffold to the BCC solution,
corresponding connection elements were available. This ensured that safe working platforms
and access points were available for all work assignments (Fig.1.50).
Figure 1.50. Adaptation of construction kit in a transverse direction (left) and PERI UP
modular scaffold connected the to the BCC equipment (right) (source: peri.com, 12.12.2015)
Figure 1.51. The PERI’s typical falsework for side spans (source: peri.com, 12.12.2015)
Due to variation in height while the constant width, a continuous change in the angle
between the outer side of the web and the cantilevers had to be made, therefore the formwork
had to be re-adjusted for each concreting section. Regarding this requirement for bridge
constructions, PERI has developed a simple and practical solution: a formable steel sheet
connects the two girder formwork elements with the result that the angle between the web and
44
carriageway formwork can be constantly changed by means of spindles. Following the change
in the cross-section, the slab formwork of the web was fixed at its respective height whilst the
element of external web formwork remained unchanged. With this construction, the required
low dimensional tolerances for the bridge cross-section could be met without any problems and
each individual segment could be realized on a customized basis.
Furthermore, the crane-independent moving procedure was also optimized in terms of
workload: after only a few simple steps, the frame could be moved forward to the next section
through the use of hydraulic presses. For this process, the construction team needed only
2 hours on site. After the respective gap closure between two piers, the cantilevered
construction units were moved back to the starting point on the pier head where they were
dismantled.
In the side spans, the bridge is constructed with the help of falsework. PERI’s ST 100
stacking towers and MULTIPROP props – connected with frames to form shoring towers –
carry the formwork in these sections (Figure 1.51).
The construction of the concrete bridge MA 78 is a good example where the formwork
& scaffolding supplier took on the responsibility for the customized planning and static
calculations of the complete solutions for the falsework ­ supported superstructure as well as
the cantilevered construction. As a result, all systems and processes were therefore optimally
matched. The solution also included technical documentation along with providing constant
back-office support for the construction team throughout the project. In addition, the on-site
presence of the supplier supervisor accelerated building progress.
45
2. Steel bridges construction
The choice of a construction method for a steel bridge is influenced mainly by the bridge
location. Regarding beam bridges, the principal methods of construction are: construction with
a crane from the ground, construction by launching, cantilever erection, placement of the
complete bridge (or of large bridge elements) with heavy lifting methods or with a barge. These
methods are described in detail in this chapter. The technical difficulties associated with each
method are highlighted. This is followed by the description of the fundamental characteristics
of steelwork construction that are common to all methods and must be carefully considered
while studying the issue. Other construction methods may also be considered on a case by case
basis. Certain methods have been developed to respond to the specific needs and characteristics
of some types of steel bridges. For example, placement of bridge elements using a trolley
suspended from a cableway above the bridge is of interest for suspension bridges, because the
pylons of the final structure can be used to support the cableway. Specific techniques have also
been developed for arch and cable-stayed bridges (e.g launching by rotation).
The erection of bridge steelwork is a particularly critical phase as far as the overall
stability of the structure and the safety of personnel are concerned. A detailed study of the
impacts of the construction method is needed. Due to the increasing capacity of lifting
equipment, the current trend is to avoid the falsework which is generally expensive, while
increasing the size of the steelwork pieces and making the maximum use of those parts of the
superstructure already in place to support those that follow. One consequence is that during
construction the parts of the steel structure may be subject to stresses from their own weight
that are higher than those they will experience in service. This effect is more pronounced if the
structural form of the load carrying structure during erection is different from the one in the
final state. Such is particularly the case for bridges that are launched or erected by the cantilever
method. During construction various actions tend to occur together and their numerical value
assumed in design is more easily exceeded. A detailed study of the construction load
combinations, as well as diligent control of each construction phase, are required because the
probability of an unforeseen or accidental action is much higher than when the bridge is in
service. All these reasons mean that the potential impacts of the construction method on the
load carrying structure should be considered in the earliest design studies (Lebet & Hirt, 2013).
It is likely that the elements needed to guarantee the stability of certain parts of the
structure in its final state will not yet be in position during the construction process. This
situation must be addressed either by planning temporary support elements, or by ensuring
through other means that structural safety in terms of stability is guaranteed throughout the
46
construction process. An example is a concrete slab of a composite bridge which provides
lateral support to the main beams compression flanges, but which is only in place and effective
after erection of the steelwork and construction of the slab. It is also necessary to check elements
of the steel structure for their resistance to local buckling when subjected to concentrated forces.
The overall stability of the superstructure during its placement (against overturning, uplift, etc.)
must be guaranteed. There may be a lack of stability during erection due to, for example, the
absence of some of the self-weight that has a stabilising effect. It is also important to check that
the various pieces of steelwork can adequately resist the various handling operations to which
they will be subjected (lifting, transportation, assembly) (Lebet & Hirt, 2013).
To ensure that a steel bridge can be safely, economically and reliably executed
(fabricated, assembled and erected), designers should be aware of the processes of fabrication
and erection, the capabilities and limitations of the steelwork contractor and how the design
choices affect those processes. Guidance on design for construction generally follows the
sequence of activities undertaken by the steelwork contractor. The objectives of “design for
construction” are: to maximize the efficiency of the construction process and to minimize the
need for clarification and change. Achieving these objectives will reduce costs, reduce the
construction period, enhance quality and increase the safety of the work. The activities for the
steelwork contractor include planning, ordering, modelling, fabricating, assembling, coating
and erecting (for more details see: SteelConstruction.info).
How the bridge is designed and how it is built are linked. Ensuring that this link is
recognised will have an effect on the safety and quality of the construction. Methods of erecting
steel bridge structures vary considerably from site to site and from project to project. The
subject of erection aspects is therefore a wide and varied subject. This chapter focuses on the
aspects that should be considered during the initial design and detailed design phases.
2.1. Construction with a crane from the ground
Construction with a crane means lifting various steelwork elements - lengths of main
beam, cross bracing, horizontal bracing - from the ground using a crane (Fig.2.1). This is an
advantageous means of erecting the steel structure because it requires less equipment on site
and a reduced amount of labour. Erection using cranes is considered the most cost effective
erection method for the majority of structures. However, it does require good accessibility
around the construction site, and is only suitable for bridges which are relatively not very high
(with a deck up to approximately 15 m above the ground). Only crawler cranes and some small
rough terrain mobile cranes are able to traverse the site with a load. The site areas adjacent to
the bridge will affect the position and size of crane that can be used and this will affect capacity
of lift and therefore choice of crane and the piece size. The area for preassembly will also
influence lift size.
The erection of an I-girder is often accomplished with a single construction crane
(Fig.2.2, left). Depending on the size, weight and capacity of the girder, the crane could use
either a single lift point located at the centre of gravity of the member or two lift points located
47
symmetrically towards the centre of gravity utilizing a spreader beam. Depending on the overall
size and complexity associated with the bridge construction, the individual girder erection could
incorporate the use of temporary supports, temporary pier brackets or hold cranes to provide
stability to the member until a sufficient number of girder lines are erected and connected via
permanent cross frames or diaphragms to form a stable system. The erection of a box girder is
often carried out with a pair of construction cranes working in tandem during the lift (Fig.2.2,
right). The cranes are typically rigged with a spreader beam that contains two beam clamps
each. The beam clamps are attached to the top flanges of the box girder near the ends of the
member. Unlike I-girders, an individual box girder may be stable once it is lifted onto its
permanent supports, due to the internal bracing system.
Figure 2.1. Construction with a crane from the ground
Figure 2.2. Erection of I-girders with a single crane (left) and erection of a box girder with a
pair of cranes working in tandem (right)
The erection may take place either with or without temporary supports. If such supports
(props, falsework) are used, they may be either continuous or discrete. The evolution of heavy
lifting equipment means that when temporary supports are needed, they tend to be a small
number of discrete props rather than continuous falsework. The bending moments that can be
resisted at the bases of the telescopic arms of mobile cranes currently used worldwide can be
as high as 15.000 kNm, which means that, e.g. a weight of 100 tones can be carried at a distance
48
of 15 m. If the spans are sufficiently short, and the size of the individual steelwork pieces allows,
the temporary supports can be avoided. An alternative method may be used if the superstructure
of the bridge is sufficiently strong and stiff in bending, or perhaps it can be supported with
temporary supports. To begin with, the structure is constructed either from the ground or
launched. Once the structure is in place, it serves as a working platform for constructing the
spans. Falsework towers are supported on the deck and allow relatively easy erection of e.g. the
arches. Such a procedure is shown in Figure 2.3.
Figure 2.3. Falsework towers supported on the deck for arch erection: scheme (top) and two
examples (bottom)
2.2. Incremental launching
This method of a steel bridge construction is about assembling the elements of a
structure in an area that is in line with the bridge axis, and located at one or both ends. As the
elements are joined together piece by piece, the structure is pulled or pushed, in stages, into its
final position (Fig.2.4). Launching is typically performed in a series of increments so that
additional sections can be added to the rear of the superstructure unit prior to subsequent
49
launches. The superstructure is completely assembled (typically one span or more at a time),
including cross frames and bracing, prior to launching operations.
Launching the steelwork over road or river, involves assembly of steelwork, typically
behind an abutment, on the highway approach. The steelwork is assembled on low resistance
rollers or a sliding system at each pier or temporary support. A haulage and restraint system is
used. The steelwork requires plenty of construction alignment to ensure adequate fit up prior to
launching and needs to have sufficient time in the programme to allow for this. However,
assembly is normally near ground level, with the use of much smaller and less expensive cranes
and minimum work at height.
The method of launching the steelwork is adopted when lifting equipment cannot access
the whole length of the bridge, or when the height of the bridge becomes too big for lifting with
cranes located on the ground. The method can be used to construct a bridge over a wide range
of challenging sites, which feature limited or restricted access, including those with the
following characteristics: deep valleys, deep water crossings, steep slopes or poor soil
conditions making equipment access difficult or environmentally protected species or cultural
resources beneath the bridge. Currently this method is used to launch girder bridges with spans
up to 150 m (Rosignoli, 2002). The launching method has also been applied to tied-arch or truss
spans, although these are fully assembled prior to launching.
Figure 2.4. Incremental launching scheme
To accomplish a girder launch, the superstructure is assembled on a roller system behind
one of the abutments in segments of sufficient lengths to maintain stability while launched.
Once a segment is assembled, hydraulic jacks are used to move the structure across the
abutment until it reaches the first pier support. After the girders are properly aligned on the first
pier, additional segments are assembled on the roller system behind the abutment and the
hydraulic jacks advance the structure until it reaches the next pier. This launching process is
continued until the entire structure rests on its permanent supports. This procedure requires a
little equipment, but the girders must be specifically designed both globally and locally to
handle the large forces generated during the launch.
In order to reduce the cantilever moments and the amount of deflection that occurs
during launching operations, one of two systems (and sometimes both) may typically be
employed. On the one hand, the contractor can construct a tapered launching nose on the leading
end of the girders (Fig.2.5, left). The launching nose reduces the dead load of the cantilever
span and utilizes its tapered profile to assist in “lifting” the mass of the girders as they are
launched forward onto the landing pier. In other cases, the contractor may choose to use a
50
relieving system utilizing temporary cable stays to reduce the deflection of the leading end of
the girders during launching (Fig.2.5, right).
The incremental launching method offers a number of significant advantages to both the
owner and the contractor, including the following: minimal disturbance of environment,
including environmentally sensitive areas, smaller but more concentrated area required for
superstructure assembly and increased worker safety since all work is performed at a lower
elevation. Launching has the advantage of allowing all the steelwork elements to be assembled
on the ground in the assembly area, which means that work can be carried out in more suitable
conditions than may be the case with alternative construction methods. This is particularly
important regarding welding operations (Rosignoli, 2002, Gohler & Pearson, 2000).
Figure 2.5. Two systems to reduce cantilever moments: launching nose (left) and temporary
cable stays (right)
Figure 2.6. Temporary devices to compensate for depth variations during launching
The method of launching has some limitations, for instance: sufficient space is available
behind an abutment and in line with the bridge axis for steelwork assembly, the bridge must be
either straight or curved in plan with a constant radius if it is to be launched from a single
51
abutment, the bridge may be straight and curved if launched from both abutments, but the
transition zone between straight and curved lengths must not be too abrupt, preferably, the main
beams should be of constant depth, because the lower surface of the steel structure should be
planar (the rolling plane). However, many structures of variable depth have been launched in
this way, using temporary devices to compensate for depth variations during launching, for long
span bridges it is preferable to have a closed cross section (box or U-shaped section closed by
temporary plan bracing) to ensure sufficient bending and torsional stiffness in the horizontal
plane (Fig.2.6).
2.3. Cantilever erection
Cantilever erection means forming the main beams by constructing cantilevers that
extend away from the piers by joining together successive pieces. Joining the two opposing
cantilevers at mid-span achieves the continuous beam. This method of erection is particularly
suitable in case of long-span bridges (>100 m), and those high above the ground or water level.
It is often used to construct bridges over navigable waterways because the bridge elements can
be brought in a barge and then lifted onto their position. The method is well adapted to cope
with all types of bridge alignment, and bridges formed from beams with significant variations
in depth.
Figure 2.7. Cantilever erection of steel girder bridges
52
Cantilever erection can also be considered for construction in a single direction starting
from an abutment. For example, the first span of a bridge, lifted into place from the ground
using a crane and a temporary support, can serve as a counterweight to the next span
cantilevering out (Fig. 2.7, top). If a span is too big, then temporary supports can be applied.
Symmetric cantilever erection away from piers (Fig.2.7, bottom) requires the steel structure to
be rigidly connected to the piers. The pieces of cantilever are lifted into place, either with a
crane located on the ground or with lifting equipment located on the structure already built. The
beneficiary of cantilever method is that this method does not require special equipment in
comparison with launching method which requires more sophisticated equipment such as huge
electric winch or hydraulic jack, wire rope, etc. Therefore, the cantilever method is less
expensive, less on risk, and faster (Lebet & Hirt, 2013).
The cantilever method of erection for continuous trusses starts with assembling the
sections over an interior pier using temporary bents or pier brackets to stabilise the panels
(Fig.2.8, left). Once the initial pier panels are erected, new panels are added to each end in an
alternating fashion until mid-span or the abutment are reached. The same procedure is repeated
again until the ends of the two cantilevers are ready for the installation of the closure pieces.
Assembling in one direction or from each side can be performed (Fig.2.8, right). Erecting a
truss in this fashion can be performed using standard construction equipment or traveling
derricks. In addition, erection can proceed simultaneously from each interior pier, and truss
geometry and deflection can be controlled using hydraulic jacks at the temporary supports.
Another commonly used method is to assemble truss section components off site, thus,
providing erection of assemblies while filling in smaller members piece by piece with smaller
equipment.
Figure 2.8. Cantilever erection for continuous trusses: start with assembling the sections over
an intermediate pier (left), assembly in one direction or from each side (right)
The construction of arch bridges with cable supports with cantilever method consists in
erecting a half-arch, which is cantilevered from the bank with temporary support from cable
stays, as shown schematically in Figure 2.9 (a). This method is only of interest for arches with
an upper deck, because the cable stays can then be supported from temporary pylons that need
53
only be of limited height. In favourable conditions, it is also possible to lift elements of the arch
into place from the ground, using a winch or a crane. For bridges of large span, or when the
access is very difficult, an aerial ropeway (Blondin) may be used (Fig.2.9b). Also for long-span
bridges, in order to reduce the time needed on site for erection, the central part of the span may
be assembled on the ground then lifted into place using large jacks. This method, which is
shown in Figure 2.9 (c), often requires the use of a temporary tie across the central part of the
arch to limit bending during the lifting operation.
Figure 2.9. Construction of arch bridges with cantilever method
When constructing an arch bridge with cantilever method, a temporary stay towers are
typically applied, located at the spring line (an end pier for tied arches) with stay cables
connecting the partially completed arch to the tower on the leading end and the tower to
counterweights or ground anchors on the trailing end (Fig.2.10). The erection begins at each
spring line and progresses towards the crown. As arch pieces are erected, additional stay cables
are added to provide temporary support until the closure piece is in place and the arches are
self-supporting. Arch pieces can be erected into place with cable and winch systems attached
to the leading end of the previous segment or with cranes located beneath the bridge, if feasible.
Using temporary stay towers is the most common erection method since arch bridges are
predominantly used to span deep canyons (true arch) or wide bodies of water (tied arch).
54
Figure 2.10. Cantilever construction of arch bridges with temporary stay towers
(source: en.ccccltd.cn, 12.12.2015)
Another method of true arch erection uses the canyon walls to support the arch as a
cantilever (Fig.2.11). With this method the arch is erected from each spring line simultaneously
towards the crown. As each arch member is erected it is "tied" back into the canyon wall using
a temporary connection. The tie member can be part of the permanent structure (as shown
below) or a temporary member/cable used solely for erection support. Once the arch is closed
and erection complete, the temporary tie members/connections are removed and the arch
becomes self-supporting.
Figure 2.11. Cantilever method of true arch erection with canyon walls support
(source: sunsteelandspray.com, 12.12.2015)
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Cable-stayed bridge erection is performed utilizing a permanent tower to support all
loads during the assembly of the superstructure. After the majority of the tower construction is
completed, the first superstructure segments are erected directly at the tower using temporary
supports attached to the tower’s cross strut. After the first segment is stabilized, additional
members are cantilevered from each end in an alternating fashion until the location of the first
permanent cable stay is passed. At this point, the first set of stay cables is installed. Alternating
cantilever erection of segments continues along with stay cable installation until the end pier or
mid-span are reached. This procedure is repeated for each tower of the bridge. When two
adjacent cantilever ends are completed, the closure pieces are installed and the erection is
finished. This erection method is common, since all of the erection forces are resisted by
permanent bridge components. Only simple beam and winch erection equipment attached to
the cantilever ends or deck mounted derricks are necessary to lift the members into place
(Fig.2.12).
Figure 2.12. Cable-stayed bridge cantilever erection with beam and winch equipment attached
to the cantilever end: general view (left) and lifting the segment (right)
2.4. Heavy lifting
Heavy lifting (or strand jacking) is not commonly used as an erection technique as it is
best suited to large lifts that cannot be carried out by crane. The temporary works for strand
jacking are usually complex and expensive. The division of the structure into an erectable
portion has to be influenced by the capabilities of the equipment. Heavy lifting (HL) is a
hydraulic lifting technique especially developed for extremely heavy loads and usually applied
for lifting of precast segmental superstructure members. The technique provides a particularly
timely and economic solution for projects based on modular construction methods and large,
heavy, pre-fabricated elements.
HL is perhaps most commonly used for lifting long midspan river sections from a barge
Fig.2.13, left). However, it is suited to other cases such as the central section of the arches,
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which are usually lifted onto the deck and transported into position beneath the jacking towers
(Fig.2.13, right).
Figure 2.13. Heavy lifting: midspan river sections from a barge (left) and central sections of
the arch (right)
Heavy lifting systems consist of proven equipment for the safe handling of heavy loads,
i.e. parts or the whole steel spans. The HL system is designed for lifting or lowering suspended
loads and for pulling or sliding loads. The main components of such a system are: a motive
unit, a tensile member with its anchorage for the load, a pump and its controls (Fig.2.14). The
motive unit anchorages have a self-gripping feature which maximises the inherent safety.
Figure 2.14. Main components of HL system: scheme (left) and view (right)
(source: vsl.net, 12.12.2015)
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The motive unit is made up of a hydraulic centrehole jack and upper and lower
anchorages. The upper anchorage is attached to the jack piston. The jack is extended during
lifting, which makes the individual strands of the tensile member be gripped by the upper
anchorage and thus moved upwards. At the start of the downward movement of the piston, the
strands are gripped by the lower anchorage as the upper anchorage opens. The load is therefore
moved in a step-by-step process. For lowering operations, the motive units are equipped with a
device that automatically controls opening and closing of the anchorages. Lifting and lowering
operations can therefore both be carried out with the same degree of safety and speed.
The process of a heavy lift is really well-organized, with all steps and details planned in
advance (Fig.2.15). The lift begins with the lifting unit in an extended position (1), the upper
anchorage engaged (2) and the lower anchorage disengaged. At the upper end of the stroke, the
piston is retracted and the load is transferred to the lower anchorage while the upper anchorage
is disengaged (3). The piston is then lifted for the next lifting step (4). All steps of the lift are
monitored, along with the movement during each lift. All jack and pump controls for the heavy
lift are connected to a central command unit for synchronisation, which allows remote
controlled operation of all lifting jacks. This ensures that the resulting lifting rate is the same
on all jacks and also provides a level control system.
Figure 2.15. The process of a heavy lift (source: vsl.net, 12.12.2015)
2.5. Placement of the complete span or large bridge
elements
It is possible to manoeuvre the complete steel structure into place or, for large bridges,
to erect sub-structures weighing hundreds of tonnes. The most commonly used methods are
moving the bridge into place with the aid of barges, by transverse sliding, or placement by
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rotation. When the bridge crosses a navigable stretch of water, the steel structure can be placed
on barges for placement. For example, if the bridge is constructed on a bank that is in line with
the axis of its final placement, one end of the bridge can be placed on a barge that serves as a
float (Fig.2.16). The bridge is then moved into position by pushing the barge towards the
abutment located on the opposite bank.
Figure 2.16. Moving the bridge into place with the aid of barges
If the bridge is assembled in an area further away from its final position, then the
complete structure can be placed on barges to be transported from the assembly area to its final
location (Fig.2.17, left). A box girder bridge can be floated into position by towing, once it has
been made watertight (Fig.2.17, right). The box is lifted using cables linked to jacks fixed to
the piers and abutments, then placed in the correct position.
Figure 2.17. The bridge span floated into position on barges (left) and part of the span floated
by towing (right)
Tied arch bridges lend themselves to placement of the complete bridge because their
support reactions are uniquely vertical. A temporary tie must be used during erection to stabilize
arches that will not be tied in their final state. Preassembly is only possible when there is
sufficient space to construct the arches on the banks or along the axis of the bridge. In urban
environments, therefore, preassembly of the complete bridge is often difficult. Two ways of
placing a complete arch bridge may be envisaged: by launching or by rotation (Lebet & Hirt,
2013).
The complete bridge span may be launched when it crosses a navigable waterway and
there is sufficient space for preassembly on one of the banks in line with the bridge axis. The
leading edge of the bridge is landed on a barge, which can then cross the waterway (Fig.2.18).
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Because the supports and structural form of an arch bridge vary according to the launching
phase and are different from those associated with the bridge in its final position, it is normally
necessary to provide the arch with temporary supports as the hangers are designed for tension
forces only.
Figure 2.18. Placing a complete arch bridge by launching
The complete tied arch bridge may be also manoeuvred into its final position by rotation
(see p.2.7). The bridge is erected on one of the banks, parallel with the bank edge. Once the
deck and arches are finished, the bridge is constructed parallel to the river and rotated into its
final position about a vertical axis through the abutment. This method is considered economical
because the work associated with erection of the arch is carried out on the ground, giving easier
access to the whole of the assembly area, with no need to use any special equipment. The same
rules for bridge launching or rotation as for arches can be applied to assembly of complete steel
truss spans or cable-stayed footbridges (see p.6.3).
It is also possible to slide the complete steel structure into position transversally
(Fig.2.19). This method is often used when a new bridge replaces an existing structure. The
new bridge can be assembled with minimal disturbance to the traffic on the existing structure.
The new bridge is erected alongside the existing structure on temporary piers and abutments.
Traffic is then diverted onto the new structure, so that the old structure can be deconstructed.
Total interruption to the traffic is brief; generally the transfer takes place over one night, which
is sufficient to allow the new structure to be slid onto the refurbished piers and abutments. Then
the new structure can be opened to traffic and the temporary piers and abutments are removed.
The most recent Polish example of this technology has been described by Palmowski &
Krzysztoforski, 2013.
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Figure 2.19. Launching the complete steel structure into position transversally – four main
stages (courtesy: PRM Mosty-Łódź)
2.6. Case study No.1. Erection of the steel superstructure
by crane
The footbridge in Rzeszow was built in the form of a ring with the outer diameter of
about 40 m and the usable width of 4 m. The ring is supported with four pillars dividing it into
four spans with the lengths of 32,6 + 29,0 + 29,7 + 31,8 m along the outer circumference
(Fig.2.20). The various span lengths result from the necessity to avoid collision with existing
streets, pavements and underground installations. The approaching ramps and staircases have
various span lengths and the usable width of 2,5 m (Siwowski & Wysocki, 2015a).
The superstructure is a steel box girder with the depth of 0,65 m and the width of 4,0 m.
The box cross section is divided with 10 mm inner webs into four chambers and laterally
stiffened with crossbeams (with 1,0 m spacing along the circle) and diaphragms (with 3,0 m
spacing). The 12 mm upper deck plate is stiffened with the longitudinal (circular) I-ribs with
the lateral spacing of 0,25 m. The bottom flange of the box girder is made of 12 mm plate
without longitudinal ribs (Fig.2.21).
The footbridge construction started with drilling foundation piles on which the concrete
footings were cast to support the concrete filled steel tube (CFST) pillars. The steel ring
superstructure was divided into 13 assembly parts to facilitate transportation, i.e.: 4 support
parts (P1-P4) and 4 span parts, additionally divided into longer segments (S1-S4) and shorter
segments (PP1-PP5). The biggest assembly part weighed about 23 tones and had the length of
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about 18 m along the outer circumference. On the construction site all parts were welded into
8 main elements as it was necessary to locate temporary supports beyond the roads so the traffic
was not interrupted during construction) (Fig.2.22).
Figure 2.20. Plan view of the footbridge
Figure 2.21. Cross-section of the steel superstructure (dimension in mm)
In order to assemble the superstructure above the crowded streets, 8 auxiliary columns
were built with typical steel scaffolding towers. At first the support elements of the
superstructure were placed on the columns and stabilised. Than the span elements were installed
at night while the streets were closed for traffic for 9 hours. All superstructure was assembled
with the mobile crane and then welded on auxiliary columns to form the ring. After welding
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had been finished and the superstructure integrated, it was connected with the CFST pillars built
underneath. Finally, prefabricated ramps and stairs were constructed and footbridge equipment
was placed on the steel deck. The footbridge erection lasted 6 months (Fig.2.23).
Figure 2.22. Steel ring superstructure division into assembly parts
Figure 2.23. Steel ring superstructure assembling on temporary supports
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2.7. Case study No.2. Rotation of the long arch span
The footbridge over the Vistula River in Cracow was completed in 2010 and spans two
riverbanks in the historical city centre, which strongly influenced both its architectural shape
and the construction technology. The footbridge is a 148 m long tied arch steel structure with a
single span main tubular arch of circular cross section and twin suspended semi-circular
orthotropic steel decks with exotic hardwood timber planks used for the wearing surface. The
superstructure has a geometry with an arch radius of R=170 m and the rise of 15,3 m. The
superstructure is very slender with H/L = 1/9.7 (Fig.2.24). The main arch is made of two
concentric steel tubes, with the outer one with diameter of 2020 mm and the inner one with
diameter of 1620 mm. Inside the annulus between these steel tubes, self-compacting concrete
was poured. The main arch has both ends fixed in new concrete abutments, which were built
behind the existing abandoned old masonry abutments from the 19-century demolished bridge.
The twin semi-circular orthotropic steel decks have a travel length of 137 m and a width
of 3,4 m and are suspended from the tubular steel arch with steel rope hangers in the “network”
arrangement. Twin inclined hanger surfaces in a semi-circular shape in the network pattern plus
steel orthotropic decks create a very stiff triangulated structural system, which enhances the
rigid body behaviour of the superstructure (Siwowski & Wysocki, 2015b).
Figure 2.24. Longitudinal cross-section and plan view of the footbridge
The 14 mm thick orthotropic steel deck with open flat plate 100 mm x 10 mm ribs spaced
270 mm is supported with 508 mm diameter tubular (circular) steel pipe crossbeams, suspended
directly from the tubular arch and perpendicular to it. Crossbeam spacing is 5,0 m along the
length of superstructure. The superstructure is X-braced horizontally with 216 mm steel tubes
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or pipes located in a plane of the cross beams. A welded gusset plate connects the centre of the
X-bracing with the midspan of the crossbeams in nodes located along the centreline of the
superstructure (Fig.2.25). There are two external ties made of two 19 x 15,7 mm cables which
are located under the each deck part. The ties are anchored in the end crossbeams and connected
to concrete abutments with special steel brackets. Both ends of the steel tubular arch are fixed
into the new reinforced concrete abutments rectangular shaped as 11 x 13 m with depth of 3,4
m. Each abutment is supported on a set of 10 concrete drilled shaft piles with the depth of about
17 m.
Figure 2.25. The members of steel arch superstructure of the footbridge (dimension in mm)
The Vistula River had to remain open for navigation and the conventional traditional
method of construction on falsework or temporary supports founded in the riverbed, which had
been assumed in the tender design, were challenged by the river administration. Therefore, it
was decided to change the construction technology and to assemble the entire superstructure on
the right bank of the river and to launch it by rotation as a rigid body by means of a floating
support into its final position on the left bank abutment. Before the superstructure rotation
began, both new reinforced concrete abutments supported on pile foundations had to be
constructed. The assembly area for the superstructure was chosen along the right river bank,
and the construction site was adjusted to aid in the rotation via floatation of the superstructure
(Fig.2.26).
The tubular arch was assembled on site on 12 temporary supports with the height of
about 2,5 to 3,5 m, located along the longitudinal central axis of the superstructure (Fig.2.27).
Falsework towers were spaced about 15 m apart. A temporary reinforced concrete slab for
65
superstructure assembly had to be built as a temporary foundation and erection platform for the
scaffolding and falsework towers. Temporary guy ropes were also used between the falsework
towers along the superstructure central axis to ensure the global arch stability during the site
assembling.
Figure 2.26. The assembly of the superstructure along the right river bank
Figure 2.27. Key stages of superstructure site assembly (description in the text)
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The entire steel tubular arch structure was divided into 15 straight sections with the
average length of 5 m and the weight of 24,3 metric tons. All arch sections were connected
together on site with welded butt joints. The concrete was placed in the annulus between the
inner and outer steel tubes. After assembling the tubular steel arch, the falsework towers or
temporary supports were removed and the erection of tubular crossbeams started from the
superstructure centre towards both ends. The twenty-five tubular crossbeams were also
supported on other 50 sets of scaffolding or temporary supports (one at each end of a crossbeam)
with the height of 2,5 to 3,5 m. The weight of the longest crossbeam located at midspan was
about 6,0 tons. Simultaneously with the assembly of the crossbeams, the horizontal X-bracing
with steel tubes was also proceeded. Next, the shop-fabricated orthotropic steel deck panels of
5,0 long x 3,4 m wide plan dimensions and the weight of 3,64 tons were installed. Each panel
end was supported on crossbeams (spaced at 5 m), and was first fixed to them by high strength
friction grip (HSFG) bolts. After that the orthotropic steel deck plate was connected with field
welding. Next, beneath the deck panels, tie cables of galvanized wire rope were installed.
Finally, the inclined wire rope hangers with turnbuckles were installed in the “network” pattern,
connecting the crossbeams to the main tubular arch girder. The steel superstructure was painted
and the timber wearing surface was installed. After assembling the entire superstructure, the
inclined vertical hangers and the horizontal wire rope ties were partially pre-tensioned
following the removal of all the temporary supports. Temporary V-shaped internal frames were
installed at each end of the tubular arch. The superstructure was ready as a rigid body for
rotation and during the erection and floatation processes it would be a tied arch. After the
erection by floatation the superstructure was converted into a fixed arch with steel tube rigidly
fixed to the reinforced concrete abutments. The key stages for the scheme of superstructure site
assembly are shown in Figure 2.27.
Superstructure assembly along with inclined hangers and the partial post-tensioning of
the horizontal tie made it possible to create a very rigid spatial structure, which could be shifted
horizontally and vertically without the danger of unacceptable deformations and / or the loss of
overall stability. However, the construction of additional temporary supporting structures was
also needed for launching to ensure stability and safety of the span during rotation. The scope
of these works comprised:
 erection of V-shaped internal frames at each end of the tubular main arch to enhance span
spatial rigidity
 construction of a temporary support with rotation table on the right river bank (next to the
abutment)
 construction of a special track structure for moving the span end from the bank onto a
floating grillage support
 construction of the floating grillage support on barges
 construction of two temporary supports on both banks adjacent to the permanent abutments.
Two V-shaped internal frames (one per arch end), connecting the arch girder with
rotation tables under the superstructure, were located at 15% and 85% of the superstructure
length. The V-shaped frames provided the correct force distribution in the tubular steel arch
and limited its bending stresses during rotation. The V-shaped internal frames consisted of 6
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inclined members made of steel pipes and an I-shaped transverse bottom steel beam. The
bottom steel beams of each V-frame were supported with specially designed rotational tables,
which were connected with steel pins of 100 mm in diameter. One of the rotational tables rested
on a temporary support structure located on the right river bank, the other - on the floating
grillage support connected to the joint between the large and small barges. After placing the
span on the two rotation tables, the adjustment of tension force in the horizontal tension tie was
needed because of the alternation of span’s support points during launching.
The special track structure for launching was erected on the right bank. Two temporary
support structures or falsework made of steel pipes and I beams were constructed and founded
in the riverbed. On these supports, a track made of four I-beams was placed to enable moving
the span end towards the floating support. The floating support consisted of a large and small
barge (large was 70 m x 10 m in plan) joined together by means of a grillage comprising Ibeams and steel wire ropes, and equipped with the rotation table at the joint between barges. At
the beginning of launching, the floating support on barges was located between two temporary
supports (“dolphins”). When the superstructure had been positioned on the rotation table
located at joint between the barges, the inner temporary support was dismantled and the
horizontal launching by rotation via floatation started. To assist the rotation via floatation, two
additional temporary supports were erected and acted as a falsework on both banks, close to the
new concrete abutments. These temporary supports enabled placing the span after launching in
the proper longitudinal axis and its final positioning to the required elevation of the deck.
Before starting the rotation, accurate scanning of the riverbed was carried out to
establish the maximum immersion depth, the “freeboard” of the barges. The freeboard was
adjusted by means of water “ballast” filling the interior of the barges. The rotation via floatation
of the superstructure proceeded as follows (Fig.2.28):
 lifting of the right-side superstructure end on the rotation table support to the elevation
required for launching
 sliding the left-side superstructure end on the track by means of jacking towards the floating
support barges
 rotation of the entire superstructure approximately an angle of 90° into the left bank
abutment; via the rotation tables, the two barges were always positioned parallel to the
current of the river; tugboats or pusher boats assisted in aligning the barges
 lowering the superstructure after launching via floatation to the required final elevation of
the superstructure and resting it on the temporary supports located on both banks
 connecting the superstructure with steel elements embedded in concrete abutments
 coupling the tie cables with turnbuckles to their short embedded parts cast into concrete
abutments along with the final adjustment of tension (tie) force.
During the rotation the floating support on rotation tables on the two barges were
secured with steel ropes, anchored in the river banks. The rotation via floatation was realized
by means of four wire ropes and hand winches with the assistance of two tugboats (pusher
boats), upstream and downstream the two barges, which enabled 100% control over the
floatation operation. The force needed to move the span towards the floating support was about
600 kN, while the force to rotate only 50 kN. The rate of rotation was about 2 m per minute
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(along the radius) and the whole launching operation lasted only about 6 hours (Fig.2.29). After
the span was connected to the abutments, equipment installation completed the footbridge
construction. The final tuning of inclined hangers in the “network” pattern was the last
operation performed.
Figure 2.28. The rotation via floatation of the footbridge superstructure (scheme)
Figure 2.29. The rotation via floatation of the footbridge superstructure
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The footbridge was built within 12 months, during which the final structural and
technological design was also prepared and completed. It was a short period for the construction
of about 148-metre-long steel bridge. The horizontal rotation via floatation has been proved to
be a reliable and efficient construction method, especially in the city centre, when crossing the
river with everyday water navigation. However, the efficiency of this construction method can
be achieved on several conditions. First, the superstructure to be launched by rotation must be
a single span bridge with the length equal to the river width. The entire structure should be
assembled on one of the banks along the riverbed and must be constructed as a very rigid body.
Enough space for the construction site on the river bank and easy access to the site are also very
important factors. As the described case showed, flood protection of the construction site should
also be kept in mind. Quite a big number of temporary structures need to be custom engineered
and built or rented for construction (the rotation table support, special bearings, the track,
barges, etc.). Finally, special staff experience for a superstructure navigation on the river is
crucial for a successful operation. All of these requirements criteria were met for the discussed
footbridge, which resulted in high contract efficiency for the employer, contractor and the third
party stakeholders, such as the city inhabitants, tourists and river navigation employees
(Siwowski & Wysocki, 2015b).
2.8. Case study No.3. Incremental launching of cablestayed superstructure
The bridge over the San River is located on the bypass of Przemyśl and is the first Polish
cable-stayed bridge in eastern Poland (Siwowski & Zimierowicz, 2013). The structure is made
up of two spans, 114 m long each, one span over the river and the other over historically
protected area. The spans are composite concrete-and-steel structures. Both spans are supported
with cables to the reinforced concrete pylon. The main dimensions of the bridge are (Fig.2.30,
2.31):
 total length Lc = 229,5 m
 theoretical span lengths – 114,0 + 114,0 m
 total width of the deck (without cables) – 21,8 m
 usable widths: carriageway 4 × 3,5 m, cyclists and pedestrians lane 2 x 2,5 m
 span depth - 2450 mm
 pylon height (measured from the road surface) – 50,1 m.
The choice of construction method was affected by the complicated conditions both in
terms of the terrain and legal issues. The river is not navigable for larger ships in the part where
the bridge was constructed. When the structure was erected in summer of 2011 the water level
was so low that it did not exceed 50-60 cm. As the river is non-navigable, the cantilever method
of construction from water was rejected as it was not possible to complete the construction in
time. Moreover, the San River area around the bridge vicinity is included in the Natura 2000
ecological zone. Due to the fact, it was legally impossible to locate any temporary supports in
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the river bed. The assembly with falsework was also rejected. Moreover, one span is located in
the area which is historically protected so any temporary supports or carrying out work in that
area was prohibited. Having considered all these conditions, incremental launching turned out
to be the only economically and technically reasonable method of construction.
Figure 2.30. Side view of the bridge
Figure 2.31. Cross-section of the bridge
Incremental launching was carried out with four temporary supports. The only option
was to erect light steel temporary supports on the banks of the river for the assembly of the
steel girders - and to avoid additional loads from the concrete deck slab by pouring it in-situ
later. Therefore the assembly of the steelwork was performed by incremental launching prior
to the concrete deck slab being poured. The final location of the steel superstructure was
achieved by using an incremental launching method with a temporary steel pylon and four
individual pairs of seven strand back-stay cables. The steel structure of the bridge and the
temporary pylons were shifted between the arms of the concrete pylon.
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The steelwork was divided into transportable segments, regarding both the capability of
the steelwork manufacturer and transport capability of the contractor as well as the lateral
spacing of cable anchorages which amounts to 12 m at the road level. The weight of one
segment was limited to 35 tones and its length – to 25 m. The individual segments of the main
girders and the cross beams were assembled by welding on-site. As the river bed of
approximately 66 m in width had to be crossed twice, it was decided to use the assembly putlog,
110 m in length, located behind one bridge abutment. The assembly of segments on the putlog
was carried out by two mobile cranes. To make the cranes move and to transport the segments,
a special service road was constructed around the assembly site. The assembly of the steelwork
was performed in 3 stages: the first stage was to make the superstructure 122,6 m in length, the
second stage – 42 m, and the third stage – 64,8 m. Before being launched onto supports, the
whole steel structure was additionally strengthened with lateral bracing (Fig.2.32).
Figure 2.32. Bridge steel superstructure on the assembly site:
top view (left) and back view (right)
In order to shorten the distance to be crossed, the temporary supports were located as
close to the river bed as possible. The design of assembly technology considered performing
only 4 temporary supports (PM-1 to PM-4), located beyond the river bed and beyond the
protected area. As the river bank was high and steep, it was necessary to use tall supports,
approximately16 m high. In order to transfer horizontal forces from launching, steel struts were
additionally applied, installed at the upper parts of temporary supports and bridge piers. Due to
span cross-section the maximum length of the cantilever during launching can be approximately
35 m. Thus, to cross the obstacle about 66 m in length, it was necessary to install a temporary
steel pylon, located on the bridge girders and fixed and moved along with the structure. The
installation pylon was made from steel pipes, and temporary stays – from prestressing cables.
Another problem to overcome was the fact that the installation pylon had to fit the clearance of
the concrete pylon, which determined its construction. During the launching operation,
stabilising ballast and a small launching nose were used to ensure the smooth launch of the
main girders on the bearings located on the temporary supports (Fig.2.33).
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Figure 2.33. Bridge steel superstructure being launched: scheme (top), top view (bottom, left)
and side view (bottom, right)
Launching was proceeded in a straight line and with upward inclination of 3%. The
launching of the steelwork was carried out in three stages, with the same segment lengths as in
the assembly operation. The first task to be done was to construct the putlog behind the southern
abutment and PM-4 temporary support (Fig.2.34, left). The steel structure was assembled on
the putlog. Simultaneously, the pylon was being erected and the upper anchorages were being
installed. Before the deck was assembled, the temporary installation pylon was delivered on the
construction site. After it was installed along with the stays, the launching operation was started.
After the first stage of launching was completed, the other bridge sections and PM-3 temporary
support were assembled on the putlog, and another stage of launching operation was carried out
(Fig.2.34, right). Before the third and the last stage of the operation, temporary supports were
erected along the river bank (PM-2 and PM-1). The steelwork reached its target location 6
months after the start of work.
To carry out launching operation two hydraulic jacks for horizontal sliding with fitting
equipment were used. The jacks were supported with two horizontal steel posts, anchoraged at
the abutment. The jacking force was transferred by two steel ropes located inside the structure,
under the crossbeams, in the vicinity of the main girders. The fixed anchorage of the cables was
performed by anchoring the cables in the crossbeams. The launching rate for individual
structure segments was 8 m per hour on average. The steel structure in its final position is shown
in Fig.2.35.
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Figure 2.34. PM-4 temporary support (left) and the superstructure with installation pylon
before PM-3 temporary support (right)
Figure 2.35. Bridge steel superstructure in place: top view (left) and side view (right)
After the steel structure was launched, the stay cables were installed and prestressed.
Two prestressing operations were envisaged: primary and secondary ones, and their order and
the value of forces in the cables were determined. Both operations were performed
symmetrically, with the method of balanced strain (strand by strand), with four light
monostrand hydraulic jacks. Primary prestressing of the cables was performed simultaneously
with concreting the bridge deck. The deck was segmented according to the concreting stages,
individual sections were concreted in the prearranged order and the demands regarding service
interruptions and the time of form stripping were set. The deck was concreted symmetrically
on both sides of the pylon. Secondary prestressing was performed after the deck was completed,
but before installing the bridge equipment. Suitable prestressing of the cables also helped to lift
the spans. The installation of the bridge equipment was the last construction stage (Siwowski
& Zimierowicz, 2013).
At every construction stage, surveyors maintained constant monitoring and control of
the displacements in the pylon in order to compare the real bridge behaviour with structural
calculations. It was not necessary to make any additional adjustments to the stay cables in order
to achieve the final road line after the completion of finishing works. The structure behaved
exactly as intended during construction - the result of a detailed analysis of the structure at the
design stage and well-planned and executed works on site.
74
3. Maintenance techniques for concrete bridges
Many bridges which were designed to previous loading specifications or which have
suffered damage or deterioration are now inadequate for modern heavy goods vehicles. As a
result, bridge owners and manager are faced with the choice of restricting traffic or carry out
rehabilitation. Over the service life of a bridge, its constituent materials (concrete, steel,
masonry) are continually subjected to fatigue and wear-tear due to dynamic vehicular loads.
Overloading due to increase in wheel loads and regular exposure to aggressive external
environment may aggravate the situation further. Poor quality of construction and lack of
regular maintenance could potentially lead to major retrofit in a bridge structure. Post-tensioned
concrete bridges may also exhibit loss in pre-stress overtime, resulting in drop in load carrying
capacities of the affected members. Defects in the constituent materials may be manifested in
the form of cracking - spalling of concrete, excessive deflection of structure, corrosion of steel
components/ reinforcement etc. It is evident that rehabilitation of bridges involves addressing
a myriad of problems and no single technique or retrofit method could offer a complete solution.
Therefore, answer lies in being able to address each individual problem with an appropriate
technique to result in a durable retrofit. This chapter along with the two following aim at
providing an overview of a variety of structural retrofit techniques available to rehabilitate
bridges.
About one third of the approximately 35.000 highway bridges in Poland were built
before 1940, and many have not been adequately maintained. Most of these bridges were
designed for lower traffic volumes, smaller vehicles, slower speeds, and lighter loads than are
common today. In addition, deterioration caused by environmental factors is a growing
problem. According to GDDKiA (General Directorate of Polish Roads and Highways, Poland),
almost 30% of the nation’s bridges are classified as deficient and in need of rehabilitation
(gddkia.gov.pl). Many of these bridges are deficient because their load-carrying capacity is
inadequate for today’s traffic. Strengthening can often be used as a cost-effective alternative to
replacement or posting.
The deterioration of the existing bridges is today one of the most important problems to
cope with by contemporary bridge engineering. It is an issue of technical, economic and social
nature and concerns bridge infrastructure in many countries, including Poland and the other
highly developed states. Many bridges are structurally deficient and functionally obsolete so
they require to be repaired or modernised.
Bridge modernisation can be classified into two fundamental groups – structural and
functional (Radomski, 2002). The first one concerns mainly the insufficient load carrying
capacity of the structure, while the latter – limited service parameters (e.g., too narrow bridge
75
deck). Structural strengthening of existing bridges is a crucial problem in Poland, where many
bridges require to be strengthened. In general, bridge strengthening can be applied mainly in
two situations – firstly, when the load carrying capacity of the structure requires to be restored
to the original level assumed at the design stage, and, secondly, when the load carrying capacity
of the structure requires to be upgraded due to the increase of the service loads. However, in
any case the strengthening methods are the same. One of the newest ones is strengthening with
the carbon fibre reinforced polymer (CFRP) composites.
The live-load capacity of various types of bridges can be increased by using different
methods, such as (1) adding members, (2) adding supports, (3) reducing dead load, (4)
providing continuity, (5) providing composite action, (6) applying external post-tensioning, (7)
increasing member cross section, (8) modifying load paths, and (9) adding lateral supports or
stiffeners. Some methods have been widely used, the others are new and have not been fully
developed. All strengthening procedures presented in the following three chapters apply to the
bridge superstructure. This chapter is devoted to repair, strengthening and rehabilitation of
concrete bridges, whereas the following two chapters deal with the same subject regarding to
steel and masonry bridges. Most of the presented examples are based on long-term author’s
experience.
3.1. Concrete repair
Concrete is strong in compression, but comparatively weak in tension. The inclusion of
steel reinforcement improves the tensile strength of the element and, consequently, its resistance
to stress. Unfortunately, in the presence of water and oxygen, steel corrodes. Therefore, in
almost all likely exposure conditions for reinforced concrete, corrosion of the steel is a potential
hazard.
The materials and methods of concrete repair are numerous, ranging from simple spot
patching to more modern strengthening systems involving advanced fibre reinforced polymer
(FRP) composites. However, the long-term performances of many of these methods have been
less than ideal. For example, when conventional patching is utilised, the chlorides within the
surrounding contaminated concrete can travel along the strands into the patched area and reinitiate corrosion. Also, the imperfect bond between the old and new concrete may lead to
separation or delamination allowing further entry of chlorides. Finally, the reinforcing steel bars
that are surrounded by old and new concretes are susceptible to galvanic corrosion due to
varying environments along the length of each bar.
Repairing concrete that is more than superficially damaged is expensive and
problematic. Since many members can be completely replaced for less than the cost of extensive
repair, aggressive replacement of deteriorated members should be pursued. Salvaging concrete
containing corroding reinforcing steel or critically saturated aggregate does not often result in
a long-lasting component since the substrate concrete repaired is only marginally better than
the unsound concrete removed. Any time there are major and extensive repairs being proposed
to concrete structures, in depth and thorough investigation of the condition of the concrete will
76
be required. This investigation must include, but is not limited to, hand investigation with a
chipping hammer, drilling into unsound concrete to determine the depth of deterioration, and
concrete cores. It is the designer's responsibility to evaluate the repair areas and determine the
most suitable repair method.
The methods available for repair of concrete bridges with corroded steel reinforcement
fall into two general categories: non-electrical (conventional) methods and electrical methods.
Conventional methods include sealers and coatings, patching, overlays, or combinations of the
above. The main electrical method is cathodic protection. Chloride extraction using electrical
methods has also been performed. In prestressed concrete, the use of electrical methods is not
common due to significant risk for hydrogen embrittlement.
In existing concrete, sealers and coatings can be used to form a seal that reduces the
permeability of concrete to new intrusions. However, the effectiveness of such external
measures would diminish substantially when they are applied to concrete members that have
already experienced significant chloride intrusion. Sealers generally fall into three categories:
penetrating sealers, coatings, and membranes. Some sealers are designed to seal existing cracks
and act similar to membranes without trapping moisture inside. More recently, there are claims
of post-application sealing of future hairline cracks in the presence of moisture through
autogenous healing.
Corrosion can be controlled by controlling the flow of electrical current. There are two
types of cathodic protection systems: an impressed current system and a galvanic (sacrificial)
anode system. An impressed current system uses an external power source to provide the
current to the protected metal. A galvanic (sacrificial) anode system uses a metal that is higher
in electro-potential relative to steel reinforcement to produce the current. More recently,
methods for retrofitting galvanic and impressed current anodes into existing concrete have been
developed and are commercially available. Also, galvanic cathodic protection methods in
combination with patch repair have been proposed.
Application of repair mortars
The basic patching process involves removal of damaged and spalled concrete, cleaning
and possible coating of steel and placement of the new patch material. The European Standard
EN 1504 defines the procedures and characteristics of products used to repair, maintain and
protect concrete structures. Concrete repair according to this standard defines the application of
repair mortars by hand as well as by machine (Fig.3.1).
Machine applied concrete repair (shotcreting) is ideal on projects with high repair
volumes. The cost efficiency is much better than application by hand and the quality of the
hardened repairs is more homogeneous. Machine application of repair mortars therefore not
only brings the benefits of increased output, but it also helps to improve the repair durability.
Machine application should be selected over manual application whenever the economic and
environmental conditions are appropriate. There are two different types of machine application:
the “dry” spray process and the “wet” spray process. In the “dry” spray process the mortar is
put into the pump hopper as a dry powder and fed by compressed air in a thin stream to the
point of application. At the back of the nozzle the powder is mixed with water, including
77
accelerator if required and propelled to the substrate at a pressure of about 2 bar. In the “wet”
spray process the mortar is first mixed and then fed in a “plastic” consistence and as a dense
stream by pump to the point of application. Air is added at the nozzle, which then dissipates the
mortar and propels it onto the substrate. Accelerators cannot be added at the nozzle.
Figure 3.1. Hand (left) and machine (right) applied concrete repair
Table 3.1. Most important selection criteria for concrete repair method
Application Method
Criteria
Machine
Hand
“Dry”
“Wet”
Cost efficiency for large repair volumes (>500kg/day)
***
***
0
Cost efficiency for small repair volumes <200kg/day)
*
*
***
Installation costs
*
*
***
Overhead application and extensive thick layer reprofiling
***
***
*
Compaction behind exposed reinforcement
***
***
*
Limited working space on site
*
**
***
Larger transport distance for material
***
**
*
Repair under dynamic load
***
***
*
Dust, noise
0
*
***
Bond strength
***
***
*
Possibility for high early strengths
***
*
**
Economical aspects
Application aspects
Environmental aspects
Technical aspects
0 = not suitable; * = possible; ** = good; *** = very good;
78
The selection of the best application method and the most appropriate repair material is
dependent on the different requirements of each project and the site conditions and its situation.
The evaluation process to decide on the best application method is dependent on different
application and performance requirements that can obviously vary greatly on different types of
project and from site to site. The Table 3.1 gives the most important criteria to consider and
then select the most suitable and efficient application method.
Concrete repair works often have to be carried out while the bridge remains in service,
and a bridge in service is subject to vibrations. The standard EN 1504 does not yet take this
situation into consideration. Repair products meeting the highest requirement may still fail in
terms of their bond behaviour when the application is done in such difficult conditions.
Therefore only the use of systems that have undergone specific additional testing is
recommended, to confirm their application characteristics, bonding and curing performance in
dynamically loaded situations.
Concrete crack repair
Following the evaluation of the cracked bridge structure, suitable methods of concrete
crack repair procedure can be selected. Successful repair procedures take into account the cause
of the cracking. For example, if the cracking was primarily due to drying shrinkage, it is likely
that after a period of time the cracks will stabilise. On the other hand, if the cracks are due to a
continuing foundation settlement, repair will be of no use until the settlement problem is solved.
Figure 3.2. Bonding the crack by the injection of epoxy: injection set (left) and establishing
entry and venting port (right)
Cracks as narrow as 0,05 mm can be bonded by the injection of epoxy. The technique
generally consists in establishing entry and venting ports at close intervals along the cracks,
sealing the crack on exposed surfaces, and injecting the epoxy under pressure (Fig.3.2). Epoxy
injection has been successfully used in the repair of cracks in concrete bridges. However, unless
the cause of the cracking has been corrected, it will probably recur near the original crack. If
the cause of the cracks cannot be removed, then two options are available. One is to rout and
seal the crack, thus treating it as a joint, or, establish a joint that will accommodate the
movement and then inject the crack with epoxy or other suitable material. With the exception
of certain moisture tolerant epoxies, this technique is not applicable if the cracks are actively
79
leaking and cannot be dried out. Wet cracks can be injected using moisture tolerant materials,
but contaminants in the cracks (including silt and water) can reduce the effectiveness of the
epoxy to repair the cracks structurally.
Routing and sealing of cracks can be used in conditions requiring remedial repair and
where structural repair is not necessary. This method involves enlarging the crack along its
exposed face and filling and sealing it with a suitable joint sealant (Fig. 3.3). This is a common
technique for crack treatment and is relatively simple in comparison to the procedures and the
training required for epoxy injection. The procedure is best applicable to approximately flat
horizontal surfaces such as bridge decks; yet, routing and sealing can be performed on vertical
surfaces (with a non-sag sealant) as well as on curved surfaces (pipes, piles and pole). Routing
and sealing is used to treat both fine pattern cracks and larger, isolated cracks. A common and
effective use is for waterproofing by sealing cracks on the concrete surface where water stands,
or where hydrostatic pressure is applied. This treatment reduces the ability of moisture to reach
the reinforcing steel or pass through the concrete, causing surface stains or other problems.
Figure 3.3. Routing and sealing of cracks: scheme (left) and cracks treating (right)
Low viscosity monomers and resins can be used to seal cracks with surface widths of
0,03 to 2 mm by gravity filling. High-molecular-weight methacrylates, urethanes, and some
low viscosity epoxies have been used successfully. The lower the viscosity, the finer the cracks
that can be filled. The typical procedure is to clean the surface by air blasting and/or water
blasting. Wet surfaces should be permitted to dry several days to obtain the best crack filling.
3.2. Strengthening & rehabilitation
In response to the demand for faster and more efficient transportation systems, there has
been a steady increase in the weight and volume of traffic using national highway systems
throughout the world. As well as increases in legal vehicle loads, the over-loading of vehicles
is a common problem and this must also be considered when assessing bridges. In response to
increased traffic flow, bridge widening is often carried out to increase the capacity of the road
network. As a result, many bridges are now required to carry loads significantly greater than
80
their original design loads. Bridge strengthening as an alternative to complete replacement can
provide an effective and economic solution in appropriate situations.
The selection of an appropriate method for strengthening a particular bridge depends on
a number of factors. The type of structure, the magnitude of the strength increase required and
the associated costs are the main parameters to be considered. Many strengthening schemes are
applicable to particular structural types and have limits on the extent to which strength can be
increased. Strengthening costs would certainly be lower than bridge replacement, but the
selection of a particular method of strengthening would need to be justified on economic
grounds. It is important to consider, not only the initial capital costs of the strengthening project,
but also the maintenance costs associated with the future in-service behaviour. The condition
of the existing bridge is an important consideration. If the bridge is in bad condition, then future
maintenance and safety problems might override the benefits of the reduced capital costs of
strengthening and provide justification for bridge replacement. The strength and condition of
the substructure must not be ignored and strengthening should not proceed without giving due
consideration to the capacity of the bridge piers, abutments and foundations. The difficulties
associated with traffic management and the costs arising from traffic delays should be
considered in the economic justification. In some cases, this may limit the use of certain
methods of strengthening.
Among the modern methods of reinforced concrete (RC) bridge strengthening one can
recognise three main groups: increasing the element cross-section with concrete (e.g.
shotcreting with additional reinforcement), bonding steel or composite strips or sheets to the
element surface, and external prestressing of the element. The selection of an appropriate
method for strengthening a particular bridge depends on a number of factors (Radomski, 2002).
The type of structure, the magnitude of the strength increase required and the associated costs
are the main parameters to be considered. Many strengthening schemes are applicable to
particular structural types and have limits on the extent to which strength can be increased.
Strengthening costs would certainly be lower than bridge replacement, but the selection of a
particular method of strengthening would need to be justified on economic grounds. It is
important to consider, not only the initial capital costs of the strengthening project, but also the
maintenance costs associated with the future in-service behaviour. The condition of the existing
bridge is an important consideration. If the bridge is in bad condition, then future maintenance
and safety problems might override the benefits of the reduced capital costs of strengthening
and provide justification for bridge replacement. The strength and condition of the substructure
must not be ignored and strengthening should not proceed without giving due consideration to
the capacity of the bridge piers, abutments and foundations. The difficulties associated with
traffic management and the costs arising from traffic delays should be considered in the
economic justification. In some cases, this may limit the use of certain methods of
strengthening.
Depending on the bridge configuration and the expected service life of the bridge after
strengthening, other factors might need to be considered before a particular scheme is adopted.
The durability, inspectability and replaceability of components of the rehabilitated bridge are
very important aspects. For some strengthening systems, the ability to monitor the behaviour
81
of the strengthened bridge might need to be considered, particularly where an innovative
method is being applied. The ability to adjust the level of strengthening in future to allow for
further increases in traffic loads might provide useful benefits. The appearance of the bridge
after strengthening is an important consideration and should not be ignored. While bridge
aesthetics have always played an important role in the design of major structures, public
perception has often been ignored for short-span bridges. It is now beginning to change and an
emphasis is now being placed on what bridges look like. The use of intermediate supports or
props, or strengthening methods which appear unsightly, while tolerable as temporary
measures, are becoming less acceptable as long term solutions.
Section enlargement
Section enlargement is placing additional layer of concrete surrounding an existing
beam. Concrete jacket is to add reinforced concrete jacket on the existing beam. Jacketing by
reinforced concrete improves resistance against applied loads and enhances the durability at
same time. Furthermore, section enlargement and concrete jacketing may be easier and cheaper
compared to other methods such as steel plate jacketing.
Sprayed concrete is one of the common enlargement methods. Using sprayed concrete
to strengthen reinforced concrete bridge beams can effectively increase their load carrying
capacity or stiffness. Furthermore, additional fibres in sprayed concrete improve the crack
pattern and ultimate capacity of RC beams. Adding fibres to reinforced concrete beams improve
flexural strength, enhanced cracking pattern, reduced tensile stress and greatly increased the
first cracking moment.
Figure 3.4. Concrete jacketing technique to rehabilitate and strengthen concrete beam (left,
source: sainkaconstructions.com, 12.12.2015) and column (right)
Similar to section enlargement method, concrete jacketing can be easy, effective and
inexpensive technique to rehabilitate and strengthen concrete structures (Fig.3.4). Concrete
jacket is achieved by adding the reinforced concrete jacket to the existing structure components
such as beams and columns. The experimental results revealed that the reinforcement should
be adequately anchored past the point of contraflexure and the support of simple beams. Also
adequate anchorage is necessary for additional stirrups near the underside of flange. The fully
82
anchored stirrups contribute fully to the strength of the jacketed beam. The effect of roughening
in interface does not influence the behaviour of jacketed beams very much.
Jacketing with reinforced concrete and section enlargement may be the relatively easy
and economic strengthening methods compared to the other methods of strengthening. They
effectively increase the load carrying capacity or stiffness. However, the addition of concrete
and steel to repair beams increases the weight of beams. So, the lightweight concrete may be
considered as better applied when strengthening the beams. Strengthening with concrete and
steel rebar might lead to corrosion in beams. Hence, section enlargement and concrete jacketing
are limited to use in harsh environment and the protecting corrosion is important work.
Attachment of external steel plates
Attaching steel plates to certain external surface of the beams is another popular
strengthening technique. Anchoring or bonding steel plates to reinforced concrete beams can
increase flexural and shear capacity. Furthermore, it can control deflections and cracking of
beams. The efficiency of steel plates is influenced by some factors such as the dimension of the
steel plate, the arrangement of bolts, and bonding method.
The steel plates or shapes are normally attached by bolting, keying, or doweling to
develop continuity between the old beam and the new material. If the beam is also inadequate
in shear, combinations of straps and cover plates may be added to improve both shear and
flexural capacity. Because a large percentage of the load in most concrete structures is dead
load, for cover plating to be most effective, the structure should be jacked prior to cover plating
to reduce the dead-load stresses of the member. The addition of steel cover plates may also
require the addition of concrete to the compression face of the member. Bolting steel plates to
the bottom and sides of beam/slab sections has also been performed successfully, as
documented by the Xanthakos (1996) or Radomski (2002). Bolting may be an expensive and
time-consuming method, because holes usually have to be drilled through the old concrete.
Bolting is effective, however, in providing composite action between the old and new material.
The obvious advantage of using this strengthening method is that it needs relatively
short installation time and the steel plates do not disrupt operations like e.g. concrete jacketing.
The disadvantages include debonding, expensive, temporary weakening, and corrosions.
Recently, the external steel plates bonding for strengthening purposes have often been replaced
with the FRP composite strip and/or sheet systems (see p.3.3 below).
External post-tensioning
The use of external prestressing as a means of strengthening or rehabilitating existing
bridges has been used in many countries and has been found to provide an efficient and
economical solution for a wide range of bridge types and conditions. The technique is growing
in popularity because of the speed of installation and the minimal disruption to traffic flow
which can, in many cases, be the critical factor in decisions regarding strengthening. The
principle of external post-tensioning is the same as that of prestressing, i.e. the application of
an axial load combined with a hogging bending moment to increase the flexural capacity of a
83
beam and improve the cracking performance. It can also have a beneficial effect on shear
capacity.
Post-tensioning as the way of bridge strengthening has been in use since the 1950s and
there are many examples of its applications throughout the world. In the many situations where
the technique has been applied, the prestress is applied through prestressing cables, either single
or grouped strand. In some applications, the stress has been applied through high tensile bars,
jacked either using hydraulic jacks or with fine screw threads.
External post-tensioning can be applied to improve the serviceability behaviour of
existing bridges. As in prestressed construction, the method can be used to delay or prevent the
onset of cracking in concrete bridge decks. It can also be used to reduce or close pre-existing
cracks. This improvement in cracking behaviour also increases resistance to reinforcement
corrosion. The resulted increased stiffness can reduce in-service deflections and vibrations. The
stress range can also be reduced and the fatigue performance can be improved. The deformation
or sag in a bridge can be reduced or removed.
Figure 3.5. Tendon configurations for flexural post-tensioning of beams: (a) eccentric tendon,
(b) polygonal tendon, (c) polygonal tendon with compression strut, (d) king post, (e) eccentric
tendons and (f) polygonal tendons
84
The use of external prestressing as a strengthening or rehabilitating method has been
popular in many countries and has been found to provide an efficient and economical solution
for a wide range of bridge types and conditions. The technique is growing in popularity due to
the speed of installation and the minimal disruption to traffic, which can, in many cases, be the
critical factor in decisions regarding strengthening. Known applications of post-tensioning of
concrete beams are presented in Figures 3.5 – 3.7. The tendon schemes in general, appear to be
very similar to reinforcing bar patterns for concrete beams (Fig.3.5). Post-tensioning also has
been used for shear strengthening, in patterns very much like those for stirrups in reinforced
concrete beams. Figure 3.6 illustrates a pattern of external stirrups for a beam in need of shear
strengthening. The types of post-tensioned external stirrups have been used or proposed for
reinforced concrete beams and for prestressed concrete box-girder bridges. In Figure 3.7 one of
the most recent applications of external prestressing for concrete bridge strengthening is shown.
Recently, the standard external post-tensioning for strengthening purposes has been often
replaced with the use of prestressed FRP composite strip systems (see p.3.3 below).
Figure 3.6. Tendon configurations for shear post-tensioning: (a) external stirrups, (b) external
stirrups with lateral tie
As with all bridge strengthening methods, there are various advantages and
disadvantages associated with the use of external post-tensioning. There are a number of distinct
advantages which have added to the increasing popularity of this method. These are listed as
follows:
 the method is economic and easy to use, particularly when single strand jacks are employed;
anchorages and deviators are easy to detail and simple to install
 both flexural and shear strength can be increased without the penalty of increased dead load
85



the ease of inspection increases the reliability of the bridge as any stress loss or damage due
to impact or corrosion can be determined by simple inspection procedures
the tendons can be re-stressed and fully replaceable
the strengthening of a bridge can be carried out without disruption to traffic flow on the
bridge.
Figure 3.7. External prestressing of a concrete bridge:
side view (top) and bottom view (bottom)
As with other methods of strengthening, there are disadvantages and it is important that
these be understood in order to make an enlightened evaluation of this method. The main
disadvantages are as follows:
 application of the method is very dependent on the existing condition of the bridge
 loss of stress due to creep and relaxation are an inherent part of post-tensioning
 installation of deviators and anchorages can be difficult, and careful detailing is required to
account for stress concentrations in the existing deck components
 where tendons need to be installed below the bottom flange, the decreased headroom is a
distinct disadvantage
 the external tendons, as in all cable stayed structures, are more susceptible to accidental
damage from fire, impact and acts of vandalism and public access to the tendons should be
prevented.
The conclusion is that the method of external post-tensioning as a method of
strengthening existing bridges has both advantages and disadvantages. Careful consideration is
required before an effective strengthening system can be devised.
86
3.3. Strengthening with externally bonded FRP composites
For more than 20 years FRP composite materials have become increasingly popular in
the bridge industry for strengthening purposes (Meier, 1992, Seible, 1995, Walser & Steiner,
1997). These materials offer several advantages such as high strength-to-weight ratio, light
weight, ease of transport and installation, thermal stability, excellent fatigue characteristics,
electromagnetic neutrality, the ability to tailor mechanical properties and non-corrosiveness
(durability). Moreover, by prestressing a laminate, the ultra-high tensile strength of carbon
fibres can be utilized and more advantages are brought to the strengthening technique (Andrä
& Maier, 2000).
The use of CFRP in bridge strengthening in Poland has developed significantly for the
last 20 years (Siwowski & Radomski, 1998). Since 1997 plenty of bridge strengthening projects
have been prepared and executed. Parallel to field applications many research projects have
been carried out since the late 90s (Siwowski & Radomski, 2015). Their results have been
gradually affecting the efficiency of CFRP strengthening methods and making them more
familiar and easier to design and use on site. The author has been deeply involved in this
development ever since, both in the university research and in the field applications. The most
interesting examples have been presented in this sub-chapter.
The first Polish applications of CFRP for bridge strengthening
An idea to use CFRP composites for bridge strengthening is at present well known and
relatively widely applied. Its first world application was recorded in 1991 in Switzerland
(strengthening of the Ibach bridge, near Lucerne), while the first Polish bridge application of
the method took place in 1997 and concerned the bridge in Przemyśl over the Wiar River
(Siwowski & Radomski, 1998). This bridge has three spans with the lengths of 10,8 + 31,0 +
10,8 m and four RC girders with the cross-section of 1,80 m / 2,80 m in height (middle/support)
and 0,5 m in width.
Figure 3.8. Bridge over the Wiar River in Przemyśl – the first Polish application of CFRP
strips for bridge strengthening – 1997: bridge after modernization (left) and CFRP
strengthening of the beam (right)
87
Due to the increasing heavy traffic towards the border post situated next to the bridge,
the bridge strengthening was required. Initially steel bonding was considered, but as the large
number of steel plates was needed, CFRP strips were taken into consideration. Since the
deficiency in carrying capacity was comparatively small (about 25 per cent) and additional
reinforcement had to be placed only in the middle of the longest span, it was decided to utilise
the new technology. The CFRP strips were arranged and glued to the bridge main girders in
one and two layers and with the different lengths according to the bending moment envelope
(Fig.3.8). The first application of the new strengthening method was preceded by the
development of the original design procedure because the method itself was beyond the Polish
standards and recommendations and its use required approval of the relevant road
administration, which was not an easy process like any time when a new method is
implemented.
The CFRP strips are normally used for the structural bridge strengthening for bending.
It should be pointed out that Poland is a world forerunner in the application of the CFRP sheets
(mats, fabrics) for shear strengthening of bridge structures (so called wrapping technology).
The first world application of the CFRP sheets took place in 1998 on the bridge over the Bystry
Canal in Augustów (Fig.3.9). The 50-year-old RC bridge has one span and two cantilevers with
the theoretical lengths of 4,3 + 12,1 + 4,3 m and five girders with the cross-section of 0,92 /1,32
m in height (middle/support) and 0,36 m in width. The bridge, located on the important road
from Poland to Lithuania, required to be strengthened at short notice due to the really intensive
traffic. The strengthening for bending was made with the CFRP strips while the strengthening
for shear with the CFRP sheets. The proof test showed that the aforementioned strengthening
combination is structurally efficient, moreover, the application of both methods is less timeconsuming.
A
B
1
100
60
430
606
606
430
1
Figure 3.9. Bridge over Bystry Canel in Augustów – the first world application of shear
strengthening with CFRP sheets (1) (dimension in cm)
Among the first Polish applications there is also the use of the CFRP strips for
strengthening the concrete bridge over the Warta River in Śrem (Fig.3.10). In this case, the
arrangement of the strips did not correspond with the bending moments envelope as it was in
Przemyśl - they were located along all the lengths of the post-tensioned concrete precast girders
(i.e. from one support to another) because of heavy corrosion of some prestressing tendons. In
the situation when a single cable is cracked due to its corrosion, the relevant internal forces are
88
transferred to the CFRP strips. Therefore, the main idea of using the CFRP strips resulted from
the bridge safety condition.
Figure 3.10. Bridge over Warta River in Śrem – strengthening of the approaching viaduct -1998
State-of-practice in Poland – some examples
Today, upgrading bridges with the CFRP strips and sheets is a widely accepted repair
and strengthening method in Poland. A large number of bridges all over the country have been
strengthened with this technique. For example, the combination of CFRP strips and sheets has
been applied for many bridges in Poland. One of them is the overpass in Czarlin near Tczew,
shown in Figure 3.11, left. Another example is complex strengthening applied on the RC bridge,
built in 1964 in Libusza over the Ropa River. The continuous three-span bridge with the span
lengths of 23,1 + 30,1 + 23,1 m and two 1,8 m girders in depth has been strengthened with
CFRP strips in positive moment regions and shear strengthening along the girders with CFRP
sheets has also been executed (Fig.3.11, right).
Figure 3.11. The combined structural strengthening for bending and shear: bridge in Czarlin 1998 (left) and bridge in Libusza - 1998 (right)
Another interesting example of CFRP strengthening technique is upgrading the pillars
of RC viaduct in Rzeszów. The main reason for upgrading was the corrosion of the
reinforcement in several columns due to leaky expansion joints. The loss in main longitudinal
89
rebar cross-section exceeded 30 per cent and the stirrups were completely broken.
Strengthening comprised shotcreting for supplementing concrete section, bonding the CFRP
strips along the main rebars and bonding the CFRP sheets round the columns to complete the
lack of stirrups (Fig.3.12).
Figure 3.12. Viaduct in Rzeszów – strengthening of RC columns – 1999: I phase - strip
strengthening (left) and second phase – sheet strengthening (right)
Prestressing systems for CFRP strips
For the recent 20 years several various prestressing systems for post tensioning with the
CFRP strips have been developed. The first one was invented by Leonhardt, Andrä & Partners
and after about 10 years of applying it on site it was improved - its second generation is named
Sika LEOBA CarboDur II (Andrä & Maier, 2000). The second system available on the market
is Sika StressHead, invented in the Swiss EMPA institute. In the early 2000s, the Swiss
company S&P Clever Reinforcement Company AG introduced a new system for CFRP strips
prestressing (Berset et al., 2002). In the USA the main development of CFRP prestressing
systems was carried out at the University of Missouri-Rolla (EL-Hacha et al., 2003). All of the
systems have anchorages. Recently a new system without anchorages has been introduced with
gradually decreased prestressing along a strip towards the ends. In this method the stiffness of
the adhesive is modified by applying the heat to the adhesive (Kotynia et al., 2013).
The Polish experience in the development of CFRP prestressing method regards two
systems. The Polish modification of Sika LEOBA CarboDur II system has been introduced by
IBDiM institute in Warsaw (Łagoda, 2005). The strengthening system is proposed as universal,
fitting the strengthened structures of different materials and various shapes. Various
possibilities of introducing prestressing force have also been taken into consideration. This
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system offers greater freedom in the realization of prestressing. CFRP strips stressing may be
executed by any device producing the force (stretching machine, press, hydraulic cylinders).
The system provides the possibility of multiple use of tensioning jaw elements. Depending on
the applied equipment, the anchorage jaws at the stressing end may be pushed or pulled. In
anchorage zones cavities in concrete are needed for the location of resisting blocks. The steel
anchorage bottom plate is glued to the concrete surface and additionally bolted with M12
chemical anchors.
Figure 3.13. The steel anchorage (left) and the main parts of the Neoxe prestressing system:
steel anchorages, leading rails, reaction blocks and tongs-shaped stretching device (without
hydraulic jack)
The second Polish prestressing system developed by Neoxe Company consists of two
main elements: special steel anchorages mounted on both ends of a single CFRP strip and a
relevant stretching device (Fig.3.13). Each anchorage is made up of two steel plates, 2 mm
thick, welded together along the edges to create a pocket. The end of the strip is placed and
fixed in a steel pocket and bonded with special epoxy-based adhesive. It is followed by gripping
steel plates and a CFRP strip in-between with small rivets. The stressing anchorage has two sets
of holes: the first set is used to fix a reaction block and transfer the tension force from the
hydraulic jack and the second set is used to fix the strip to the strengthened structure. The strips
with determined length are delivered on site as ready-to-install, i.e. with two steel anchorages
mounted on both ends. Before strip tensioning the passive anchorage is bolted to a beam. During
tentative CFRP strip fixing (without glue) the stressing anchorage is inserted in-between two
leading rails which have been earlier fixed to the flange to mount the stretching device. The
leading rails allow the movement of the anchorage towards the force action and simultaneously
hold it down to the structure base. The rails on one end are bolted together with a beam
constituting the stationary reaction block for the stretching device, whereas the second reaction
block is bolted to an active anchorage. The tongs-shaped stretching device is opened out by a
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hydraulic jack. The ends of the tongs are inserted into the reaction blocks of the active
anchorage and afterwards opened, tensioning effectively a CFRP strip (Siwowski et al., 2010).
3.4. Case study No.1. Strengthening by section
enlargement
The RC viaduct was constructed in 1951. Its original design live load according to the
old Polish standards was 150 kN truck. For about 40 years of service, live load on the viaduct
has more than doubled. As the viaduct is situated along a significant road, the local roads
authorities decided to strengthen the structure up to 300 kN, i.e. to the present design live load
capacity standard for bridges on national roads. The viaduct slab superstructure is complicated
both geometrically and statically. The slabs of each span are skewed at a skew angle of 66°.
The 4-span frame with spans of 10,0 + 13,0 + 10,0 + 8,4 m is made up of a continuous RC slab
0,5 m deep, supported with 5 pairs of columns cast in place with the superstructure. Open
abutments and piers are founded by footings directly on the subsoil. There are hinges between
columns and footings (Fig.3.14). The width of the entire deck slab is 10,4 m with 7,6 m
carriageway and two pedestrian lanes of 1,25 m. Due to the insufficient maintenance, the
technical condition of the viaduct was very bad.
The actual carrying capacity of the viaduct determined by FEM analysis test was less
than 150 kN, i.e. below the lowest class according to the current Polish standard (Siwowski,
1992). Such low capacity was mainly due to the lack of a transverse reinforcement across the
slab and longitudinal bars in the negative moment regions of the frame. Therefore the basic
strengthening work was to add the required number of rebars in both the top and bottom parts
of the concrete slab section (Fig.3.15). An additional reinforcement was estimated using the
allowable stress method with the assumption of the internal forces according to the previous
static analysis for a capacity of 300 kN. The top reinforcement, both transverse and longitudinal,
was placed in the additional layer of concrete, placed on the existing deck and composited with
old concrete by means of gunfixed studs. A reinforcement mesh was welded to the studs and a
new 12 cm thick layer of concrete was cast with appropriate falls for drainage.
Figure 3.14. Longitudinal section of the viaduct (dimension in cm)
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Figure 3.15. Cross-section of viaduct: before strengthening (left) and after strengthening
(right) (dimension in cm)
The bottom reinforcement (only transverse required) was fixed underneath the slab
using specially designed steel anchor plates and left uncovered. Before fixing the bars, the
bottom surface of the slab was cleaned with sandblasted. Some cracks and hollows with open
reinforcement were found after cleaning the surface. The cracks were repaired with epoxy resin
injection. Corroded rebars were blasted and protected. Then the hollows were filled with epoxy
mortar. Finally a mesh of rebars was fixed and the whole bottom surface of the concrete slabs
was covered by shotcreting. After the concrete layer had set, steel plates were fixed to both side
surfaces of the slab (Fig.3.16). Two ways of fixing were used: bonding steel plates to concrete
surface with epoxy resin and bolting them with M-28 bolts fixed in the concrete. An additional
bottom transverse reinforcement was attached to the steel plates. At both sides of the slab φ 16
mm rebars were bolted thanks to the threaded ends. Bolting enabled the rebars to be tightened
and let them contribute to the load distribution. The steel plates and rebars were protected and
left uncovered. It will facilitate, if necessary, adjusting stress in the reinforcement and its
replacement when corroded.
Fig.3.16. Placement and fixing of additional transverse rebars: cross-section (top) and side
view (bottom)
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Apart from the work associated with direct strengthening of the spans, the concrete slab
was adapted to the modern maintenance and service requirements. During modernisation,
waterproofing and asphalt pavement were replaced and new pedestrian sidewalks were created.
Transverse and longitudinal falls shaped in the concrete upper layer ensured good pavement
drainage (Siwowski, 1992).
3.5. Case study No.2. Rehabilitation of precast concrete
bridge
The construction of fully prefabricated bridges has been the predominant technology of
concrete road bridge building in Poland since late 60s of the last century. Structures with precast
concrete beams constitute 18% of all Polish bridges and 67% of the concrete ones, constructed
in 1956-1990. Generally, within the Polish public road network there were about 5300 precast
concrete bridges (Biliszczuk & Machelski, 1989). A general tendency to industrialise bridge
construction led to the practical application of over 20 different systems of precast bridges.
Regarding to the span length, the bridges with the following types of precast beams were
constructed: reinforced - up to 15 m, prestressed - up to 28 m, post-tensioned - up to 36 m. Most
of the precast bridge systems are fully prefabricated with minimum of cast-in-place concrete.
Moreover, multispan viaducts were mostly constructed as a row of simply supported beams
with many expansion joints. At the beginning, the deck slab on beams was precast and joined
to them by being welded to the suitably protruding reinforcement elements. The deck slab was
then usually covered only with a thin (3-4 cm) concrete layer. Some time later the reinforced
deck slab began to be made on the beams and thus combined with them. However, several years
passed until the latter method become widespread and replaced the former one. By that time,
as many as a few thousands totally prefabricated bridges had been constructed in Poland and
now, due to their low durability, they need urgent reconstruction. The analysis of the causes of
damages in prefabricated spans shows that their poor durability results, first of all, from their
faulty design solution and unsatisfactory workmanship. Bridges were then designed for
allowable immediate strength disregarding service conditions as well as aggressive
environment.
The unproven design of many prefabricated bridges, poor quality of construction and
materials and mass construction of such structures caused the bad state of repair of many
bridges in Poland. It is estimated that more than 30% of the precast bridges require the urgent
repair or strengthening. To work out methods of their rehabilitation and restore their service
value became an exigency. The presented methods, which have been set up at several
construction sites, seem to be the most effective.
Methods of improving durability of precast concrete bridges
The defects of prefabricated bridges (built in the 60s and 70s of the 20th c.) required,
after a relatively short period of 10-15 years a major repair. Initially, the repair consisted in the
94
reconstruction of the original state, i.e., damaged elements were removed and replaced by new
ones of the same kind. However, this procedure was not effective as it only temporarily
eliminated the effects, but not the main causes of the damage. By the end of the 80s, the way of
modernisation of the precast bridges underwent considerable change consisting in the
elimination of the faults pointed out in until now applied solutions (Siwowski, 1995). The main
changes were as follows:
 making the structure continuous: eliminating expansion joints and changing the static
diagram significantly enhance the durability of the object
 increasing the stiffness of spans and piers: making the structure dynamic and fatigue
resistant
 good waterproofing of the superstructure: stopping water penetration through the concrete
deck slab.
Making the structure continuous aimed mainly at getting rid of faulty expansion joints
as the main reason for prefabricated bridge destruction (Fig.3.17). The first way was to join the
spans in a chain by means of a flexible slab, covering a gap, shaped in the thickened deck slab
of a bridge (buried joint). It formed a continuous beam diagram regarding horizontal forces and
thermal influence, but it was still a chain of simply supported beams regarding live load. This
method did not work properly. Theoretical assumption concerning the behaviour of a flexible
slab allowed its cracking, which really occurred. Consequently, they underwent urgent damage.
Thus, they stopped being applied.
Another method of eliminating expansion joints was performing full continuity of the
deck slab without changing the existing system of bearings. The static diagram of the structure
was changed from a chain of simply supported spans into a row of quasi-frame segments of
several spans. The technologically most complicated solution, necessitating the use of
temporary supports, was to obtain a full continuity of all spans in a continuous beam system or
a frame system. A monolithic joint, shaped above the pier, made it possible to place some extra
reinforcement on negative moments as well as protect and strengthen the ends of precast RC
beams. The change of the static diagram enabled to upgrade the load carrying capacity of the
whole structure and to achieve its considerable stiffening.
Except monolithic joints, the upgrading of span stiffness was achieved by casting, on
the whole deck, a new reinforced concrete deck slab, about 12 cm thick. The slab was made as
composite with the existing precast elements or, possibly, with the old deck slab. It ensured the
cooperation of two slabs in carrying live load, which often increased the load capacity of the
bridge. In the new slab extra reinforcement was embedded, transverse and longitudinal falls
were shaped and new gulleys fastened. Over the gap between the span and the abutment, the
new deck slab was extended behind the back wall using brackets, thus eliminating expansion
joints.
The complex updating of the deck slabs included also additional sealing of the structure.
Additional waterproofing layers were put on, particularly in the curb area. The sidewalk was
modernised by shaping reinforced concrete put-on slabs while all hollow bricks, openings and
extra installations were removed. Besides, leak-proof pavement both on the carriageway and
sidewalk was laid. Described above rehabilitation steps executed on a broad basis and up to
95
proper quality standard restored the right condition of the structure and often upgraded its
carrying load capacity (Siwowski, 1995).
Figure 3.17. Four methods of making a continuous structure of simply supported precast
spans
Case study of precast concrete bridge rehabilitation
The rehabilitated bridge is a four-span structure with supports spaced at 16,3 m and a
total length of 68,6 m, with a curbless deck of 11,0 m service width. Each span is made up of
six precast beams, spaced axially at 2,0 m. The beams support precast deck slabs of 2,0 x 3,0 x
0,12 m, covered with thin 3 cm concrete layer. The whole deck has a one-sided 1,5% crossfall
and a 4,0% longitudinal fall. The bridge had no draining gulleys.
The faulty structure of the spans and severe service conditions caused considerable
damage of the spans, supports and deck equipment. The condition of the main beams varied
regarding the span and the cross-section of the beams. The areas near the supports were the
most deteriorated because of the leaky expansion joints. Corrosion of the main reinforcement
caused cover destruction and occurrence of considerable concrete loss and delamination. As in
the beams, corrosion affected the areas of the precast deck slabs located near the leaky
expansion joints. The blooms, decolourization, scalling and small concrete losses were
96
observed. The other parts of slabs were in good repair and showed no corrosion (Siwowski,
1995).
The caps of piers were the most damaged elements of the bridge structure. The leaky
expansion joints as well as the lack of efficient drainage of the pavement caused extensive and
deep concrete losses, exposure of the corroding rebars, scalling on the whole concrete surface,
etc. Besides, hammerhead piers with such a wide cap form too flexible support for bridge spans.
The bridge flexibility could be easily felt when heavy lorries were crossing the bridge.
The main rehabilitation stages were as follows:
 providing continuity of spans by joining them with piers in a frame (Fig.3.18)
 covering the whole superstructure with a new reinforced concrete slab
 introduction of changes in the cross-section of the bridge
 extending the deck slab beyond the back walls of the abutments.
Figure 3.18. Modernisation of the supporting area of piers: a) before reconstruction; b) after
reconstruction (dimension in cm)
The main goal of making continuity of spans was to upgrade the longitudinal and
transverse rigidity of the structure and to eliminate expansion joints, which meant sealing the
deck. Concrete cast-in-place knot over the piers create a frame structure. A strongly reinforced
knot of considerable size stiffened also the cap of the pier (Fig.3.18). A curbless cross-section
was replaced with a new one with curbs separating pedestrian lanes from those for the traffic.
97
The total service width did not change. The sidewalks were constructed as reinforced concrete
put-on slabs separated from the road pavement with stone curbs. All hollows, pipes and empty
chambers were removed from sidewalk area. The new deck slab made as 10 cm thick RC layer
was composite with the existing one by means of special steel studs. The new deck slab sealed
the deck, increased the load capacity of the span and enabled to properly anchor rebars of the
joints over the piers. To eliminate side expansion joints, the new deck slab was extended beyond
the back walls of the abutments. Besides, end crossbeams were executed which grasped the
ends of the main beams over the bearing seat of the abutment (Siwowski, 1995).
3.6. Case study No.3. Two concrete bridge strengthening
methods – comparison
In this sub-chapter two modern concrete bridge strengthening methods, i.e. CFRP
composites bonding and external prestressing, have been compared. The comparison has been
made for the existing reinforced concrete bridge, build in early 50s, which had to be
strengthened due to heavy traffic increase. The detailed technical design supplemented with
cost analysis for the both solutions was the basis for this case study (Siwowski, 2003).
The bridge is situated along the national road No.19 in Dukla over Wisłoka River. It is
the monolithic RC structure with three-span continuous beams, with the span lengths of 22,0 +
29,0 + 22,0 m. The total length of the bridge is 85,9 m and the total width of the deck before its
rehabilitation was 9,90 m, including 7,0 m carriageway and 2 x 1,25 m pedestrian lanes. The
superstructure is made up of four main beams with the 1,80 m constant depth and altering width
from 0,40 m in midspans to 0,80 m over supports. The main beams are braced with crossbeams
with the section of 0,30 x 1,20 m, situated every 4,0 m along the bridge. On the grid the RC
deck is supported, monolithically casted in place with the rest of superstructure. Two massive
concrete abutments and two concrete bent piers, founded directly on the rock bed by footings,
are the substructure of the bridge. The general view of the existing bridge before rehabilitation
is shown in Figure 3.19, left.
The detailed assessment of the bridge technical condition was carried out to achieve
basic data for rehabilitation design. As a result of this expertise, comprising both material
investigations and analytical calculations, the good technical state of the existing RC structure
had been approved and their upgrading was considered. Sample testing proved the application
of C-25 class concrete. The reinforcement was not menaced due to corrosion, because the
chloride contamination in concrete had not passed the allowable limit. However, the static
analysis showed that for required service load level the calculated stresses due to bending were
higher than ultimate strength of materials, about 18 % for rebar steel and 10 % for concrete in
compression. Additionally, the calculated shear inner forces at the supports were about 30 %
higher than allowable shear resistance of concrete. It meant that the carrying capacity of the
bridge was insufficient with respect to contemporary requirements and the spans needed to be
strengthened (Siwowski, 2003).
98
Figure 3.19. The general view of the bridge before (left) and after (right) rehabilitation
CFRP strengthening proposal
The bridge had to be strengthened according to Polish code for truck with total weight
of 40 tonnes. To fulfil this requirements, the CFRP strengthening with strips and sheets was
proposed, supplemented with the enlargement of main beams cross-sections at the support
compressive zones. The strengthening works comprised the bonding of about 150 m CFRP
strips on the bottom surfaces of beams as bending reinforcement and about 200 m2 CFRP sheets
on the side surfaces of beams as shear reinforcement.
1006
125
28
350
350
IN THE MIDSPAN
AT THE SUPPORT
110
125
4
2%
2%
2%
2%
50
2
2
2
203
80
2
40
3
50
140
3
240
1
240
140
1
240
Figure 3.20. The cross section of the bridge superstructure with proposed elements to be
upgraded: (1) increasing the compressive zones at the support, (2) CFRP sheets, (3) CFRP
strips, (4) additional RC deck (dimension in cm)


Besides, the rehabilitation works also included:
execution of a new RC deck slab, laid on the existing one and connected together, to upgrade
the deck carrying capacity
adaptation of the deck width to contemporary road requirements
99



execution of new modern deck equipment, i.e. waterproofing, pavements, barriers, curbs
geometrical adaptation and strengthening of supports
repair of concrete in existing super- and substructure, i.e. crack injection, hollow fillings,
surface protection.
The proposed rehabilitation design assumed the execution of works under traffic, what
was required by the road administration. The bridge with proposed elements to be upgraded is
shown in Figures 3.20 and 3.21.
7525/2
5
3
5
B
A
A
A
B
1
1875
4
6
2280/2
4
2
2200
2900/2
Figure 3.21. The side view of the bridge with proposed elements to be upgraded: (1) protected
surface of beam concrete, (2) reshaped and upgraded bent of pillar, (3) adapted abutment, (4)
CFRP strips, (5) CFRP sheets, (6) increasing the compressive zones at the support (dimension
in cm)
The concept of bridge strengthening with CFRPs was not accepted by the road
administration, mainly due to “not reliable” and “not approved” CFRP technology at that time
(2000) and high initial cost of materials. These reasons were quite odd for designers, because
they had realised and tested several CFRP strengthening projects before, without any doubts
with regard to the method effectiveness and reliability.
Strengthening with external prestressing
Due to the conservative administration position the next proposal of bridge upgrading
was prepared. The client changed one of the initial assumptions and allowed the rehabilitation
works to be executed under the closure of the bridge. It enabled increasing the strengthening
efficiency assumed in calculation. In this circumstances the external prestressing appeared the
most effective method of strengthening (Siwowski, 2003). Following their acceptance, the
detailed calculations and technical drawings with specifications were presented to the client,
approved and implemented on site.
The external prestressing of RC beams with BBR CONA External system was proposed
(Fig.3.19, right). The tendon alignment was quasi-curvilinear, with linear sections between
deviation points. It was due to necessity of shear strengthening at the support regions of beams.
With regard to low shear carrying capacity of beams, the number of deviation points was
100
comparatively large, i.e. five points in central span and three points in side spans. To ensure the
proper tendon alignment the steel deviators were installed. The tendon anchorages were created
in concrete blocks, casted under the deck plate, behind the support crossbeams. The other
elements of bridge rehabilitation were performed similarly as in the first concept described
above. Thanks to the bridge closure the execution was much easier.
The comparison of two strengthening methods
Both methods described above result in the same structural effect – the increase of
bridge carrying capacity to class B according to the Polish standard, i.e. 40 tonnes. Since the
methods are widely used for bridge strengthening at present and seem to be applied in the future,
it is useful to compare them. The necessity of preparing two technical designs along with cost
assessments for the same bridge created good basis for making the accurate comparison of both
methods (Siwowski, 2003). The data included in Table 3.2 were collected by the author from
the client and contractor of the bridge.
Table 3.2. The cost comparison of strengthening methods (in €)
No.
The partial cost
1 Direct initial cost
(execution of works)
CFRP strengthening
About 100 000 €
(materials and bonding)
2 Indirect initial cost
No cost
(temporary road
(small limitations for traffic on the
closing, environmental bridge)
and social costs)
3 Maintenance cost
No cost
(protection of
(no maintenance action required on
strengthening
CFRP)
elements)
4 TOTAL COST
About 100 000 €
External prestressing
About 65 000 €
(RC anchoring blocks, steel deviators
and prestressing system)
About 15 000 €
(signing and maintain of temporary
detour, repair of rural roads serving for
temporary detour)
About 30 000 €
(corrosion protection of steel deviators
and ducts every 5-10 years)
About 110 000 €
The costs presented in Table 3.2 apply only to structural strengthening and do not
include the rest of rehabilitation work. As it is shown, the total cost of CFRP strengthening is
only due to execution of works. No other costs would be necessary to maintain the added
strengthening elements in the future service life of the bridge. However, the high initial cost to
be invested directly for strengthening works, forced the road administration to look for the other
solution. Similarly, the bridge contractors, when responsible for design and execution of works,
e.g. in D&B tendering procedure, considerably often propose the conventional but initially
cheaper method rather than more expensive CFRP strengthening.
Though the comparison has been made for only one particular case, it has confirmed
that the costs of both strengthening methods, when estimated in terms of service life, seem to
be similar. According to the case study literature, this fact is already recognised worldwide.
Thus, the decision which method to choose for bridge strengthening is based not only on
economical but rather on technical aspects. The comparison of technical elements of both
strengthening technologies is shown in the Table 3.3.
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Table 3.3. The technical comparison of strengthening methods
No.
Subject to compare
1 Technology
2
Durability
3
Aesthetics
4
Strengthening range
5
6
Strengthening areas and
sections
Time of execution
7
Design
8
Unreliable factors
CFRP strengthening
Easy and quick execution, not
complicated technology
High corrosion and fatigue
resistance
Strengthening elements invisible,
bridge aesthetic not changed
Maximum 30 % for individual
elements, passive method, no
stress regulation or crack closure
Suitable for shear (sheets) and
flexural (strips) strengthening
On average two weeks
External prestressing
Skilled labour required, rather
complicated technology
Cable fatigue corrosion and corrosion
of steel couplings and deviators
Strengthening elements visible,
depreciating the bridge aesthetic
Up to 50 % depending on the section
location, active method with stress
regulation, crack resistance increased
Suitable mostly for flexural
strengthening
On average one month, because of the
execution of concrete blocks
Complicated, no domestic
Easy on the basis of domestic
standards available
standards for prestressed concrete
Mechanical destruction, creep and Mechanical destruction, ineffective
ageing of glue
cable injection
This comparison does not answer the question which method is better. And it is
impossible to answer this question in general terms. Strengthening of a bridge is in each
particular case an individual task which requires detailed economical and technical analysis,
based at least on the factors given in the both tables. CFRP strengthening is technologically
easier to execute and maintain, more durable and does not involve any social costs connected
with long-lasting temporary detour. On the other hand, it is initially more expensive and has a
limited range of application. External prestressing is a conventional, even traditional method of
strengthening, with well recognized advantages, such as its effectiveness and the possibility of
stress regulation. Yet, it has also limitations, the most severe of which is the need for bridge
closure and the detour implementation as for the case in Dukla. As usual, in such cases the
optimal solution could lie in between. According to the author’s experience the optimal solution
for RC beam strengthening could be the mixed method: CFRP strengthening limited to shear
areas and external prestressing limited only to flexural strengthening (Siwowski, 2003).
3.7. Case studies No.4. Strengthening concrete bridges with
prestressed CFRP strips
The first Polish on­site bridge application of the CFRP prestressing system was carried
out on a small single span concrete slab bridge in Przesławice over Pokojówka Creek, built in
1948. The length of the span is 5,5 m, the total width is 8,2 m and the concrete slab thickness
is 0,55 m. As the actual carrying capacity of the bridge was far below the road administration
requirements (15 tonnes against required 40 tonnes), it was decided to strengthen the bridge.
102
Twelve strips of 614 HM type transversely spaced with the distance of 0,80 m were used for
strengthening. The strips with the length of 4,72 m were prepared and equipped with steel
anchorages before delivery on site. The M­16 x 215 type anchors were used for bolting each
anchorage to the bottom surface of the concrete slab (Fig.3.22). Before strip stressing the
adhesive was applied on the CFRP strips and the concrete surface. The strips were tensioned
with a hydraulic jack up to 75 kN, required for prestressing (Fig.3.23). The prestressed strips
assembly was followed by bonding three passive CFRP strips, which were used as
complementary transverse reinforcement of the slab. After the CFRP strips installation the
entire bottom surface was covered with some concrete surface protection material (Siwowski
& Żółtowski, 2013).
The next application of the CFRP prestressing system was carried out on the concrete
bridge in Słupiec over the Breń River. Original bridge had load capacity of 15 tonnes which
was not enough for road class and had to be strengthened up to 40 tonnes. Along with being
strengthened, the bridge had to be widened to accommodate modern parapets and crash barriers.
The bridge static diagram is a one­span frame on tall piers with cantilevers. The main span is
21,8 m long with two cantilevers of 7,3 m each. The bridge is curved in plan at the angle of 74 o
which results from the river and road axes. The total bridge width originally was 7,5 m and
needed to become 9,7 m after widening.
Figure 3.22. The final arrangement of additional CFRP reinforcement (dimension in cm)
103
Figure 3.23. Steel anchorages bolted to bottom surface (left) and stretching device (right)
The bridge superstructure is cast-in-situ reinforced concrete with two beams and slab
deck configuration. The concrete class is C25/30 according to core testing results. The main
steel reinforcement of beams is made up of φ40 mm rebars. The following phases of the bridge
strengthening were executed on-site:
 removing all possible dead loads from the bridge (pavement, old parapets, etc.)
 carrying out all necessary concrete repairs (crack injection, holes filling, etc.)
 installation of CFRP strips with Neoxe Prestressing System, i.e. bridge strengthening
 concreting of a new deck slab (widening)
 installation of new insulation and pavement layers
 load testing to prove the strengthening effectiveness.
Figure 3.24. Arrangement of prestressed CFRP strips (top and bottom surface respectively)
(dimension in mm)
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The arrangement of the CFRP strips for strengthening is shown in Figure 3.24. The total
number of 40 strips were used on the top slab surface – to reinforce both beams element for the
negative bending moment in support zone and 20 strips were used in span on the bottom and
side surfaces of each beam to strengthen them for positive moments (Fig.3.25). After the CFRP
strips installation, the entire bottom surface was covered with concrete surface protection
material and the new deck slab was concreted on the existing deck covering the top CFRP
reinforcement.
Figure 3.25. Prestressing the strip on bottom surface (left) and on the top surface (right)
Figure 3.26. Parallel installation of strips in two layers
Due to the large amount of prestressing force required, there was not enough space for
Neoxeplate 614HM strips to be installed in a parallel layout, one next to another. A special
variation of anchorage plates was developed to allow installing one Neoxeplate strip over
another. Firstly, the shortest tape was installed and stressed, followed by the longer tapes, laid
directly on the former ones (Fig.3.26). The layout of bolts on anchorages allowed one tape to
pass over another with minimum level deviation. This deviation was designed to be 4 mm by
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1000 mm inclination – small enough to produce negligible bending stress at CFRP strips.
Therefore Neoxeplate 614HM strips were arranged in stacks, greatly increasing prestressing
force per the element perimeter. The shop-assembled and tailor-made steel anchorages were
designed especially for this application. The special Hilti HST M-16x215 type anchors were
used for bolting each anchorage to the concrete surface. For each steel anchorage four anchors
were used. Before strip stressing the adhesive was applied on the CFRP strips and concrete
surface. The strips were tensioned with a hydraulic jack up to 75 kN.
Recently, for the first time in Poland, the pioneer strengthening of post-tensioned
concrete bridge has been implemented with prestressed CFRP strips anchored with a gradient
method (Kotynia et al, 2013). The road bridge in Szczercowa Wieś over Pilisia River, built in
1965, required reconstruction and strengthening. The existing structure of the bridge was made
up of five post‐tensioned precast concrete I‐girders, 18,4 m in length, supporting a 160-mm
reinforced concrete deck. Bridge reconstruction required widening the structure with two new
post-tensioned girders and strengthening existing girders in flexure and shear. Ten CFRP
prestressed strips and 90 sheet wrappings were applied on the girders. The gradient anchorage
system, developed at EMPA, Switzerland, was used for prestressing. This method, based on the
epoxy resin ability to cure rapidly under high temperatures, provides for a sector-wise heating
followed by a gradual decrease of the initial prestress force towards the strip ends. The complete
application of one strip using the gradient method required less than 4 hours, up to three strips
were applied daily. Due to the use of the gradient anchorage system, all mechanical anchoring
devices can be removed immediately and re‐used at the next strengthened member. No loss of
initial prestressing force in the CFRP laminates was observed during the strengthening process.
The successful application of bridge strengthening with this system was demonstrated on site
for the first time. (Fig.3.27).
Figure 3.27. CFRP strip installation on the PC girder bottom (left) and the bridge after
modernization (right) (courtesy: R.Kotynia)
The use of CFRP laminates (strips, sheets) for strengthening and repair work have
developed significantly in Poland for the last 20 years. Composite materials are becoming
increasingly popular in the bridge industry for strengthening purposes. These CFRP materials
offer several advantages such as resistance to corrosion, a high strength to weight ratio, and
almost unlimited delivery length (in sheet form), thus eliminating the need for joints. However,
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the Polish experience, presented herein, has confirmed the fact that the very high tensile strength
of CFRP composites is not fully utilised when they are applied in a passive form. To improve
the effectiveness of the CFRP composites, the original techniques of their prestressing have
been developed and successfully implemented in Poland.
The Polish experience of the CFRP application for bridge strengthening has shown
several characteristics of the method, namely: it is very effective for structural strengthening
itself and technically relatively simple in the engineering practice, moreover, it can be applied
at short notice. The latter is economically significant and can be considered as a decisive factor
in many situations due to minimising so called social costs. In most cases described above, all
the process of the preparation and location of the CFRP material takes a few hours. In general,
it means that the time of bridge strengthening with CFRP can be performed in several days
only. It should be emphasized that the simplicity of its use and time effective factor of
application of the method compensate a relatively high cost of the material and, therefore, the
method is highly economically effective. Moreover, by prestressing the strip, the ultra-high
tensile strength of the CFRP can be utilized and more advantages are brought to the
strengthening technique (Siwowski & Radomski, 2015).
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4. Maintenance techniques for steel bridges
Steel bridges constitute a large number of the existing bridges worldwide. Corrosion
due to lack of proper maintenance and fatigue sensitive details are major problems in steel
bridges. Main reasons for repair of steel bridges might be also aging, accident damage,
aggressive environmental factors, etc. Damage to steel members typically results from
corrosion, fatigue, and impact. If the damage from any of these causes is extensive, either a
portion or the entire member may have to be replaced. Moreover, many steel bridges are in need
of upgrading to carry larger loads and increasing traffic volumes. Some of the reasons which
might necessitate strengthening of bridges are increases in service loads and intensity, changes
in codes/standards or extra safety requirements.
Two strategies which might be taken when it comes to structurally deficient bridges
include: replacement of a bridge with a new structure and upgrading of the structure to a
required level. The first option is usually very expensive and involves a great deal of traffic
disturbance. The second option, which includes strengthening and repair/retrofit of structurally
damaged or weak elements, is on the other hand a more local approach which targets the points
causing the total deficiency of the structure. The latter approach is of course, more economic
and causes less disturbance in traffic. However, the decision of using which strategy is
dependent on the results of life cycle cost analyses.
Current methods of repairing/retrofitting steel bridges typically utilize steel plates that
are bolted or welded to the structure. However in many cases, welding is not a desired solution
due to old unweldable steel and fatigue problems associated with weld defects. On the other
hand, mechanical details such as bolted connections, which have better fatigue life, are time
consuming and costly. Moreover the lack of chemical compatibility of steel materials used for
repair can cause corrosion problems. Replacing the deteriorated bridge members, particularly
decks, to today’s design standards and durability requirements are more economical alternatives
to replacing the bridges themselves. Reasons such as this provide significant impetus for the
development of bridge decks made of new advanced materials that are durable, light and easy
to install. Lightweight deck replacement is also a recognized strengthening technique for
bridges that have structural limitations on the load carrying capacity, but have nonetheless
sound steel stringers, cross-beams and main girders. Modern repair, replacement, strengthening
and rehabilitation techniques as well as bridge redecking methods are presented in this chapter
with supplementary three case studies.
In one of these case studies the life cycle cost analysis (LCCA) has been used for
selection of rehabilitation method. In the face of growing public scrutiny, officials at
transportation agencies are under increasing obligation to demonstrate their stewardship of
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taxpayer investments in the construction and maintenance of highway infrastructure, including
bridges. Many agencies are investigating economic tools such as life-cycle cost analysis that
will help them choose the most cost-effective alternatives and communicate the value of those
choices to the public. Any transportation agency can use LCCA to determine the design
alternative that will accomplish a project's objectives at the lowest overall cost. By factoring in
all costs over a project's total multiyear life cycle, not just the initial construction investment,
LCCA helps to ensure that an agency can avoid selecting an alternative based solely on the
lowest initial cost. Agencies typically use LCCA to choose among design alternatives that
would deliver the same level of performance during normal operations over the project's life
cycle.
Many state and local highway agencies worldwide have successfully applied LCCA to
analyze options for investments in highway infrastructure, particularly for decisions concerning
the reconstruction, rehabilitation, preservation, and maintenance of bridges. LCCA concepts
are even built into some bridge management systems, and some highway administrations
recently developed a software tool called to support the application of LCCA in bridge design.
Such software system incorporates probabilistic evaluation of multiple variable inputs
including costs, service lives, and economic factors to estimate the likelihood of net present
value (NPV). The exemplary methodology of LCCA presented in this chapter (p.4.6) is based
on the simple and flexible life-cycle cost model consistent with the US standard method for
performing life-cycle costing (Ehlen, 1997). Obviously it can be applied to determine the bridge
maintenance alternative not only for steel bridges as presented in p.4.6.
4.1. Repairs & replacement
Crack repairs
If a crack has been detected (most often in welded connections) and the causes have
been examined thoroughly, retrofit measures must be undertaken to repair the cracked structural
member. If certain details have been evaluated to be weak parts of the structure in terms of
fatigue, the fatigue performance may be increased by applying adequate strengthening
measures. In general the repair and strengthening methods have to consider the cause of the
damage or failure, and reliability-based decisions have to be undertaken to control the repair
and strengthening process for critical details.
The following listing contains the most important repair and strengthening methods for
welded structures:
 stop holes
 removal of crack by grinding
 re-welding
 surface treatments such as TIG (Tungsten Inert Gas) dressing, hammer peening or grinding
 adding steel plates or CFRP strips
 bolted splices using high strength preloaded bolts
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

shape improving
modification of the connection detail.
Table 4.1 proposes in a similar way as to (Kühn et al., 2008) the repair methods to be
applied for repair of fatigue cracks in relation to the crack failure main causes.
Table 4.1. Applicability of repair methods for fatigue cracks in welded structures
Repair and strengthening methods
grinding
re-welding
surface treatments
adding plates
bolted splices
shape improving
stop holes
modification connection
Causes of fatigue cracking
weld defects
G
G
N
G
E
G
N
G
lack of fusion
F
G
G
G
E
E
G
E
cold cracks
F
G
G
F
E
G
G
E
restraint
F
F
G
G
E
G
G
E
vibration
F
F
G
G
F
G
F
E
web gaps
G
F
G
F
N
N
F
E
geometrical changes
F
F
F
G
E
N
F
G
web breathing
N
F
F
F
F
N
G
E
E: Excellent
G: Good
F: Fair
N: Not good
The following listing contains the most common repair and strengthening methods for
riveted and bolted structures:
 stop holes
 strengthening by means of pre-stressed bolts or injection bolts
 adding additional structural members, e.g. filler plates, cover plates or angles
 repair-welding (verification of weldability needed)
 adding CFRP strips
 changing the static system
 restore of the bearing conditions.
In each single case a verification of the efficiency of the chosen method is
recommended. Table 4.2 proposes in a similar way as to (Kühn et al., 2008) the repair methods
to be applied in riveted structures in relation to the crack failure main causes.
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Table 4.2. Applicability of repair methods for fatigue cracks in riveted structures
Repair and strengthening methods
pre-stressed bolts
adding plates
repair-welding
fibre reinforcement
changing static system
stop holes
repair bearing
Causes of fatigue cracking
riveting process
G
G
F/N
G
G
-
-
cracks at holes
E
E
N
F
F
N
-
cracks in gross cross section
N
E
N
G
F
N
-
thin connection plates
E
N
N
F
F
N
-
out-of-plane bending
N
N
N
G
F
E
-
secondary stresses
N
N
N
G
-
G
E
local stress concentration
-
-
-
-
-
F
-
frozen joints
-
-
N
-
-
-
E
poor detailing
F
G
-
G
G/F
-
-
E: Excellent
G: Good
F: Fair
N: Not good
-: not applicable
Figure 4.1. Weld repair in bottom flange (top) and in web (bottom)
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The most common technique that can be used to repair through - thickness cracks is
splice plates (additional cover plates). Splice plates add material to either increase a crosssection or provide continuity at a cracked cross-section. The philosophy of splice plates for
fatigue crack repair is to add cross-sectional area, which in turn reduces stress ranges. For
instance, if a fatigue crack grows across the full depth of the bridge girder, there are two ways
in which it can be repaired. A vee-and-weld can be specified, but the base metal that is weldrepaired will most likely have a shorter fatigue life than the original detail (Fig.4.1). To ensure
the weld repair will have adequate fatigue resistance, splice plates can be added after the repair
is made to decrease the stress range that contributed to the original cracking thus protecting the
repair (Chajes et al., 2004). The example of crack repairing with splice plates is shown in Figure
4.2.
Figure 4.2. Steel girder crack repairing with splice plates: damaged girder (left) and splice
plate repair (right)
Repair of deformed members
Structural members in bridges can be damaged by accident by vehicles or manufacturing
in the factory. Depending on the limitation and codes, the deformed structural member can be
repaired with mechanical and thermal repair methods. The mechanical repair is the most
common repair method and the process is carried out by application of external loads to the
deformed member. The deformed member is loaded in opposite direction to the deformation.
The use of this method results in a loss of the yield strength in the member. There are additional
requirements to national and international codes and recommendations for the mechanical
method, namely:
 mechanical straightening is not permitted below the temperature of -20°C
 maximum external forces should be continuous about 15 minutes in the last step of the
operation
 no cracks and defects caused by mechanical repair are acceptable
 first bending deformations, then torsional deformations should be removed.
The location and size of the deformation and the depth of the plate are decisive in
determining tools and equipment to be used in the repair operation. For example, in case of
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small deformations and thin plates, hand tools are sufficient to use. For large deformations and
members, hydraulic jacks, power winches, chain blocks, pulley blocks, rigging screws etc. can
be used to create external force.
Figure 4.3. Four main heating methods used in thermal straightening process
Figure 4.4. Deformed I girders (left) and their heat straightening in workshop (right)
In thermal (heat) straightening process, damages such as distortion and buckling in
wrought iron and steel members can be repaired. This method is based on using heat (under
control) to the plastically deformed component. Heating and cooling are repeated and the
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deformation is gradually straightened. To heat the deformed material, oxy-acetylene torches are
used. There are different heating methods, including: vee heating, strip heating, line heating
and spot heating (Fig.4.3). The complex heat straightening of deformed I bridge girders has
been described by Sobala et al., 2014 (Fig.4.4).
Replacement of structural members
For individual members, it will be necessary to determine whether the best option is to
repair or replace. In making this decision, cost must be considered along with factors such as
traffic maintenance, convenience to the public, service life of the structure, whether the
rehabilitation is long-term or short-term activity, and the practicality of either option.
A structural member can be replaced due to damage as long as repair process is not
economical and not technically feasible. The main idea behind the replacement process is that
a new structural member is manufactured in the factory and then is installed or assembled in
the existing bridge. In this process galvanic corrosion between the new and old structural
member can occur and therefore this problem should be considered and avoided. Another issue
is jointing, for instance if welding is chosen the weldability of the existing structural member
should be investigated. Welds and high strength friction grip (HSFG) bolts are mostly used to
join the old and new structural member. Cutting the damaged member out of the existing bridge
causes redistribution of the internal forces and consequently the geometry of members in the
bridge can change, therefore this should be investigated and analysed. In such cases, if
necessary, hydraulic jacks, power winches, etc. can be used to support the existing bridge
temporarily. From economic point of view the replacement of a deteriorated member is more
beneficial than its repair since replacement requires less labour (for example see p.4.5). Time
is also a decisive factor in case of bridges with heavy traffic.
4.2. Strengthening & rehabilitation
Addition of steel cover plates
One of the most common procedures applied to strengthen steel bridges is the addition
of steel cover plates to existing members. Steel cover plates, angles, or other sections may be
attached to the girders by means of bolts or welds. The additional steel is normally attached to
the flanges of existing sections as a means of increasing the section modulus, thereby increasing
the flexural capacity of the member. In most cases the member is jacked up during the
strengthening process, relieving dead load stresses on the existing member. The new cover plate
section is then able to accept both live load and dead load stresses when the jacks are removed,
which ensures that less steel will be required in the cover plates. If the bridge is not jacked up,
the cover plate will carry only live load stresses, and more steel will be required.
The technique is widely applicable to steel members whose flexural capacity is
inadequate. Members in this category include steel stringers (both composite and noncomposite), cross-beams, and girders on simply supported or continuous bridges. Cover plating
is most effective on composite steel-concrete members. Typical strengthening of the bottom
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flange of a girder is shown in Figure 4.5. In (a) and (b), horizontal plates, in (c) two vertical
plates and in (d) two triangular plates are added to the member to increase the capacity. Typical
girder strengthening is shown in Figure 4.6. Figure 4.7 shows how the cross-section of a truss
members is increased by welding extra material.
Figure 4.5. Strengthening of the bottom flange area of a steel girder with cover plates
Figure 4.6. Typical girder strengthening with steel plates
There are a number of advantages to the application of steel cover plates as a method of
strengthening existing bridges. This method can be quickly installed and requires little special
equipment and minimal labour and materials. However, in certain instances these advantages
may be offset by the costly problems of traffic control and jacking of the bridge. As a minimum,
the bridge may have to be closed or separate traffic lanes established to relieve any stresses on
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the bridge during strengthening. Still another potential problem if welding is used is that the
existing members may not be compatible with current welding materials.
Figure 4.7. Increasing the truss members cross-sections by welding extra steel plates or shapes
Strengthening by steel plate is a popular method due to its availability, cheapness,
uniform materials properties (isotropic), easy to work, high ductility and high fatigue strength.
However, several disadvantages of steel plate including the transportation, handling and
installation of heavy plates, corrosion of the plates, and limited delivery lengths of plates which
necessitates the work and difficulty of forming joints, the need for massive and expensive false
work to hold plates in position during adhesive cure, and the need to prepare for steel surface
for bonding are very apparent.
Strengthening using CFRP strips
CFRP strips have essentially replaced steel plates as CFRP has none of the previously
noted disadvantages of steel plates. Although CFRP strips are expensive, the procedure has
many advantages: less weight, strengthening can be added to the exact location where increased
strength is required, strengthening system takes minimal space, material has high tensile
strength, no corrosion problems, easy to handle and install, and excellent fatigue properties.
However, there are some limitations regarding this method including: insufficient design codes
for adhesive joints, dependency on workmanship’s skills and a lack of knowledge about the
long term performance. Application of this method in the ambient temperature below 5°C is
not recommended since the curing process stops at this temperature. Strengthening of structures
using CFRP bonding technique is rather new and the success of the operation depends on the
quality of the application and experience of the staff. A number of different approaches have
been investigated to assess the effectiveness of CFRP materials for the repair and strengthening
of steel bridges. Hollaway & Cadie (2002) and Xiao-Ling & Lei (2007) presented state-of-the116
art articles on the retrofit of steel structures using FRP. Recent research has showed that there
is a great potential for CFRP materials to be used in the retrofitting of steel structures.
The year of 2007 witnessed the first Polish application of CFRP strengthening of a steel
bridge. The five-span riveted steel truss bridge in Chełmno over Vistula River with the lengths
of continuous spans of 84,0 + 3 x 96,0 +84,0 m and an upper deck were strengthened with
CFRP strips (Fig.4.8). The steel truss was to be strengthened with a conventional method:
increasing the cross-section of corroded members with steel plates and HSFG bolts replacing
old rivets. It required the insertion of tens of thousands of bolts preceded by the removal of the
same number of rivets. Rivets removal could cause an internal forces distribution and a potential
truss members deformation. The method was very expensive and complicated to implement on
site. Alternatively, a structural reinforcement that did not involve the rivets removal was
proposed - the flexible CFRP strips, narrow enough to be glued between rivet heads on the
upper and lower chords of the main truss girders. The tension zones of both chords were
strengthened with four (upper chord) or two (lower chord) strips of HM 614 type and the lengths
between 15–20 m (Łagoda et al., 2009).
Figure 4.8. Strengthening of upper (left) and lower (right) chords of steel riveted truss with
CFRP strips
Several investigations have discovered that this method can be significantly improved
by prestressing the strips. However, only a few studies have examined the effect of prestressing
on the behaviour of steel structures strengthened with CFRP plates. Recently, Walbridge et al.,
2011, has demonstrated that steel beams can be successfully strengthened or repaired using
prestressed CFRP materials. The results showed a more significant improvement after
reinforcement with the prestressed CFRP laminates compared to the non prestressed. The first
Polish application of this technology for steel bridge strengthening has been more deeply
described in the first case study in this chapter (see p.4.4).
Addition of supplemental supports
Supplemental supports can be added to reduce span length and thereby reduce the
maximum positive moment in a given bridge (Fig.4.9). By changing a single-span bridge to a
continuous, multiple-span bridge, stresses in the bridge can be altered dramatically, thereby
117
improving the maximum live load capacity of the bridge. At the same time, however, a negative
moment is created which must be taken into account. In situations where the added support
cannot be placed at the centre, reductions in positive moments are slightly less. Even though
this method may be quite expensive because of the cost of adding an additional pier(s), it may
still be desirable in certain situations. This method is applicable to most types of girder bridges,
such as steel, concrete, and timber, and has also been used on truss bridges. Each of these types
of bridges has distinct differences.
Figure 4.9. Strengthening by supplemental supports: scheme and structural effect (left) and
examples of application (right)
Depending on the type of bridge, there are various limitations regarding this
strengthening method. First, due to the conditions directly below the existing bridge, there may
not be a suitable location for the pier (e.g. when the bridge to be strengthened passes over a
roadway or railroad tracks). Other constraints, such as soil conditions, the presence of a deep
valley, or stream velocity, could greatly increase the length of the required piles, making the
cost prohibitive.
Modification of simple spans
According to this strengthening method, simply supported adjacent spans are connected
together with a moment and shear-type connection. Once this connection is in place, the simple
spans become one continuous span, which alters the stress distribution (Fig.4.10, left). The
desired decrease in the maximum positive moment, however, is accompanied by the
development of a negative moment over the interior supports.
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Figure 4.10. Strengthening by modification of simple spans: scheme and structural effect
(left) and example of application (right)
This method can be used primarily with steel bridges, although it could also be used on
concrete precast beam bridges (see p.3.5). The girder cross-section and the type of deck used
will obviously dictate construction details. Thus, the main advantage of this procedure is that it
is possible to reduce positive moments (obviously the only moments present in simple spans)
by working over the piers and not near the midspan of the girders. This method also reduces
future maintenance requirements because it eliminates an expansion joints and one set of
bearings at each pier where continuity is provided. The main disadvantage of modifying simple
spans is the negative moment developed over the piers. To provide continuity, regardless of the
type of stringers or deck material, one must design for and provide reinforcement for the new
negative moments and shears. Providing continuity also increases the vertical reactions at the
interior piers. The main design consideration concerns how to ensure full connection (shear and
moment) over the piers.
The bridge in Figure 4.10, right, can serve an example of such strengthening and
modernisation. The existing bridge was made up of four simply supported spans, 24,3 +24,6 +
24,6 + 24,3 m. The spans were made up of four steel castellated beams, on which the wooden
deck was placed. Modernisation and strengthening comprised the installation of two extra HEB
900 steel girders in each span and the execution of reinforced concrete knots (joints) to create
a continuous beam out of the span system. The steel girders were additionally braced with the
steel crossbeams. Simultaneously, a reinforced concrete slab was performed on the bridge,
composite with the existing and the new steel girders. The concrete slab reinforcement in
support sections were designed to carry the negative moments induced by continuity. The
modernisation increased the bridge capacity from 15 to 40 tonnes and widened its deck by 2 m.
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Post-tensioning of girders/stringers
Post-tensioning has been used to repair or strengthen most common steel bridge types.
Most often, post-tensioning has been applied to steel stringers, cross beams and girders, and
case histories for strengthening of steel bridges date back to the 1950s. Typical strengthening
schemes for stringers, cross beams and girders are contained in Figure 3.5 (chapter 3). The
simplest and, with the exception of the king post, the oldest scheme is (a) - a straight, eccentric
tendon. Since early 1950s scheme (a) has been applied to many bridges in Europe, North
America, and other parts of the world. Scheme (a) is the most efficient if the tendon has a
smaller length than that of the member, so that the full post-tensioning negative moment is not
applied to regions with small dead-load moments. The variation on scheme (a) for continuous
spans, scheme (e) has been reported in use for deflection control or strengthening since the late
1970s.
Figure 4.11. Post-tensioning system of steel bridge girder
Figure 4.12. Cross-sections of the strengthened steel superstructure
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One of the most interesting applications of this method was the strengthening of the
bridge over Vistula River in Kiezmark in 2005 (Kaliński et al., 2005). The orthotropic steel
spans of 91,0 +130,0 + 91,0 + 52,0 m in length were strengthened by post-tensioning with
external tendons (Fig.4.11).
Three layouts of tendons were applied (Fig.4.12):
 angle tendons above the supports – post-tensioning in order to diminish the stresses in the
orthotropic deck slab and improve deflection values
 horizontal tendons in the main span – post-tensioning in order to diminish the tensile
stresses in the lower chord of the girders and improve the deflection values
 horizontal tendons above the supports – post-tensioning in order to diminish the tensile
stresses and cracks in the reinforced concrete slab which strengthens the deck.
An additional measures for spans strengthening were the execution of a reinforced
concrete slab on the existing steel orthotropic deck, connected with studs, and the welding of
cover plates on the bottom flanges of girders in positive moment regions.
Figure 4.13 shows external post-tensioning for strengthening of crossbeams of a through
truss bridge by means of high strength bars. The simple steel anchoring blocks were used and
mounted on bottom flange by means of HSFG bolts which replaced several rivets on each end
of the crossbeam. All the crossbeams were post-tensioned in the same manner, which resulted
in the increase of bridge deck capacity.
Figure 4.13. External post-tensioning of steel cross beam with high strength bars: scheme
(top) and bar and its anchorage (bottom) (dimension in mm)
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Post-tensioning of trusses
Post-tensioning as a strengthening method was first applied to steel trusses in the early
1950s. Typical strengthening schemes for trusses are presented in Figure 4.14. Scheme (a),
concentric tendons on individual members, is usually proposed to strengthen the most highly
stressed tension diagonals by post-tensioning (see p.4.5). This scheme tends to be uneconomical
because it requires a large number of anchorages, and very few truss members benefit from the
post-tensioning. Scheme (b), a concentric tendon on a series of members, has been the most
widely used form of post-tensioning for trusses. The polygonal tendon in (c) has been applied
for strengthening purposes in the continuous-span bridge, but it has been used version of (d) in
single span truss. The upper chord of a truss is unable to carry the additional compression force
induced by the post-tensioning, and, therefore, a free-sliding compression strut is sometimes
added to each top chord to take the axial post-tensioning force. Scheme (e), the king post, has
been suggested for new as well as existing trusses.
Figure 4.14. Tendon configurations for post-tensioning trusses (left, see description in the
text) and two examples of the king post strengthening (right)
4.3. Bridge redecking with lightweight deck
One of the more fundamental approaches to increase the live load capacity of a bridge
is to reduce its dead load. Significant reductions in dead load can be obtained by removing an
existing heavier concrete deck and replacing it with a lighter-weight deck. The concept of
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strengthening by dead load reduction was used primarily on steel bridges of almost all types,
however, this technique could also be used on bridges constructed of other materials.
Lightweight deck replacement is a feasible strengthening technique for bridges with
structurally inadequate, but sound, steel stringers or cross beams. If, however, the existing deck
is not in need of replacement or extensive repair, lightweight deck replacement would not be
economically feasible. Lightweight deck replacement can be used conveniently in conjunction
with other strengthening techniques. After an existing deck has been removed, structural
members can readily be strengthened, added, or replaced. Composite action, which is possible
with some lightweight deck types, can further increase the live load carrying capacity of a
deficient bridge.
Modern bridge redecking systems must meet the criteria of ease of construction,
lightweight, economically acceptable schemes, relatively short construction time and long
service life expectancy. Most systems are prefabricated elements (panels) developed to achieve
economy through the repeated use of fabrication facility and to reduce on-site construction time
and labour. Lightweight deck replacement is also a recognized strengthening technique for
bridges that have structural limitations on the load carrying capacity, but have nonetheless
sound steel stringers, crossbeams and main girders. Composite action is possible with most
modern lightweight deck systems and can improve further the live load capacity. The recently
developed redecking systems can be grouped according to material used. The groups are: (1)
conventional materials as concrete, steel and timber and (2) modern advanced materials as:
engineered cement composite, glulam timber, aluminium alloys and FRP composites. The letter
groups of redecking systems is described below in more details.
The contemporary progress of metal engineering, which led to the development of new
generation aluminium alloys with excellent strength and durability, had let to wider utilisation
of this material in civil and transportation engineering (see p.6.1). Particularly effective is the
use of aluminium alloys in bridge redecking (Siwowski, 2009). The removal of deteriorated
heavy RC deck and the replacement with lighter one, engineered with aluminium, allows
avoiding the strengthening of the super- and substructure and thus cuts the total cost of
modernisation. Furthermore, the excellent corrosion resistance of aluminium alloys brings the
saving of cost, spent for maintenance during service life of a bridge, eliminating also during
that time a lot of environmental issues due to painting for corrosion protection. Additionally the
application of aluminium deck shortens the closing time of the bridge, needed for carrying out
the rehabilitation works. It reduces the social costs induced by traffic congestions.
Recognising the potential benefits that aluminium could offer the transportation
industry, since 2005 author has undertaken the research program to develop and implement an
aluminium bridge deck system, which would be feasible and applicable in domestic conditions
(Siwowski, 2008). The aluminium deck panel consists of the hollow extrusions with the crosssection shown in Figure 4.15. On the basis of analysis of the similar deck extrusions, a triangular
one-voided section of profiles with the height of 0,16 m and the width of 0,12 m was accepted.
These dimensions were limited by the recent capability of Polish aluminium extruder, who
could fabricate extrusions with the section inscribed in a circle with the 0,2 m maximum
diameter of piston. After the comprehensive material studies, the 6005A-T6 aluminium alloy
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was chosen to fabricate the extrusions, because of its optimal both the mechanical and
anticorrosive properties.
30o
130
4
3
R6
7
9
3.20
4
167
R6
6
3
67.89o
R4
spawanie MIG
2.10
2
130
30o
Figure 4.15. Cross-section of the aluminium deck extrusion (left) and the deck panel made of
extrusions (right) (dimension in mm)
Figure 4.16. Experimental setup of the panel (left) and load patch simulation (right)
The tests were carried out on the deck panel, 2,10 m wide and 3,2 m long, which was
made up of 16 extrusions welded together with the MIG butt welds (Fig.4.16). The dimensions
of the individual panel were accepted with the assumption of its use in the redecked bridge. The
linear support of the panel on steel beams was arranged in the experiment. The spacing of
supports was about 2,0 m, that suits the most frequently applied spacing of the main girders (or
stringers) in the existing plate-beam bridges. The results of the laboratory tests confirmed the
adequate stiffness and strength of the aluminium bridge deck panel under service load. Results
clearly demonstrate that aluminium bridge deck panels are feasible alternative to RC decks from
the standpoint of strength, serviceability and dynamic response (Siwowski, 2008).
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Although the resulting deck is geometrically orthotropic, the panel is typically oriented
with extrusions parallel to the supporting girders and the direction of traffic. The deck
constructed in this way cooperates very well with the girder (or stringer), creating a composite
system. When installed in this manner, stresses developed under load can be generated by three
different mechanisms, namely (a) longitudinal bending of composite girder - system I stresses;
(b) transverse panel bending - system II stresses; (c) transverse bending of the panel top plate system III stresses. The connections can be made for example by means of a galvanized bolts,
carrying shear forces. The panel is suitable for the application on both an aluminium and steel
girders. In the latter case the cover elements are applied on contact area of both metals in order
to avoid the galvanic corrosion.
Figure 4.17. FRP bridge deck: an pultruded example (left, source: fiberline.com, 12.12.2015)
and on-site application for redecking (right, source: best.umd.edu, 12.12.2105)
Figure 4.18. The cross-section of three arrangements of webs in a FRP deck panel (RUT)
The growing need for the increase in road bridges durability has recently resulted in the
research on new, durable, lightweight and easy to handle bridge decks, made of fibre reinforced
polymers FRP (Fig.4.17). The RUT was also conducted its own research program, in which
three structural solutions of sandwich FRP bridge deck were elaborated, produced and tested
under static and fatigue load. After publications and already existing solutions had been
reviewed, three different arrangements of webs were selected for the numerical analysis and
testing (Fig.4.18). The panel structure comprised outer sheets, inner webs (stiffeners) and a
foam core. Materials engineering was also included in the project, which came down to the
choice of reinforcing fibre number and orientation in the certain elements of the deck panels.
All three deck panels were manufactured with vacuum assisted resin transfer moulding
125
(VARTM) process (see p.6.2). On the base of initial test results, the stiffness, load carrying
capacity and dynamic behaviour of panels were estimated and the best solution for further
research was chosen. The best solution (Fig.4.18, right) has been tested in full scale to estimate
its behaviour under service, ultimate as well as fatigue load. The panel fulfilled the required
criteria for ultimate capacity, serviceability and safety, therefore its application in prototype
bridge redecking is planned.
4.4. Case study No.1. Bridge strengthening with
prestressed CFRP strips
CFRP materials used in strengthening of steel structures provide an excellent solution
for short-term and long-term retrofits, which combines benefits of passive bonded CFRP
laminate systems with the advantages due to external prestressing like improvements in the
serviceability and ultimate strength of the weakened steel structure. The use of prestressed
CFRP strips has been examined in previous studies and found to be a highly effective approach.
Developing a proper and practical prestressing procedure and method for predicting the
carrying capacity enhancement are the main concerns of the current research. The main problem
of strengthening with prestressed strips are the anchor zones. Without mechanical anchorages
there are peeling failures at the ends. Therefore the continuous development of CFRP strips
anchorage systems has been recently observed (see p.3.4). An initial experimental study
investigating the feasibility of new strengthening approach has been conducted at RUT. The
objective of this research was to investigate the prestressed CFRP laminate system for
strengthening of steel bridges, including anchorage capacity checking, estimation of prestress
losses, stress distribution and bending effect of the strengthened girder. The research revealed
the feasibility and efficiency of the new strengthening method. The initial testing results of the
new system revealed that the concept might be used to strengthen steel girders. The ultimate
strength of steel anchorages is as high as 70% the ultimate tensile strength of CFRP strip, which
seems to be quite enough to be applied in reliable strengthening system (Siwowski & Paśko,
2014).
The first on-site application of the system was carried out on a single span composite
steel-concrete bridge (Fig.4.19, left). The span length of the bridge is 15,3 m and total width
8,2 m (after widening, executed simultaneously with strengthening works). The span is made
up of five I NP 550 rolled steel beams and 0,25 m thick concrete deck slab. The cross-section
of existing steel girders was increased before bridge construction with steel cover plates welded
to the bottom flange of I-beam. The general state-of-repair of steel girders after more than 30
years of service was very good, without any considerable corrosion losses or cracks and with
renovated painting system. The material tests performed on concrete core samples taken from
the deck slab revealed about 40,0 MPa concrete compressive strength. Concrete slab surface
shown some small defects and cracks underneath, which were repaired with polymer cement
concrete (PCC) during modernization.
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Fig. 4.19. The bridge (left) and anchorage after installation on the girder’ bottom flange
(right)
Figure 4.20. Details of steel I-girder strengthening with prestressed CFRP strips
(dimension in mm)
127
The actual carrying capacity of the bridge was evaluated and the calculations revealed
that it was only 15 tonnes of a truck weight allowed to pass the bridge, which means generally
the lowest class according to the Polish bridge code. The road administrator’s expectation was
to strengthen the bridge up to 40 tonnes. The concrete deck slab and I -beam top flanges and
webs could carry higher service loads, while the allowable stresses were considerably exceeded
in the bottom flanges of steel beams. Therefore prestressed CFRP strips were proposed as a
strengthening mean, mainly due to high effectiveness of this method, immediate installation
without long traffic interruption and no temporary on-site works required. The CFPR strips
were installed on the girder bottom flanges and prestressed with the system hydraulic jack (see
Fig.3.13). Each steel I-girder was strengthened with one CFRP strip of HM type with 60 × 1,4
mm cross-section. The steel anchorages were bolted to the bottom flanges with six M-20 HSFG
bolts. The details of the steel girder strengthening are shown in Figure 4.20.
CFRP strip tensioning procedure was as follows. The strips with both anchorages
mounted on their ends were prepared in the workshop and tested on the special testing beam
for 1,5 design load. In ready-to-install form the strips were delivered on site. The assembling
started with bolting the passive anchorage, followed by stressing anchorage and reaction block
fixing to the flange. The adhesive mix was applied on the strip and bottom flange surface before
installing the strip. After installing the stretching device the strip was tensioned with the
hydraulic jack up to 75 kN. It was followed by the stressing anchorage bolting in its final
position on the bottom flange. The steel anchorages were protected against corrosion with
relevant paint. The stressing anchorage after installation on the girder is shown in Figure 4.19,
right.
The strengthening process was time-efficient and lasted only one working day on site.
Traffic was not limited during the operation and the bridge carrying capacity was upgraded
straight after the repair works were finished. Meanwhile, all slab concrete surfaces were
repaired and bridge equipment such as drainage, insulation, pavement and guardrails were
replaced. The first application of the new CFRP prestressing system clearly demonstrated that
this strengthening method can be successfully applied to steel bridges in normal site conditions
and without any special temporary works and traffic limitations. The system tested on the actual
steel girder bridge showed more than 60% efficiency of the strengthening method. The
superstructure stiffness enhancement was also observed during static load test carried out on
the bridge after modernization (Siwowski & Paśko, 2014).
4.5. Case study No.2. Truss bridge rehabilitation
The bridge over the Vistula River in Nagnajow is a key component of National Road
No. 9 in Poland. It was built in 1959-1961 as five-span continuous steel Warren truss with RC
deck slab (Fig.4.21). The spans of steel truss are 72,0 + 3 x 90,0 + 72,0 m long. Two truss
girders with the height of 9,0 m and the spacing of 8,40 m are built of welded members with
riveted connections. Upper and lower chords have hollow box section and diagonals have both
box and I-shaped sections depending on compression/tension inner force. The main truss
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girders are transversally stiffened with portal frames located over each support and horizontal
lateral X-bracing in both chords levels. The rolled I-shaped and channel profiles are used for
lateral bracing. The deck is made of RC slab with the thickness of 0,18 – 0,24 m, supported on
steel grid of riveted cross beams with the depth of 1,20 m and three welded stringers with the
depth of 0,90 m. The deck slab which is composite with steel grid is cut every 27,0 m to release
temperature stresses.
The bridge was selected for the comprehensive rehabilitation with a main goal to
increase its load carrying capacity up the highest class according to the Polish bridge code (50
tonnes truck). The assessment of actual bridge condition and evaluation of load carrying
capacity made it possible to establish the scope of required rehabilitation works, which were
necessary to fulfil road administration expectations. The scope of work included (Siwowski &
Żółtowski, 2008):
 repair, strengthening and replacement of relevant members of truss girders
 strengthening of riveted connections
 repair and strengthening of deck grid along with RC deck slab replacement
 execution of new bridge deck equipment.
Figure 4.21. The steel truss bridge before rehabilitation
Truss members strengthening
Two strengthening methods of truss members were applied in the rehabilitation:
member section enlargement with additional welded plates (passive strengthening of upper
chord and compression members) and external prestressing (active strengthening for tension
members). The scheme of member section enlargement as well as strengthening works
execution are shown in Figure 4.22. It was performed with simultaneous members relieving
after RC deck slab demolition in order to ensure the cooperation between the old and new added
parts in carrying dead load of the deck after rehabilitation.
Tension diagonals were strengthened by external prestressing with high strength rods
(Fig.4.23). This work was executed at the final stage of rehabilitation, after the execution of the
new RC slab together with full deck equipment. Active prestressing enabled the full utilisation
of the member capacity also taking into account the losses of prestressing force. The rods of 26
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and 36 mm in diameter were applied along with the system nuts and anchorage plates. The
tailor-made steel anchorages were located under the deck slab (active) and on the upper chord
(passive). The prestresing began in the middle of trusses and was performed towards both ends
of the bridge at a time. The rods were protected against corrosion with special HDPE pipes and
additionally against vandalism with steel tubes (lower parts of rods as high as 2 m above the
deck level). Stressing anchor blocks were equipped with easy removable caps with elastic stuff
in order to adjust prestressing force during the operation.
Figure 4.22. Scheme of member section enlargement (left) and view of strengthening
members (right) (dimension in mm)
Figure 4.23. External prestressing of tension members: bar tensioning (left) and strengthened
members (middle, right)
Truss members replacement
Post tensioning steel bars were selected to relieve tension member during the
replacement procedure. The bars were prestressed up to the value of axial force in the member
in question resulting from dead load. It was assumed that the new RC slab had already been
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made, all cover plates of truss members had been welded, and the crane necessary to perform
the replacement operation was standing on the bridge deck. Because the connection of the new
member was designed with HSFG bolts, some unexpected slip and relaxation was assumed.
Additional safety margin of 50% axial force was applied due to uncertainty of current state of
forces when under construction. The assumption was made that temporary structural
monitoring system would be deployed during the replacement and post tension force would be
adjusted if necessary. Prestressing of the replaced member was realised with four high strength
M-24 bars and special anchor blocks fixed to the upper and lower chord of the truss (Fig.4.24).
After relieving the deformed member was replaced with the new one of similar shape but redimensioned to larger thickness, providing necessary enlargement of member cross-section.
Figure 4.24. Tension member replacement with post tensioning steel bars for relieving
Replacing compression member was much more complicated procedure than replacing
tension member. The difficulties were mainly caused (but not only) by the necessity to relieve
compression member and stabilise the truss geometry during the replacement procedure. Three
different approaches were considered (Siwowski & Żółtowski, 2008):
 replacing the member without relieving
 erection of temporary stiffening structure for relieving member and stabilising its nodes
 relieving the member by means of the temporary support and imposed load.
The first approach was discarded due to excessive internal forces in truss members in
question. While the internal forces were close to the ultimate load carrying capacity of truss
members, the process of cutting the compression member would lead to uncontrolled dynamic
impact to the structure. This would certainly increase forces in members as compared to the
state of its static equilibrium without compression member. The second option required the
design and manufacture of special stabilizing structure, which would act as temporary
reinforcement and stiffening of truss girder. This option was discarded because of complicated
execution and the need for purpose-built structure to be used only once. Finally, the last option
was chosen to build a temporary support under truss girder and to apply member force
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regulation (Fig.4.25). Hydraulic jack was used to push the truss node up to the level, when the
axial force in compression member was completely reduced and thus the member was relieved.
After relieving the deformed member was removed and replaced with new element.
Figure 4.25. Compression member replacement: temporary support under truss girder (left)
and deformed member removal (right)
Replacing a member in steel truss girder was a complicated and dangerous task,
requiring constant monitoring during operation. A temporary structural monitoring system was
deployed to assist in operation of replacement of deformed compression member. The system
consisted of: displacement transducers, strain gauges in key structural members and pressure
sensor in hydraulic jack. All these measurements were completed with geodesy survey. The
continuous monitoring of displacements was set-up in 5 points of each truss girder. Strain
measurements were performed in 3 members, equipped with 4 strain gauges each. The new
member was controlled after its installation, during lowering the jack on the temporary support.
The value of force in hydraulic jack was controlled by monitoring the value of oil pressure. The
displacement of the jack was measured with displacement transducer. The structural monitoring
system allowed for continuous control of replacement operation and on-site decision making.
Comparison of recorded and expected values before and after operation allowed to check the
safety of replacement operation and to verify whether the truss has been repaired effectively.
Deck strengthening and slab replacement
After old deck slab demolition the existing cross beams were strengthened with steel
plates (splices) bolted to lower flanges with HSFG bolts. The special epoxy glue was also used
to create the additional bonding between old and new steel elements. The deck grid
strengthening procedure included also the assembly of three additional stringers in order to
relieve two existing ones. The tailor-made sections of new welded stringers were bolted to
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existing cross beams with HSFG bolts. On the top flanges of cross beams and stringers shear
studs were installed to connect the new RC slab in composite action with steel grid. When the
new steel grid had been executed the concrete slab was cast on the formwork rested on the grid.
Special sequence of slab concreting was established to reduce rheological effects (creep,
shrinkage). Finally, the insulation layer and two pavement courses were placed on the slab
following anticorrosion works undertaken on the whole steelwork. The new steel sidewalk
brackets were also installed out of truss girders with the prefabricated concrete slabs laying on
them.
Strengthening of riveted connections
Strengthening of relevant riveted joints was executed by replacing rivets with HSFG
bolts with additional use of epoxy glue. This hybrid bolted/bonded connections were
extensively tested in Poland and proved to be the best method for riveted joints strengthening
(Łagoda, 1993). The existing riveted connections of cross beams were partially dismantled (in
turn: web, bottom flange and top flange). Due to very dense grid scheme of deck beams no
additional supports were needed during dismantling. The procedure of execution of hybrid
bolted/bonded connections was as follows (Siwowski & Żółtowski, 2008):
 removing old rivets in relevant part of joint
 geometric inventory of existing holes pattern to produce precise new lap plates
 sandblasting of both surfaces
 epoxy glue placement on clean surfaces
 assembly of lap plates in the joint along with temporary bolts mounting
 final HSFG bolt tightening after glue hardening.
Figure 4.26. Procedure of execution of hybrid bolted/bonded connection: temporary
assembling of lap plate (left) and plate after bolt installation and gap welding (right)
Strengthening of truss girder joints was also executed with the same technology as for
the deck grid with additional enlargement of gusset plates together with supplementary bolts
installation (Fig.4.26). During joint strengthening the existing connection was partially
dismantled by removing 50% rivets. The remaining rivets ensured the relevant load carrying
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capacity required for dead and technological load on the bridge. The geometrical adjustment of
new lap plate was then performed following surface cleaning, glue placing, bolt installation and
tightening. In a similar way the second part of truss connection was strengthened. After the
whole joint had been tightened two parts of additional lap plate were welded together.
The proof test results revealed the high effectiveness of all strengthening and repair
methods used on the bridge. Its load carrying capacity and stiffness were considerably increased
and the dynamic behaviour remained very good. The present load carrying capacity fulfils
requirements of the road administration for bridges located on international road network in
Poland. It seems that the bridge will serve the next 60 years without the need for extensive and
costly maintenance works.
4.6. Case study No.3. Rehabilitation method selection
based on LCCA
The use of advanced materials is viewed as a potential long-term solution for steel bridge
strengthening and rehabilitation. But new construction materials for bridges must be selected
with great care and foresight over the conventional construction materials. Some minimum
technical criteria must be first satisfied, such as the material's ultimate strength, stiffness, code
compliance and expected life of the structure under a set of defined environmental conditions.
Currently, for the alternatives that provide the same technical performance, construction costs
are typically used to compare and ultimately decide on the design strategy. But the alternative
with higher initial construction costs, as it usually happens in case of advanced materials, may
have significantly lower operation, maintenance, and repair costs, and therefore life cycle costs.
The life-cycle cost analysis (LCCA) allows the highway agency to determine which alternative
is cost effective over its intended life.
The LCCA methodology used in case study
The simple and flexible life-cycle cost (LCC) model consistent with the standard
method for performing life-cycle costing has been used in this chapter (Ehlen, 1997). It is based
on the LCCA methodology for new-technology materials in construction sector. The
conventional cost categories have been included in the LCC model, i.e.: initial construction
costs, operation, maintenance and repair costs and finally disposal costs. The LCC of each
alternative is computed as the sum of individual project cost items, each cost discounted to
base-year, present-value Euros. Figure 4.27 illustrates how individual project costs can be
classified (Ehlen, 1997). The project LCC is the sum of all project costs. This total is first
divided into agency, user, and third-party (social) costs, representing a ‘‘Level 1’’ classification
of costs by who pays the cost. Each of these Level 1 groups is then divided into construction,
operation, maintenance and repair and disposal costs, representing a ‘‘Level 2’’ classification
according to the period in the life cycle. Finally, each of these life-cycle-period groups is
divided into elemental, non-elemental, and new-material introduction groups, representing a
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‘‘Level 3’’ classification according to which component generates the cost. The classification
also allows designers to compare each material’s cost advantages by comparing these cost
groups across materials.
Figure 4.27. Classification of the project LCC’s
Once the all costs are compiled, the user compares the life-cycle costs of the alternative
materials for bridge strengthening & rehabilitation. The alternative with the lowest life-cycle
cost, all other factors being equal, is the cost-effective material. The user utilizes the cost
classification to compare the technical advantages and disadvantages of each alternative in lifecycle cost terms.
Bridge rehabilitation assumptions
The subject of this study is the bridge described in p.4.5 of this chapter. The service life
assessment of the steel truss revealed that the safe life-cycle period could be estimated for about
60 years, so it was assumed for life-cycle analysis. The scope of work and life-cycle period
defined the project objectives and performance-based requirements. In the next step the
alternatives that satisfied the project objectives and performance requirements had been
identified. The main activity in rehabilitation project was the replacement of deteriorated RC
deck slab along with the strengthening of steel deck beams (stringers and cross beams). Three
deck replacement alternatives were considered in this study: new reinforced concrete deck slab
(base case), steel orthotropic deck and aluminium deck made of extruded shapes (Fig.4.14). All
alternatives satisfied identically minimum performance requirements of the project (load
carrying capacity, serviceability requirements), but they differed about material durability
(which means the service life), unit deck slab weight and relevant scope of steelwork
135
strengthening, and finally type of deck equipment and its durability. The assumed service life
is 30 years for RC deck and 60 years for both metal decks and unit weight of the considered
material alternatives of new deck is 525, 151 and 83 kg/m2, respectively. For each alternative
an individual life-cycle scenarios (or maintenance strategy) had to be established (Siwowski,
2012).
Table 4.3. Life-cycle scenarios for deck replacement alternatives
Life
cycle
time
T
T+15
T+30
T+45
T+60
Deck replacement alternative
RC deck slab
Orthotropic steel deck
Aluminium deck

Existing RC deck slab
demolition

Execution of new
monolithic RC deck slab

Strengthening of existing
steel deck beams (100%)

Strengthening of steel
truss girders (100%)

Anticorrosion protection
of concrete deck slab

Execution of deck
equipment

Pavement and
waterproofing replacement

Concrete deck repair

Anticorrosion protection
renovation

Existing RC deck slab
demolition

Execution of new
monolithic RC deck slab

Anticorrosion protection
of concrete deck slab

Execution of deck
equipment

Pavement and
waterproofing replacement

Concrete deck repair

Anticorrosion protection
renovation

Existing RC deck slab
demolition

Disposal of concrete
debris

Existing RC deck slab
demolition

Execution of orthotropic
steel deck

Strengthening of existing
steel deck beams (30%)

Strengthening of steel
truss girders (50%)

Anticorrosion protection
of steel deck

Execution of deck
equipment

Pavement and
waterproofing replacement

Anticorrosion protection
renovation

Existing RC deck slab
demolition

Execution of aluminium
deck made of extruded shapes

Strengthening of existing
steel deck beams (30%)

Execution of deck
equipment

Pavement and
waterproofing replacement

Steel deck repair

Anticorrosion protection
renovation

Pavement and
waterproofing replacement

Aluminium deck repair

Pavement and
waterproofing replacement

Anticorrosion protection
renovation

Pavement and
waterproofing replacement

Existing steel deck
demolition

Recycling of steel scrap

Existing aluminium deck
demolition

Recycling of aluminium
scrap

Pavement and
waterproofing replacement
The timing of cyclic activities which covers annual maintenance and user costs/activities
during normal operations was assumed according to the Polish road agency guidelines. These
136
activities include annual maintenance works, detailed inspections carried out every 5 years and
partial (not major) repair works carried out every 7 years. The timing of non-recurring future
activities which covers all rehabilitation, restoration, and resurfacing activities was determined
in the most conventional method. The author’s engineering judgment and road agency expert
opinion let to estimate the rehabilitation strategy for each alternative under evaluation. These
life-cycle scenarios for deck replacement alternatives are shown in Table 4.3. Deck material
and equipment durability (except aluminium alloy) was assumed according to the Polish
standards. The performance and durability of aluminium alloy in bridge decks was based on the
comprehensive author’s study (Siwowski, 2008).
Finally, timing of rehabilitation works which covers user delay time during work-zone
operation due to traffic limitations and environmental burdens generated by this delay time had
to be estimated for each alternative. The traffic limitations in work-zone were assumed to take
place in years T, T+15, T+30 and T+45, due to the closure of one bridge lane for pavement and
waterproofing replacement, minor deck repairs and execution of new concrete deck slab along
with its equipment (in base alternative). Timing for work-zone operations is different for each
alternative and strongly depends on deck material and scope of maintenance works. Numbers
of operation days based on contractor’s judgment and road agency expert opinion as assumed
in LCCA are shown in Table 4.4.
Table 4.4. The rehabilitation works timing for deck replacement alternatives
Life
cycle
time
T
T+15
T+30
T+45
Deck replacement alternative
RC deck
Deck: 225 days
Superstructure: 75 days
Total: 300 days
Deck: 60 days
Deck: 225 days
Deck: 60 days
Orthotropic steel deck
Deck: 180 days
Superstructure: 60 days
Total: 240 days
Deck: 90 days
Deck: 120 days
Deck: 90 days
Aluminium deck
Deck: 180 days
Superstructure: 30 days
Total: 210 days
Deck: 30 days
Deck: 60 days
Deck: 60 days
LCCA calculation and results
For all LCC calculations in this study BridgeLCC software was used (Ehlen, 2003). The
software uses a life-cycle costing methodology based on the cost classification scheme showed
in Figure 4.27. The first step of LCC analysis is to identify, classify, and estimate all costs that
occur over the life-cycle. As stated above this total cost can be divided by who pays the cost
into agency, user, and third-party (social) costs, which representing a ‘‘Level 1’’ classification
of costs (Fig.4.27). Agency costs mainly due to construction/rehabilitation works were
estimated on contractor proposals offered in public procurement. In aluminium alternative all
costs estimates were obtained from aluminium producers, fabricators, contractors and recyclers
dealing with this specific construction material. For this alternative also the new technology
introduction costs were estimated and included in LCC analysis. These costs were generated
from testing, evaluating and accepting by agency for use.
137
The user and third party (social) costs data and estimates were obtained according to the
Polish road agency instructions. The user costs to drivers during the works are the sum of driver
delay costs, vehicle operating costs, and costs due to the increased incidence of automobile
accidents. Driver delay costs and vehicle operating costs are based on the additional time that
drivers and vehicles spend in traffic during bridge rehabilitation. Accident costs were neglected
because the rehabilitation works did not affect the trip length. Average daily traffic was
assumed for the Polish national road No.9 in 2010 and traffic volume in T, T+15 T+30 and
T+45 years was forecast according to the national instructions used in feasibility studies for the
road sector. There was no indication of third party costs for this bridge. The bridge is situated
in a remote location, not surrounded by businesses, residences or special environmental zones.
Construction activity is not likely to affect anyone but drivers. However, for research purposes,
the third party costs generated by emission of fuel combustion products were estimated and
assumed in the LCC analysis. Table 4.5 shows the values of these parameters used in LCC
analysis (Siwowski, 2012).
The economic data includes the base year and the length of study period, the currency
to be used, and the inflation and real discount rates. The base year is the first year of the study
period (typically the first year of construction/rehabilitation). It also serves as the year on which
all life-cycle cost (present value) calculations are based. The 2005 year was assumed in the
analysis. The length of the study period is the duration over which all costs are analysed. As
stated above, 60-year study period (which means the whole service life in this case) was
assumed. The analysis was prepared in Polish currency (PLN) and then converted into Euro (€)
with the conversion rate 1€ =4PLN. The current inflation and real discount rates were assumed
at 2% and 6% respectively, according to the national instructions used in feasibility studies for
the road sector. The same economic data were used for all project alternatives.
Table 4.5. Project parameters used in user cost estimates
Item
1
2
3
4
5
6
7
Project Parameter
length of affected roadway
over which cars drive
average daily traffic, measured
in number of cars per day
normal traffic speed
traffic speed during bridge
work activity
number of days of road work
hourly time value of drivers
hourly vehicle operating cost
(average)
Value
1 km
7625 cars
72.3 km/h
43.7 km/h
see Table 4.4
8.58 €/h
0.81 €/km
Each alternative deck’s LCC is the sum of all costs that are incurred over the life of the
bridge, i.e. over 60 years. Table 4.6 shows the computed total LCC for each alternative, with
cost breakdowns by level categories, while Table 4.7 shows the unit costs per 1 m2 of deck area
and total cost savings in comparison to the basic case (RC deck alternative). The aluminium
deck has the lowest LCC (5.917.030 €), making it the cost-effective bridge deck. Looking at its
138
Level 1 breakdown of costs, the total agency cost is more than the RC deck 3.948.490 € versus
3.664.501 €) but it has much lower total user costs to drivers (1.967.129 € versus 3.616.260 €).
However the steel orthotropic deck has a similar qualitative relationship to the concrete deck
and its total LCC is only 6,5% higher than aluminium deck (6.301.168 € versus 5.917.030 €).
Comparing the Level 2 costs for aluminium and concrete decks, the first deck has lower initial
construction costs (4.468.107 € versus 4.785.434 €), lower OM&R costs (1.458.130 € versus
2.486.906 €), and much lower disposal costs (-9.207 € versus 10.946 €).
The comparison of direct agency costs seems to be the most interesting from the point
of view of public procurement, because only the cheapest alternative wins the tender in Poland.
In Figure 4.27 the LCC breakdown by cost bearer due to initial construction works is shown.
As it could be expected, the aluminium alternative is the most expensive one, i.e. 10% higher
than RC and 18% than steel one. Relatively small differences are due to limited scope of
strengthening works generated by aluminium deck replacement option. However, when only
initial construction costs are typically used to compare and ultimately decide on the design
strategy, this alternative would never win the public tender.
Table 4.6. Total LCC’s by deck alternative (€)
Classification level
Level 1
(by entity that
incurs costs)
Level 2
(by life cycle
period)
Level 3
(by elemental
breakdown of
project)
Cost breakdown
Total LCC
Agency
User
Third Part
Construction
Operation,
Maintenance & Repair
Disposal
Elemental
Non-elemental
New technology
introduction
RC deck
7 283 286
3 664 501
3 616 260
2 526
4 785 434
Deck replacement alternative
Steel deck
Aluminium deck
6 301 168
5 917 030
3 672 168
3 948 490
2 626 887
1 967 129
2 113
1 411
3 799 900
4 468 107
2 486 906
10 946
7 238 774
44 513
2 497 838
3 430
6 256 655
44 513
1 458 130
-9 207
5 760 017
44 513
0
0
112 500
Table 4.7. Unit costs of 1 m2 of deck and total cost savings (€)
Parameter
Unit costs of
1 m2 of deck area
Total net
cost saving
Cost saving on
1 m2 of deck area
Deck replacement alternative
Aluminium
RC deck
Steel deck
deck
2 269
1 963
1 843
0
982 119
1 366 257
0
306
426
Figure 4.28 illustrates the total LCC breakdown by cost bearer. The aluminium deck is
the most expensive alternative for agency but the cheapest one for users and third party. The
139
agency bears the additional costs due to high initial price of aluminium products and the new
technology introduction. However, compared to the cheapest alternative, the difference is quite
small – only 7,7%, which seems to be reasonable to pay for minimising the user and third party
costs. The aluminium alternative considerably decreases the user costs in comparison with the
RC and steel alternatives – 84% and 33% respectively. These differences are due to less
maintenance works in OM&R period and short time of traffic limitations for deck replacement
with aluminium alternative. The third party costs are quite small and similar for each alternative
because of the same manner of work zone traffic organization during rehabilitation works. They
are mainly due to combustion emission caused by traffic speed decrease and time of traffic
limitations.
LCC by Cost Bearer
€4,000,000
€3,500,000
Life-cycle cost (€)
€3,000,000
€2,500,000
€2,000,000
€1,500,000
€1,000,000
€500,000
€0
Agency
User
Third Party
RC deck
.
Steel deck
Aluminium deck
Figure 4.28. Total LCC breakdown by cost bearer
Figure 4.29 shows the total LCC breakdown by life cycle period. When considering the
user and third party costs in initial construction period, the aluminium alternative is not the most
expensive one. The higher (more than 7%) cost generates the RC deck, which is the cheapest
one when only agency cost is taken into account. Whereas the lowest cost is incurred by the
aluminium alternative in OM&R period. In this 60-years long period the cost of aluminium
deck is over 70% lower than OM&R costs of RC and steel alternatives. However the high
sensitivity of these LCC on assumed maintenance strategy should be remembered. The total
disposal cost are relatively low for each alternative, but only the aluminium alternative can
generate profit in this life cycle period.
140
LCC by Life-cycle period
€5,000,000
€4,500,000
€4,000,000
Life-cycle cost (€)
€3,500,000
€3,000,000
€2,500,000
€2,000,000
€1,500,000
€1,000,000
€500,000
€0
€-500,000
Init. Constr.
OM&R
Disposal
RC deck
.
Steel deck
Aluminium deck
Figure 4.29. The total LCC breakdown by life cycle period
€7,500,000
€6,500,000
€5,500,000
€4,500,000
57
53
45
37
33
29
25
Steel deck
21
17
13
9
5
1
RC deck
41
€3,500,000
49
Life-cycle cost (€)
Cumulative costs in constant €
Year
Aluminium deck
Figure 4.30. The cash flow of cumulative costs – break-even point
In Figure 4.30 so called break-even point is determined, in which the aluminium
alternative is getting to be the cheapest in present value terms. It is the year T=31 when the
replacing of RC deck and the major repairs of metal decks were assumed. It should be
141
emphasized that for the case of the bridge under consideration the RC deck is always the most
expansive alternative during the considered life cycle period. It is due to high initial user costs
in construction period and the necessity of deck replacing after a half of life cycle period. The
need for replacing is caused by the lowest durability of RC deck assumed according to Polish
standards. Figure 7 illustrates clearly the influence of life cycle period length on LCC results.
In case of 30-years life cycle period economically the most effective alternative is the steel
deck, though the aluminium deck is only a few percent more expensive (Siwowski, 2012).
The results of LCC analysis for three deck replacement alternatives have deterministic
character, because they are based on individually estimated values, unit costs, timing and
economic rates. However, most of these values, times and costs are not known in detail. Some
tools are available to deal with this uncertainty, e.g. breakeven analysis and sensitivity analysis.
Breakeven analysis indicates the maximum or minimum values of key parameters necessary
for an alternative to be cost-effective. Sensitivity analysis measures the effect of key parameters
changes on total LCC.
The goal of the whole life costing assessment was to evaluate the life-cycle costeffectiveness of three bridge deck replacement options. The conclusions reached in this study
are based on the relative performance of the three deck systems. Some key assumptions that
affect the results include a 60-year bridge life for all systems at the beginning, as well as regular
and periodic maintenance works (inspection, repairs, replacement) at different intervals for all
alternatives. Results from these economic LCCA suggest that the aluminium deck system is, in
general, superior to the conventional decks from whole life costing standpoint. The whole life
assessment seems to be the best method to compare advanced construction materials with their
more conventional “competitors”. Despite the simplifications used for the sake of this analysis,
the presented LCCA model could serve as a tool to evaluate bridge strengthening and
rehabilitation alternative from a holistic sustainability perspective – integrating environmental,
social and economic indicators. Such a selection procedure could enhance robust investment
decisions and infrastructure sustainability.
142
5. Maintenance techniques for masonry bridges
Bridges have been important throughout history in connecting cultures, sharing ideas,
and providing the backbone of transportation networks. It is necessary to restore and preserve
these structures for their particular functionality and cultural heritage value. The focus of this
chapter is to present and discuss numerous ways of strengthening and repairing masonry bridges
ranging from minimum intervention to complete reconstruction. Two case studies were selected
to discuss and demonstrate the application and process of various methods.
The development of various strengthening and repairing techniques has been necessary
as historical masonry bridges vary significantly. There are various causes of degradation and
failures, so various needs for upgrading load-bearing capacity must be considered in defining
the conservation work for a particular bridge.
The main structural element of a masonry arch bridge is the arch and thus many
strengthening techniques are based around stabilising and improving the performance of the
arch. When considering the way an arch behaves statically and under loads, several principles
should be considered to design strengthening intervention. These principles reflect the
geometric theories that scientists have experimented with for several centuries. The main idea
is that the geometric form must force the structure to be subjected predominantly to compressive
forces and allow an appropriate path for the line of thrust. Deformations and tensile forces
should be limited. The principles of strengthening masonry arch bridges are presented and
widely discussed by Beuerman (2009).
Interventions may use a variety of materials such as concrete, steel, epoxy resins, soils,
mortars, stones, and bricks. With the introduction of these materials to the historic structure,
compatibility of these materials one with another and with the older materials must be
considered both for the immediate future and for years to come. Compatibility means that the
interaction between the materials and the elements of the structure, whether chemical or
physical, is proper and does not threaten bridge stability and appearance. Incompatibility may
lead to local stresses, alteration of load paths, or over stiffening.
It is the responsibility of an engineer to analyse and determine the best method for
intervention. Before a decision is made, the cause of any and all known deterioration must be
understood. Then, an assessment of the entire bridge’s stability must be made. The assessment
should be made with the assistance of non-destructive or, if appropriate and necessary,
destructive testing to determine material properties. Then, numerical analysis, modelling in
computer software, and other similar methods should be used to determine capacity and stability
of the bridge. The assessment of what effect a repair will have on the behaviour of the existing
143
structure should also be determined. A variety of factors may influence the choice of
strengthening or repairing methods other than just the type of deterioration the bridge has
experienced. When selecting and designing a bridge repair and strengthening, it is necessary to
consider these recommended requirements (Garrity, 2001). Typical terminology for masonry
arch bridges is shown in Figure 5.1.
Figure 5.1. Typical terminology for masonry arch bridges
5.1. Repairing techniques
Grouting
Grouting is used to fill voids in the arch ring or spandrel walls. It is often used to fill
voids caused by ring separation (or cracks) in multi-ring brick arches or between the ring and
backing/fill. Grouting itself does not provide any substantial increase in load capacity, but rather
restores the bridge to a former condition and protects the structure from further deterioration.
The repair is only minor and is usually executed along with some strengthening technique.
It is important to repair masonry with adequate and compatible mortars because the
properties of a mortar determine the durability, compressive strength, flexural and tensile bond
strengths of the masonry. Although hydraulic lime mortars are more common in historic
masonry, non-hydraulic (or fat) limes were also used. To follow the conservation principles and
provide a compatible interaction, it is advised to apply lime mortar rather than artificial
hydraulic cements for the fabrication of masonry repair mortars.
Grouting can be performed relatively quickly and with little or no traffic disruption.
Grout can be applied to easily accessible areas with hand tools. It is recommended to preserve
the aesthetic look of the bridge.
144
Injections
Similar to grouting, injections use grout to fill voids in the fill and backing (above the
arch and in the piers or abutments), deeper than near-surface. The injection fill can increase
load capacity by improving load distribution to the arch and abutments or piers, and by
increasing the weight of the piers or abutments to resist horizontal thrust. It can also be a
preventative measure to slow further deterioration of the structure. Injections will reduce the
amount of water percolation through the structure. The precautions that were discussed in
grouting repair apply for injections as well. For installation, a matrix of holes is drilled into the
structure, flushed with water to clear debris, and then injected with grout starting at the lowest
point and working upwards (Fig.5.2). If injection holes are properly plugged, the repair will
have no negative effects on the appearance.
Figure 5.2. Injection scheme
Replacing brick or stone
Often individual units (brick or stone) or a small section in the masonry will deteriorate
significantly, particularly the edge units which are highly exposed. It is also possible that local
stresses have caused a unit to detach and protrude from the structure. Both cases can reduce the
effective section in the arch ring locally and cause more stress in these locations. To regain the
effective section in the arch ring and help prevent further deterioration, these units should be
replaced. Replacing brick or stone should be executed with compatible units, not only in terms
of material and mechanical properties, but also in colour, size and appearance. Proper selection
of replacement material will improve the appearance and respect the original structure.
145
5.2. Strengthening techniques
Saddling
Saddling involves excavation of the fill and casting of an in-situ concrete arch, which
may be reinforced, on top of the existing arch. The concrete is typically of a weaker strength to
provide better compatibility with the masonry. The technique is often combined with spandrel
wall repairs, or fill and backing repair and it also allows for waterproofing of the structure. The
minimum saddle thickness in which to provide adequate improvement is estimated at 150 mm.
Concrete saddles may be poured monolithically with varying cross-section thickness or may be
poured with uniform thickness (Fig.5.3). The use of fibres in the concrete may provide some
advantages.
Figure 5.3. Concrete saddle of uniform thickness
The new arch formed by the concrete saddle is usually designed to act compositely with
the existing arch. This will increase the effective thickness and improve distribution of loads.
Only nominal reinforcement is likely to be used in this case and some technique of connecting
the new arch to the existing arch is needed, such as ties. The ties ensure proper continuity and
transfer of forces between the saddle and existing structure. They should be installed into the
arch ring, abutment and spandrel walls as needed. The existing abutments sometimes do not
have enough capacity for the addition of a saddle. It is therefore necessary to strengthen the
abutments in conjunction with the saddling technique.
Sprayed concrete
Sprayed concrete is traditionally used to increase the thickness of the arch ring in an
effort to increase load capacity and to stabilise and protect masonry. Sprayed concrete is usually
applied to the existing intrados of the arch ring. In some rare cases, however, the original
intrados ring of masonry is removed and replaced with a sprayed concrete lining to prevent loss
of clearance under the arch.
Sprayed concrete can be applied in three different processes: dry, wet, or composite.
Pre-mixed concrete is sprayed at high velocity and adheres on impact, filling crevices and
146
compacting material already sprayed. Plasticisers are usually added to the mix in order to gain
the right consistency for such application. The concrete is applied in a layer between 150 mm
and 300 mm thick and usually reinforced with a mesh (usually of nominal size steel). The
method reduces the size of the arch opening. It is necessary to include additional abutment
support for the concrete arch by adding to the existing abutments, attaching a pedestal support
to the existing abutment or by cutting into the existing abutments if they are stable (Fig.5.4).
Figure 5.4. Providing additional abutment support for sprayed concrete arches
Prefabricated liners
Prefabricated liners are typically made of corrugated metal or glass reinforced cement
and attached to the intrados of the arch. The space between the liner and arch ring is filled with
concrete or grout. The liners provide an increase in load capacity by supporting the arch and
giving it more resistance. With the concrete between the liner and arch ring the thickness of the
arch is increased, which in turn also increases the load capacity. In addition, when filling the
space between the existing arch and liner, cracks, missing mortar and voids will also be filled
in. A liner must be manufactured for the shape of the arch. The liner is attached to the arch by
supports at the springers or with some kind of bolted anchor system. Prefabricated liners reduce
the headroom beneath the arch and the width of the waterway. The appearance of the bridge is
majorly affected by the procedure and is not preferable for maintaining an authentic appearance.
When corrugated metal is used, even when galvanised or coated, corrosion is possible.
Figure 5.5. Corrugated metal liner
147
Near-surface reinforcement
Near-surface reinforcement used to strengthen masonry bridges is made up of stainless
steel reinforcing bars, which are grouted into pre-drilled holes or pre-sawn grooves in the
exposed near-surface zones of the masonry, where tensile stresses arising from external loads
or settlement effects are likely to result in cracking. The reinforcement helps improve lateral
load distribution and increases the transverse flexural strength of the arch.
Procedures should be followed as for grouting and injections. Once grout has time to
set, transverse holes are then drilled into the arch ring. Stainless steel reinforcing bars are
installed into the holes and then grout is pumped into the holes, encapsulating the reinforcement.
Next, longitudinal grooves are sawn into the intrados of the arch and grout is injected into each
groove. The stainless steel reinforcing is then installed with spacers to provide proper placement
and to ensure that each bar is fully encapsulated with grout. Additional grouting is injected over
the reinforcing. The selection of grouting material that is compatible with the existing structure
is necessary to ensure no increase in local stresses or premature bond failure at the interface of
the grout, masonry and reinforcement.
Another technique which utilises near surface reinforcement in the arches is similar to
an anchoring technique (see below). Retrofitting reinforcement is used to increase the bending
capacity of the arch at critical positions and to stabilize transverse cracking. Steel is placed at
an approximate tangent position to the critical positions in the arch ring. The reinforcement is
usually installed from above using accurately positioned drilled holes through the fill and into
the arch ring (Fig.5.6).
Figure 5.6. Near-surface strengthening scheme
FRP strengthening
Recently many researchers have suggested the use of FRP in the form of surface
reinforcement for masonry structures. The technique has been used in the reinforced concrete
field for more than ten years before researchers began experimenting with it in the masonry
structures (see p. 3.3). The typical way to use FRP in strengthening masonry arches has been to
apply sheets at the intrados and/or extrados. The sheets are continuous across the surface of the
arch to enhance the capacity. The FRP reinforcement will not prevent masonry from cracking,
148
but rather transfers the tension force across the crack, preventing the cracks from opening and
creating plastic hinges. Furthermore, FRP does not alter structural behaviour and is removable.
FRP can be applied in one of three different systems (pre-cured, wet lay-up, or prepreg).
Each system can provide slightly different mechanical and geometric properties, however, wet
lay-up and prepreg systems have been discussed as very similar. Properties should be obtained
from the individual manufacturer to be considered in design. The application of each system is
quite easy and only requires simple bonding to the surface at the pre-defined locations.
FRP strengthening is certainly not as labour-intensive and has longer life than other
strengthening techniques. Particularly when FRP is only applied to the intrados, installation can
be completed very quickly and usually without traffic or service disruption. When application
is required on the extrados, traffic and service disruptions should be planned on.
Anchoring
When significant longitudinal cracking becomes present in the intrados of an arch or
spandrel walls are detached, tilted or bulging from its backing, anchoring is a viable option for
restoring shear transfer and continuity (Fig.5.7). Ring separation occurs more often and a type
of anchored called radial pinning can be used to restore the loss in integrity caused by ring
separation and prevent further separation. Both processes may require replacing or resetting
some units in the intrados or spandrel wall, as well as repointing and/or grouting.
After replacing and resetting, oversized holes are drilled with a rotating drilling device
through the full width of the bridge or through the ring to a pre-defined depth into backing and
fill. After holes are drilled, stainless steel rods encased in a sleeve are placed in the holes and
then grouted under low pressure. The rods are secured to steel anchorage plates at each side of
the arch. To avoid negative visual effects, the plates are usually set in from the face of the stone
and plugged with grout and a cylinder of stone from the borehole (Oliveira & Lourenco, 2004).
Figure 5.7. Strengthening of the masonry arch: (a) adopted anchor scheme and (b) horizontal
anchors
Horizontal anchoring through the arch ring should not affect services; yet, with radial
pinning and spandrel wall anchoring services should be located and appropriate steps taken to
prevent damage. Drilling horizontally through the fill can be difficult. Otherwise, construction
is relatively simple and typically less expensive than excavating the fill and backing to do
149
repairs. Once construction is complete there is little change to the appearance as long as the
anchor plates are capped or plugged.
Relieving slabs
Relieving slabs are flat reinforced concrete slabs placed on top of the fill (Fig.5.8). The
technique improves the bridge through better distribution of loads on the arch and alters the line
of thrust to allow appropriate load transfer to the abutments. In some cases a compressible layer
is installed under the central section of the slab to relieve the arch of more live load. Lateral
pressure on the spandrel walls is reduced and the slab allows for a good waterproofing to keep
water out of the structure.
Relieving slabs are similar to saddles, yet, they are not placed directly on the extrados
of the arch. It is recommended at minimum to remove the current road surface and place the
slab over the fill. On the other hand, if the same height is desired or improvements are needed
in the fill and/or backing for possibly load distribution purposes, fill will need to be excavated
to a pre-defined depth. Excavation should be executed symmetrically on both sides of the
crown. Relieving slabs will be cheaper and easier to construct than saddles.
Figure 5.8. Relieving slab
5.3. Case study No.1. Rehabilitation of 19-century masonry
viaduct
The bridge is situated in the old city of Przemyśl, the border city in eastern Poland.
Similar to the other buildings in the area, the viaduct has been recognized as the part of the 19century heritage. Therefore the city road administration made an attempt to rehabilitate and
revitalize the 130-year-old structure, which survived to our days without any serious
constructional alternations and significant damages. For the last 50 years the brick-stone viaduct
was shielded by modern buildings (stories, garages, pavilion, etc.), hidden and not maintained
and thus its heritage significance has been completely forgotten. Thus the structure underwent
150
progressive deterioration, which recently caused local failures. Thanks to the commitment of
local road administration and the researchers from the RUT, the comprehensive rehabilitation
project has been successfully prepared and implemented (Siwowski & Sobala, 2008).
Description of existing bridge
The viaduct was probably built in years 1860 –1870, when the national railway line
between Cracow and Lvov had been built by the Austrian–Hungarian Empire. The original
constructional arrangement of the viaduct consisted of four parts (Fig.5.9): the central part with
two 6-metre-long brick vaults (A), the symmetrically situated two side parts with three 5-metrelong brick vaults (B1 and B2), the approaching part with three 5-metre-long brick vaults (C),
situated on the east side of the viaduct and two stone retaining walls: eastern (D1) and western
(D2).
15.40
20.47
20.22
20.51
20.50
20.50
14.50
TOR nr 1
20.58
TOR nr 2
17.76
TOR nr 3
D1
C
B1
A
B2
17.78
D2
188.97
Figure 5.9. Scheme of Most Kamienny in the old city centre of Przemyśl
According to historical research study, carried out by the author, the static diagram of
the viaduct during its service life was changed a few times, as shown in Figure 5.10. Probably
during the construction the decision was taken to change the length of the western vaults due to
encountered difficulty with the execution of foundations and also in order to carry a new street
under the structure. The next significant reconstruction of the viaduct took place in 1947, when
the railway line was electrified. In order to adjust the clearance under the central span, the total
removal of two brick vaults was needed. On the rest of its piers the new span with encased
rolled steel beams was constructed. At present the whole structure consists of the six main parts:
the encased beam viaduct over the tracks (A), the eastern brick vaulted trestle (B1 and C), the
eastern stone retaining walls (D1), the western brick vaulted trestle (B2) and the western stone
retaining walls (D2). The total length of the viaduct between edge points of the approaching
walls is 189 m and therefore it is one of the longest masonry viaducts of this kind in Poland.
The viaduct over the railway line has one span with total length of 11,9 m and the
construction height of 0,58 m. The span contains ten I-shaped steel rolled beams 0,43 m,
encased in concrete slab. The total width of the deck is 12,8 m with 8,1 m carriageway, 2 x 1,95
m pedestrian lane and 2 x 0,4 m cornice with steel railings. The span is rested on the stone piers,
which were abandoned during central brick vault demolition in 1947 (Fig.5.11).
The eastern brick vaulted trestle with the total length of 41,0 m consists of two equal
sections with three semicircle brick vaults. The vaults with the length of 5,0 m between its
supports have the width of 13,2 m and the height of 0,6 m. On the vaults there are stone spandrel
walls with the stone cornice. The brick railing walls are located on the cornice with the height
151
of 0,9 m and the width of 0,5 m. The vaults are rested on the stone piers with the width of about
1,5 to 2,5 m. The sand fill is placed on the brick vaults followed with the asphalt insulation
layer. The vault spans have been drained off gravitationally with the decorative stone gargoyles,
located above each pier. The width of road and pedestrian lane is the same as on the central
span, but due to the brick railing walls the total width of the spans increases to 13,2 m.
a)
C
B1
A
B2
b)
c)
Figure 5.10. Alterations of viaduct scheme: a) likely design stage, b) stage after construction
in 1870, c) stage after reconstruction in 1947
The western brick vaulted trestle with total length of 20,9 m, differs slightly from the
eastern one, and consists of one section only with two semicircle brick vaults and the part of
the stone retaining wall. The vaults have the length of 6,0 m between its supports. Other
geometrical parameters of the western trestle are the same as in the eastern structure.
The stone retaining walls along the eastern approach are curvilinear, adapted to the
bridge street widening. The length of the southern wall is 58,1 m and the northern one is 77,0
m long. The walls were made of stonework, topped with the stone cornice. The brick railing
walls were executed on the cornice with the height varying from 0,9 m to 3,0 m and the width
of 0,5 m. The width of the road and pedestrian lanes is the same as on the trestle, but toward
the approaching street the road widens up to 21,5 m. The stone retaining wall along the western
approach is situated only on the northern side, because on the southern side the old public
buildings are situated. The stonework of the wall is similar to the eastern walls. The total length
of the wall is 36,0 m and the width of he road and pedestrian lanes is the same along this part
of the street 2,0 + 8,0 + 2,0 m wide. Due to its poor technical condition the northern wall was
strengthened with two stone buttresses (Siwowski & Sobala, 2008).
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Figure 5.11. The viaduct before rehabilitation in 2000: partially brick up vaults (top left),
viaduct over railway line (top right), stone retaining wall (bottom left), viaduct surroundings
(bottom right)
Evaluation of technical condition
The detailed evaluation of technical condition of the viaduct, completed with the
comprehensive material testing, the geometrical inventory and analytical calculation has been
carried out. On this basis three main groups of damages have been recognised (Fig.5.12). The
first failure group results from the lack of sufficient load carrying capacity of the viaduct. This
group includes longitudinal cracks of brick vaults, inclination and deflection of the walls due
to local instability and extensive loss of brick and stone. Additionally, the 1947 reconstruction
changed the static diagram and induced the cracks in the brick piers of the viaduct.
The second group of damages results from the low durability of originally used
materials, which caused extensive corrosion. The results of chemical, physical and biological
corrosion could be found on the surface of brick and stone in the form of erosion, local leakage,
stain, wet spots and loss of joint material. The corrosion-induced damages have been mainly
caused by bad drainage of the viaduct vaults, i.e. the lack of proper transverse and longitudinal
slopes of pavements, the destroyed and obstruct system of inlets and pipes towards gargoyles
and permanent moist in the vault fill. The wet fill increased the dead load and, along with
thawing-freezing in winter, caused the destruction of the vaults.
The last group of damages was caused by the lack of permanent and correct
maintenance. Additionally, the administration allowed to brick up the vaults which impeded
natural ventilation and drying of walls. Also the big trees, growing next to the viaduct and
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completely shading brickwork, impeded drying processes. Moreover, the repair actions
undertaken in the last 30 years have been only provisional and their results have been not
durable as the improper repair materials were used (Siwowski & Sobala, 2008).
Central span - corrosion of encased
steel beam
Brick arch - vertical cracks and
stone displacement
Brick arch - deterioration of structural material
Brick arch - longitudinal crack
Brick railing walls - deterioration of structural material
Figure 5.12. Typical damages of structural elements
In order to estimate the basic material properties, needed for an analytical evaluation,
the comprehensive material testing was performed. The class of brick and joint mortar was
defined. The compressive strength of sandstone and concrete was also defined. Additionally,
the chemical tests were carried out in order to estimate the chloride and sulphate contamination
and the depth of concrete carbonation in the central span. The tests proved that, in spite of bad
visual condition, most of the viaduct materials have appropriate strength and durability to be
utilised in the viaduct rehabilitation. Another outcome was that the “new” material applied in
the viaduct repairs for the last 30 years was the worst and therefore should be removed. On the
contrary, the old 19-century material was quite good and the decision was taken not to remove
most of it.
The actual carrying capacity of the viaduct was defined on the basis of the detailed static
and strength analysis, taking into account the results of material tests. It was far below the
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lowest load class according to the Polish bridge code. The restrained section of vaults and
middle span of encased beam were the critical parts of the structure. The remaining elements
could carry higher loads but due to the proceeding corrosion process, the carrying capacity
could decrease as well. After the recognition of actual technical condition and load carrying
capacity of the viaduct, the city road administration decided to prepare the rehabilitation design
and to execute comprehensive refurbishment.
Design and execution of rehabilitation work
Following the professional evaluation and research, the rehabilitation design was
prepared in agreement with the city heritage conservator. He recommended the most
insignificant interference in the structure as possible; moreover, suggested limiting
rehabilitation operations to refurbishment only. Fulfilling the conservator's recommendations
and due to extensive scope of needed strengthening works, the administration decided to close
the viaduct for vehicles and create a city passage for pedestrians only. Thus it was possible to
preserve the historical shape and character of the viaduct without any essential changes and also
to prolong its service life.
Figure 5.13. Design visualisation: railings (top left), central span (top right), stone retaining
wall (bottom left) and deck (bottom right)
In the close cooperation with the city architect, the architectural and structural design of
rehabilitation and refurbishment works was prepared (Fig.5.13). The comprehensive
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architectural design comprised also the revitalisation of the area close to the viaduct, which had
been much neglected although located in the city centre.
As far as the encased viaduct over the railway is concerned the total removal of side
brackets and bridge equipment elements was executed. The new RC deck plate with the width
corresponding with the other parts of the vaulted structure was cast as composite with the
existing span (Fig.5.14). It also enabled proper slopes in order to drain the pavement. New
concrete handrails were cast monolithically with the deck. To harmonise them with the rest of
the railings the clinker facing was used. The new insulation layer and granite sett pavement
were placed on the deck. Additionally, the existing concrete structure was cleaned and
prevented by special coating to enhance its durability (Siwowski & Sobala, 2008).
13.30/2=6.65m
Sandstone caps, 2002
Clinker facing of RC
railing, 2002
2.50m
3.50m
Granite pavement, 2002
2%
2%
Central span, 1947
Widening of central span light weight concrete, 2002
5.51
Stone support, 1870
Railway track
Figure 5.14. Typical cross section of central span over railway tracks
(section A acc. to Fig.5.9)
In the vaulted trestles the stone spandrel walls with brick railing walls were replaced
(Fig.5.15). In the rest of brick vaults the comprehensive material repair with loss filling, mortar
joints replacement, surface cleaning, desalination and protection layer placing were carried out.
The crack injection was also needed in the intrados. To relieve the walls cell geocomposite
layer was additionally placed inside the fill just under the pavement structure. The drainage
system was rearranged changing location from lower to upper part of the fill. Similarly to the
central span, the bridge equipment elements (i.e. insulation, pavement, drainage, etc.) were
replaced. To cut away possible drainage leakage towards the fill and to ensure its long durability
the special membrane was installed. Finally, the new lighting was installed on the renovated
deck.
The scope of refurbishment works executed on the stone retaining walls along the
approach was very similar to the ones executed at the brick vaulted trestles. The thorough
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material repair with loss filling, mortar joints replacement, surface cleaning, desalination,
protection layer placing and crack injection was carried out. To relieve the walls cell
geocomposite layer was also used. Drainage and the deck arrangement were executed in the
same manner as for trestles. The whole viaduct after rehabilitation is shown in Figure 5.17.
13.20/2=6.60m
Sandstone caps, 2002
Brick railing wall, 2002
2.50m
3.50m
Granite pavement, 2002
2%
2%
Demolition level
Stone retaining wall, 1870
Ground level
Figure 5.15. Typical cross section of approaches between stone retaining wall
(sections D1 and D2 acc. to Fig.5.9)
Sandstone caps, 2002
13.20/2=6.60m
2.50m
Brick railing wall, 2002
3.50m
Granite pavement, 2002
2%
2%
Demolition level
Brick arch, 1870
Stone support, 1870
Ground level
Figure 5.16. Typical cross-section of brick arch trestle (sections B1, B2 and C acc. to Fig.5.9)
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Figure 5.17. The viaduct “Most Kamienny” after rehabilitation: general view (left) and the
granite pavement (right)
Old masonry bridges, more than 100-years-old, are usually in bad technical condition
and thus appear as the structurally deficient and/or functionally obsolete parts of existing road
infrastructure. Therefore most of them need to be rehabilitated and/or refurbished. On the other
hand, they are mostly heritage structures under the conservator's protection. Thus the methods
of their rehabilitation are atypical, strictly based on the detailed technical state evaluation,
supported with the geometrical inventory and thorough material testing. The lessons learnt from
the various case studies are also very important field of knowledge to be considered in the
rehabilitation project. The case study discussed hereby extends this field and can be used as the
benchmark for this kind of works to be undertaken on similar masonry bridges (Siwowski &
Sobala, 2008).
5.4. Case study No.2. Reconstruction of the brick arch
viaduct
The viaduct was a brick, one-span, vaulted structure built in 1928. The consequence of
railway line electrification was not sufficient vertical clearance, thus the structure was to be
demolished and replaced with the new one, situated at some distance apart. It was to be
approximately 33 m long, made up of three spans with the lengths of 9,0 + 12,0 + 9,0 m. The
intermediate supports were to rest on direct foundation and each end support on bored piles.
The spans were to be prefabricated reinforced concrete girders.
After the inspection and the initial static analysis of the structure, the author found out
that there was no need to demolish the whole brick structure and build a new one on the
condition that the old structure was adapted up to the modern railway line standards improving
the traffic conditions. This conclusion was based on the following facts: horizontal clearance
of the old viaduct was larger than the clearance required for two-track electrified railway line
and the state of repair of the brick structure was good. The suggestion was accepted. The
rebuilding process was presented in Figure 5.18 (Siwowski & Trojnar, 1993).
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First of all, filling was removed and the spandrel walls, vault and the upper parts of the
abutments were taken down as far as 0,5 m below the vault in the skewbacks. Then the
abutments were rebuilt with reinforced concrete in which bridge seats, back walls, side walls
and attached cantilever wings were shaped. Additionally, the wings of the new part of abutment
were strengthened with a crossbeam. The width of the rebuilt parts is the same as the old brick
abutment. Prefabricated girders of prestressed concrete of a length of 18,0 meters, rested on the
rebuilt abutments, support the reinforced concrete deck slab. At the assumed length of span, the
support points are considerably moved out from the face of the old abutment towards
embankment, thus their position is very close to that of the vertical components of strut, which
loads on the foundations of the vaulted structure.
A difficult task was to work out the project for reconstruction and adaptation of the
supports as no relevant information on either their foundations or the loads anticipated in the
original design of the viaduct was preserved. The possibility to use the old supports in the rebuilt
viaduct was assessed on the basis of the following considerations: foundations of the viaduct
were adapted to transfer the load of the strut structure to the subsoil. The dead load with regard
to the aboveground part of the old viaduct can be determinated. The pressure of the old viaduct
on the subsoil applied for several score of years resulted in consolidation of the ground, so an
increase in the load transferred to the ground by 30% above the permissible value was assumed
to be safe.
Figure 5.18. The viaduct after reconstruction: (1) lower outline of demolished brick vault, (2)
adapted parts of brick abutments, (3) new RC parts of abutments, (4) new span of precast
prestressed beams
Before design work began, the aboveground parts of the viaduct were supervised and
their weight calculated. The vertical component of the vault strut developed by the dead load
and transferred to the abutments was determined. The effect of the live load on the strut value
was ignored in favour of safety. Then the contact stresses in the old abutment under the rebuilt
part due to the weight of the new structure, live load and the pressure of the filling on the new
part of the abutment were checked. Considering the intractability of the brick wall supporting
the new part, it was assumed that it would be subject to passive earth pressure. The stresses in
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the wall turned out to be much lower than those permitted for the old brick wall constituting a
part of the old structure. Then there was found distance X from the point at which demolition
work was stopped to the section below which the ratio of the vertical loads after reconstruction
to the previous ones would be less than 1,3. Distance X defines therefore the boundary level
below which the stress increase due to new vertical loads will not exceed 30%, i.e. the value
allowed for the subsoil.
The inequality to calculate the value of X was:
where:
A - vertical load on old abutment, transferred by new structure (including live load)
B - vertical load on abutment in old viaduct, on the section, where demolition work stopped
(excluding live load)
Q - weight of the part of abutment along height X.
The analysis neglected friction and passive earth pressure on the lower part of the
abutment as they would not change after the reconstruction. The value of Q was calculated from
the above inequality X = 2,0 m was obtained dividing the weight Q by the listed area of
horizontal section of the lower part of the abutment. Since the footing of the abutment is situated
much lower than assumed for the calculation, the stresses in the subsoil due to vertical load will
be smaller than the allowed ones, when the rebuilding has been completed.
Then the effect of the moments of forces on stresses was evaluated. It resulted from the
analysis that before the reconstruction the moments increased the edge stresses in the subsoil
on the side of the embankments. The calculation showed that after the reconstruction the
moments of forces on this side would reduce the edge stresses beginning from 3,3 m below the
point where the demolition of the abutment was left off. As the footing of the abutment
foundation is definitely lower than 3,3 m, after the reconstruction the edge stresses will be
smaller than before. Similar reasoning was used to assess the edge stresses in the subsoil on the
side of the railway line. It turned out that after the reconstruction the stresses would not exceed
the former ones. The above analysis made it possible to decide that parts of the supports and
the foundations of the old brick viaduct could safely be used in the reconstruction of the old
structure.
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6. Advanced materials in bridge engineering
Advanced engineering materials (or high-performance materials) are materials that
provide specific performance advantages in comparison with the counterpart conventional
materials. It is often difficult to categorise materials strictly into one group or another group,
but we often define materials as follows:
 standard materials which are used in products that are exposed to noncritical environments
and in low-stress applications
 standard engineering materials which are used in products that must have general bearing
and wear properties
 high - performance materials or advanced engineering materials which are used in products
that must have superior properties (extreme service environments, superior chemical
resistance, corrosion resistance, wear resistance, and loading capacity).
There are several types of materials currently called high-performance materials, e.g.
high-performance concrete (HPC), high-performance steel (HPS), high-performance
composites, high-performance aluminium alloys and high-performance ceramics. They all have
in common outstanding properties compared to the conventional materials. In short time these
materials may not be so high-performance, since the materials science is rapidly developing.
While research laboratories are still exploring the ways to exploit these materials, some of them
are ready to use. As little data is available on the long-term results of many high-performance
materials and as the initial cost is often relatively high, states are reluctant to take on such a
venture independently.
Future trends in bridge design can be classified in four main categories: enhancement
of existing materials (high performance steel, high performance concrete), development of new
materials (fibre reinforced composites, metal matrix composites), new structural association of
materials and structural control. With the development of high-strength and possibly
lightweight materials, bridges will become increasingly more slender and lighter, especially for
long-span bridges and fixed links. Many important fixed links still remain to be planned and
built. Therefore the future will see a wide variety of new and improved structural materials.
Moreover, corrosion has been identified as the major cause of the deterioration in bridge
structures and therefore it is the most important factor responsible for the large majority of
structurally deficient bridges. Improving the life of bridges can be achieved using alternative
high-performance materials, such as fibre-reinforced composites, aluminium alloys or gluelaminated timber. Since composites, aluminium and glulam are lighter than steel and concrete,
do not rust or need painting or protective coatings, and require lower fabrication and erection
time – hence the lower cost – they have distinct advantage over other construction materials.
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Also self-compacting concrete (SCC) is a recent development that shows potential for future
applications. It meets the demands placed by the requirements of speed and quality in concrete
construction. Bridges built using SCC have shown promising results. High strengths and
adequate durability were obtained using SCC. This chapter presents the basic information on
the application of these HP materials in bridge engineering along with the relevant case studies.
6.1. Aluminium alloys
Aluminium has a long record of performance for new or rehabilitated bridges
worldwide. A bridge with aluminium deck was first built in the 1933 in the USA. Since then,
almost 100 bridges have been built and/or rehabilitated using aluminium and this number is still
increasing. The low weight of aluminium enables prefabrication and easy transport of large
assembly elements. In case of bridge reconstruction, aluminium decks can be made and
assembled very quickly and thus the period of bridge closure and traffic limitation can be
minimised. As a result, on-site work is minimal and carried out quickly. If a bridge deck can be
replaced in a matter of days, rather than months, it is claimed that the economic benefit is often
more than the cost. Aluminium also has a higher strength-to-weight ratio than concrete, thus an
aluminium deck weighs usually only 20% of the weight of a concrete deck. This difference
enables a lot of bridges to be additionally strengthened during modernisation, without supports
adaptation costs. The maintenance costs are also lower because anti-corrosion protection is not
necessary.
The achievements of the world aluminium industry in applications in aviation,
astronautics and shipbuilding create huge possibilities of innovation in bridge construction.
Though nowadays the higher cost of aluminium alloys than traditional materials is the barrier
in their implementation in construction industry, the change of this relation into the advantage
of aluminium is a matter of time. Already today, the economic effects of replacing traditional
heavy bridges or bridge decks with the light ones made of aluminium alloys are remunerative.
The numerous benefits of aluminium, particularly its strength, light weight, noncorrosive properties and low maintenance costs, ready fabrication and extrudability, make it a
metal for future consideration in bridge applications. Five main fields of aluminium application
in bridge engineering can be distinguished nowadays: replacement and renovation of
deteriorated bridge decks, extension of existing bridges, movable bridges, floating bridges and
residential area bridges and footbridges (Siwowski, 2006). Another application concerns
military, but the use of aluminium in military structures is out of the scope of this chapter. The
other, most recent examples are described below.
In Sweden, the light aluminium bridge deck system called SapaFront has been
developed to replace deteriorated reinforced concrete decks in composite steel-concrete bridges
(Benson, 1998). The deck is made of multi-voided extrusions connected one to another by
means of tongue and groove to form an orthotropic slab (Fig.6.1). This type of connection
allows for rotation of one extrusion in relation to the next. The ends of extrusions are torsionally
restrained. The triangular hollow profile has been developed with very small unit weight and,
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simultaneously, very high bending and torsion stiffness. There are two extrusion dimensions
available: one having a cross-section of about 0,25 x 0,05 m, with a weight of 50 kg/m2, and
the other having a cross-section of about 0,30 x 0,10 m, with a weight of 70 kg/m2. In most
cases, the extrusions have been mounted with bolts on the deck beams or, in some cases, directly
on the main bridge girders. The deck span between the supporting beams varies from 1,2 to 3,0
m. No welding is used in the system. The deck surface is covered with acrylic-based material
or asphalt. The system has been used for more than 15 years and has been applied to a variety
of bridges.
Figure 6.1. Sweden aluminium bridge deck system called SAPA FRONT: deck panel made of
extrusions (top) and two types of extrusion (bottom)
In the mid 90s in the USA, an aluminium bridge deck, called Alumadeck was developed
(Reynolds, 1997). The deck system was made up of several separate prefabricated deck panels
which were assembled on-site. Each deck panel consisted of a hollow two-voided extrusion
0,30 m wide and 0,20 m high (Fig.6.2, left). The deck panels were fabricated by shop-welding
individual extrusions together at the top and bottom flanges to obtain the designed dimensions.
The deck system was essentially isotropic, which provides significant bending strength in both
the longitudinal and transverse directions. As the deck system is slightly stiffer in the direction
of the extrusions, the deck is normally oriented with the extrusions running parallel to the bridge
girders. Splice plates were used to join the deck panels together along their longitudinal sides.
To achieve composite action between the aluminium deck and the steel girders, and in order to
prevent galvanic action between the two materials, a haunch with nominal thickness of 0,05 m
was constructed on the girders’ top flanges. Shear studs on the girders penetrated into the
interior of the aluminium deck. The bond between the deck and shear studs was achieved by
injecting magnesium phosphate grout into the full length of the extrusions located above each
girder. Another aluminium bridge deck systems made of welded extruded profiles have also
been developed in the Netherlands (Soetens & Van Straalen, 2003) and Japan (Okura, 2003).
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Fig.6.2. American bridge deck called Alumadeck (left, source: alumabridge.com, 12.12.2015)
and the Polish aluminium deck system developed at the RUT (right)
Based on the experience gained from the worldwide aluminium bridge applications, a
research program in Poland has been undertaken to develop and implement an aluminium
bridge system, which would be applicable and feasible in domestic conditions. The main aim
of the study is to design an aluminium bridge deck made of extrusions welded together (Fig.6.1,
right). The several service load, ultimate load and dynamic tests have been carried out on the
prefabricated 2,10 x 3,20 m deck panels, in order to examine and evaluate the panel behaviour
under standard truck load and when loaded to failure. Results from the study clearly
demonstrate that aluminium bridge deck panels are feasible alternative to RC decks from the
standpoint of strength, serviceability and dynamic response. The panel may be recommended
for the use on deteriorated bridges, which need deck replacement (Siwowski, 2009).
Figure 6.3. The Forsmo bridge, Norway
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As a part of a Norwegian research program, the first contemporary all-aluminium road
bridge has been built in 1996 in Forsmo, where the 60-year-old bridge made of steel beams and
reinforced concrete deck was in such a poor condition that it had to be replaced to extend its
capacity (Lundberg et al., 1998). The bridge superstructure, 39 m in length, consists of two box
girders, where the bridge deck functions as the upper flange. The box girder is primarily made
of flat bar extrusions with integrated stiffeners. The same extrusion is used in building up panels
for the web and panels for the bottom flange. The extrusions are placed in longitudinal direction
of the bridge. Each 3,0 metres, the twin box girders have cross-bracing inside the girders and
between the girders to keep the shape of the girders, and to distribute the loads between the two
girders. The bridge deck is made of hollow three-voided extrusions 0,25 m wide and 0,12 m
high. The four webs in each extrusion are set with an angle of about 60o and act as a framework
in the cross direction of the profiles. Extrusions were welded in top and bottom flanges to create
an orthotropic deck. The panels of the deck were welded to the upper flanges of the box girders,
with the extrusions running perpendicular to the bridge girders (Fig.6.3).
Figure 6.4. The Helmond bridge (left) and the Riekerhaven bridge in Amsterdam (right)
Nowadays, one of the wider aluminium bridge applications is the use of this light metal
in construction of movable bridges. Thanks to the use of this lightweight construction material,
the energy consumption during the moving operation is lower. Several movable aluminium
bridges have recently been constructed and opened for service in the Netherlands, for instance
twin bridges in Helmond and Riekerhaven bridge in Amsterdam, all unique bascule bridges,
completely constructed in aluminium (Soetens & Van Straalen, 2003). The bridges in Helmond
were constructed between 1999 and 2000. These are 10-meter-long structures with two lanes
for cars and one cycling lane (Fig.6.4, left). The superstructure was transported in one piece by
ship from the workshop to the construction site. The Riekerhaven bridge in Amsterdam, which
opened in 2003, is a bascule bridge, consisting of two spans of 10 and 13 meters. The
superstructure is made up of extruded trapezoidal profiles for the deck and plates for the main
girders which are 0,90 m high (Fig.6.4, right). The aluminium on the spans is completely
unprotected, only the side surfaces and railings are anodized for aesthetic reasons.
In search of an innovative solution for transportation infrastructure in the Netherlands,
a design for a single-lane floating road for passenger cars with a top speed of eighty kilometres
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per hour has been developed (Brown, 2003). The road consists of rectangular aluminium main
modules measuring 5,3 x 3,5 x 1,0 m (L x W x H). These are interconnected into a road. The
stiff, floating road construction can be used safely and comfortably. The modules can be
transported easily by lorry to any required location. Different configurations make the floating
roads be used in all kinds of environments and situations. The main modules are equipped with
a special edge structure to prevent vehicles from falling into the water. The structure also serves
as a vehicle guidance rail and a protection against collisions and spray water from wave action.
The road is anchored with steel pipe piles. The polystyrene foam filling in the modules make
the modules not sink for a long time if they are damaged. A floating drawbridge can be built
into the road. It can be remote-controlled and has a 10-metre-wide passage. To connect the
floating road with the land, a ramp was designed which is able to move with any fluctuation on
the water level. A pilot stretch of the 70 m floating road was realised near Hedel in mid 2003
(Fig.6.5).
Figure 6.5. The aluminium floating road near Hedel, the Netherlands
(source: roadstothefuture.nl, 12.12.2015)
A good example of a residential area bridge made of aluminium alloys is the
Lockmeadow footbridge in Maidstone, UK (Firth, 1999). The bridge was opened in 1999. The
system was developed specifically for this project and is now being adopted elsewhere. The
bridge deck is a shallow aluminium construction, made up of extrusions laid side by side and
stressed together to form a wide aluminium plank. There are no secondary elements or added
finishes, because the extrusions are designed to fulfil all the functional requirements of the deck,
including the non-slip top surface. The deck system was conceived and developed to satisfy the
following criteria: visual interest and slenderness, maximum component standardization and
repeatability, speed and simplicity of handling and assembly on site, low self-weight, no
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secondary structure, no added finishes and low maintenance requirements. The lightweight
solution also minimised the foundation costs. The dimensions of the deck extrusions are 300
mm in depth and 105 mm in width. Closed-section extrusions of this size are at least twice as
expensive as those with open sections, so an assembly of back-to-back open channel sections
was adopted, leading to formation of a cellular deck cross section. The top flange was ribbed to
create a non-slip surface. However, as it was undesirable to add ribs to the soffit, it was not
possible to use the reversible doubly symmetric shape and make use of “tongue and groove”
type interlocking details. Instead, small continuous slots were introduced in the outstands to
receive shear keys, where necessary, in order to improve transverse rigidity. Additional internal
horizontal flanges were required in the extrusions in order to reduce out-of-plane bending
effects in the webs under the heads of the transverse prestressing bars. Thus, the sections form
a pattern of X-bracing, which improves the transverse bracing and shear behaviour (Fig.6.6).
Figure 6.6. Cross-section of the Lockmeadow footbridge in Maidstone, UK
High-strength aluminium alloys have been the materials of choice for transport
infrastructure for several decades (Siwowski, 2006). The need to enhance structural efficiency
through lower weight and higher durability led to further improvements in aluminium alloys,
many of which were achieved via new manufacturing processes such as double aging, reversion
aging, and controlled combinations of heating and mechanical deformation between quenching
and aging. These practices, along with tight controls on alloy composition, have increased the
strength, durability, and corrosion resistance of aluminium alloys. Today’s aluminium alloys
are 1,5 times stronger than the early alloys used for bridges and, considering inflation, have
essentially the same cost.
Several new aluminium structural materials have been developed for use in recent
decades. These new materials are the result of recent innovations in powder-metallurgy
technology and materials science and engineering. New production processes have been
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developed to compact powdered aluminium into billets which can be extruded, forged, or rolled
into aluminium mill products. These new aluminium powder metals are also being
mechanically alloyed and reinforced with whiskers and fibres to produce very unique structural
properties. Generally, the new materials are superior in quality and corrosion resistance to
conventional ingot aluminium mill products. Fibre reinforced aluminium materials which are
stronger and stiffer than steel are produced, and yet weigh only 1/3 as much. Potential structural
applications for these new materials are being defined and evaluated in structures. A major
revolution could be started by either a major cost-reducing production breakthrough, or the
advent of larger production volume applications for one or more of these new aluminium
materials. This revolution may also include the design of more hybrid aluminium structures in
which a mix of these different aluminium material systems is used to maximize the structural
properties and minimize the total life cycle costs. All of these imminent structural materials
engineering developments will have a major impact on the design structures in XXI century.
Building and construction have been relatively constant for a number of years, but the
need for significant attention to rehabilitation and rebuilding of the world’s transport
infrastructure may change that. Opportunities to replace deteriorated bridges with aluminium
decks and/or girders without strengthening the foundations and piers could generate a
significant market for aluminium plates and extrusions, as more and more bridges worldwide
are reported to be in serious to urgent need of replacement. Aluminium allows highway
agencies to increase the life of existing bridges, saving billions of dollars by replacing and
upgrading their deck systems. And the replacements can be made in days rather than months,
greatly reducing the downtime and disruption of bridges and roads for motorists. Aluminium
will continue to be an important component of bridge structures. Growth in aluminium usage
seems assured by its high value to fabricators and consumers. Briefly summarized, the
potentially fruitful fields of effort fall into three major categories evolving from the three
industry sectors: raw materials production technology, manufacturing technology and
application technology. The following applications are primary:
 bridges and bridge decks, by retrofitting deteriorated concrete and steel bridge decks with
aluminium, to increase load carrying capacity and reduce future maintenance cost while
preserving foundations and girder structures
 aluminium in earthquake-resistant structures, because of the effect of reduced inertial forces
on structures in earthquakes, notably for elevated roadways, bridges, and high-rise
buildings.
While alloys have been developed for specific market (automotive, aircraft, marine,
etc.), wider application may be found. Many of these materials possess relatively high strength,
exceptional corrosion resistance and are readily weldable. Many would find application to the
building of the new infrastructure. New alloys, materials, and processing technologies are being
used to produce better components with significantly lower life-cycle costs. For example, the
aluminium-lithium alloy 2097 has 3 times the fatigue life, 5 percent lower density, and 7 percent
higher stiffness than the original material, aluminium alloy 2024, used in early aluminium
bridges. Since the replacement alloy is more fatigue resistant, it decreases the frequency and
cost of downtime for element replacement. If the conventional aluminium-alloy elements fail,
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the material is being replaced with a metal-matrix composite (MMC), for example an
aluminium alloy reinforced with silicon-carbide or aluminium oxide particles (discontinuously
reinforced aluminium - DRA), which is about 50 percent more stiff than monolithic aluminium
alloy (Starke & Williams, 1999).
Many bridge engineers incorrectly believe that aluminium alloys are not sufficiently
robust to survive the rigours service requirements for highway bridge components. Most of the
above presented examples clearly demonstrate that aluminium alloys can provide safe and
durable bridges with minimal life-cycle costs. Today aluminium alloys for bridge applications
are available in a broad range of strengths, ductility and fracture resistance and are fully capable
of meeting the most demanding requirements. Aluminium can be readily extruded to various
configurations to suit the needs of countless possible structural applications. Combining this
versatility with its light weight, corrosion resistance, ease of fabrication and erection and low
life cycle cost, makes aluminium an ideal material for bridge applications.
Future prospects for utilisation of aluminium in the infrastructure will increase, given
the physical characteristics, flexibility of product forms and relative ease of fabrication. Efforts
to reduce first costs should continue, and result in more competitive initial cost. Aluminium
alternatives should be examined more favourably as life cycle cost methodologies become
accepted and mature (Siwowski, 2006).
6.2. FRP composites
Since 20 years the lightweight and high strength fibre reinforced polymer (FRP)
composites have extend the life of critical infrastructure throughout the world, while
minimising ongoing maintenance costs. First used in the 1950s for ships, FRP composites have
undergone ongoing development to become highly valued in aerospace and military
applications for their lightweight and high strength properties. Civil infrastructure applications
began in the 1980s with the research on carbon fibre reinforced polymer (CFRP) strengthening
of bridges, buildings and other structures, and the design and construction of the first FRP
pedestrian bridge in China (Shrivastava, 2009). More recently, pedestrian and road bridges,
bridge enclosures and other structural applications have been undertaken using FRP composite
components.
FRP is a composite material made of a polymer matrix reinforced with fibres. The fibres
are usually glass, carbon or aramid and have many advantages over other more traditional
construction materials due to their high strength-to-weight ratio, ability to be moulded into
various shapes and resistance to environmental conditions that therefore result in low
maintenance costs. Structural applications of fibre reinforced polymer composites have recently
been attractive in the civil engineering community due to their superior material properties such
as high specific stiffness (carbon fibre composites), high specific strength, and high corrosion
resistance. However, all-composite structural systems have specific shortcomings such as high
initial costs, low stiffness (glass fibre composites) and existence of catastrophic failure modes.
To make the best use of materials and to overcome their shortcomings, the combinations of
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FRP and conventional materials have recently been investigated by a number of researchers
(Zogh, 2014). According to them, the most effective use of FRP composites in civil structural
applications is in the form of a hybrid construction with concrete. The review of various FRPconcrete composite beam/deck systems and their construction applications in civil engineering
is presented, and several problems regarding the future research are discussed e.g. in (Hejll et
al., 2005) and (Yang, 2014).
The use of FRP materials for primary structural components in bridge construction has
developed over approximately 20 years. This development has been driven by the need for swift
complete or partial reconstruction of existing bridges to minimise disruption to the transport
network and communities, in conjunction with the generally acknowledged durability of FRP
materials. The development of FRP bridges has focussed on a small number of basic concepts:
large custom-made vacuum-cured elements, multiple-bonded/bolted small pultruded elements,
and large pultruded bonded/bolted elements. Each of these concepts has various advantages and
disadvantages in terms of design freedom and installed cost, summarised in Table 6.1.
Table 6.1. Advantages and disadvantages of different FRP bridge technology
FRP bridge type
Multiple-bonded
small pultruded
elements
Multiple-bonded
large pultruded
elements
Advantages





Vacuum-cured
elements


Disadvantages
Variable structural forms
Cost effective due to module
repetition

Cost effective due to module
repetition
Cost effective due to minimum
number of bonds
Reduced risk due to minimum
number of bonds
Maximum variability of
structural form and laminate fibre
architecture
Reduced risk due to minimum
number of bonds





Labour intensive due to large number
of bonds
Increased risk due to large number of
adhesive bonds
Structural forms limited
Few different types of element
available
Specific compliance tests required for
bespoke elements
Initial cost of mould/die
In terms of overall material cost of a bridge, an extensive analysis, although often
required, will usually have a relatively small impact on the installed cost. The most effective
method to reduce the installed cost of FRP bridges is to investigate the most appropriate design
concept that gives a simple fabrication method and installation procedure. This procedure has
been applied for the first Polish FRP road bridge, which was the final result of the research
project carried out in 2013-2016 by the scientific consortium led by Mostostal Warszawa SA.
The main goal of this research was to develop and demonstrate the FRP road bridge (combridge.pl). The innovative hybrid idea of a FRP-concrete composite structural system for bridge
girders was proposed. The bridge, material selection, VARTM manufacturing technique used
for girders production and some results of the initial research on the hybrid box girder have
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been briefly presented below. The output of the research project gave a very promising future
for the FRP composite bridge application in Poland (Siwowski et al. 2015).
The bridge deck with a span of 22,0 m and a width of 10,4 m, is formed with four simply
supported FRP girders with an overlying lightweight concrete slab, 0,18 m thick, reinforced
longitudinally and transversally with glass fibre reinforced polymer (GFRP) bars (Fig.6.7, 6.8).
The FRP girder has a U-shape cross-section with a maximum width of 1,55 m and a depth of
1,0 m. The top flanges are 0,35 m width each, and the bottom flange is 0,73 m wide. The top
flanges and the webs have thickness of 23 mm, while the bottom flange is 19 mm thick. The
top flange is made of GFRP but the bottom one has mixed CFRP/GFRP structure. The web is
made as a sandwich panel with a foam layer, 15 mm thick, between two glass fabrics. To
increase the torsional stiffness of the girder and prevent shear-bending buckling of its webs, six
internal diaphragms are placed along the length of the girder. The diaphragms are built as a
sandwich panels with the same structure as the webs. The sandwich panel is also adhesively
connected to the top flanges of the girder to be used as a formwork during the concrete slab
casting. The top reinforced concrete slab is connected to the FRP girder through galvanized
steel shear connectors. The support zones of the FRP girder are filled with lightweight concrete
to create support crossbeams and to ensure transverse stiffens of the span.
Figure 6.7. General view of the bridge
The FRP laminates, which form the walls of the box girder, have vinyl ester matrix and
hybrid glass-carbon reinforcement. Carbon and E-glass in the form of a woven fabrics were
chosen as the reinforcement of FRP. The advantage of vinyl ester resins, which were chosen a
material for the matrix, is that it combines the positive aspects of epoxy and polyester. The
lower FRP flange has got the hybrid structure and comprises 28 layers in total (17 layers of
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carbon fibers with 0̊ orientation and 11 layers of E-glass fibers with ± 45̊ orientation). The top
FRP flange has got 32 layers of E­glass fibers with 0 ̊ and ± 45 ̊ orientation. The web has got 9
layers of E­glass fiber with ± 45̊ orientation and foam with a thickness 15 mm.
Figure 6.8. Cross-section of the bridge deck
Table 6.2. The material constants
No.
1
2
3
The
type of
fabric
Glass
Glass
Carbon
Fibre
orientation
Fabric
weight
Nominal
thickness
of
laminate
[deg]
[kN/m2]
[mm]
± 45
0
0
1200
1200
600
0,8
0,80
0,60
Modulus of laminates
in the longitudinal /
transverse directions
Ex
[GPa]
12,06
42,13
115,76
Poisson’s ratio
Ey
[GPa]
12,06
10,87
5,72
νxy
[-]
0,49
0,29
0,41
νyx
[-]
0,49
0,075
0,021
Shear
modulus
Gxy = Gyx
[GPa]
3,04
2,71
3,32
Table 6.3. The material strength
Normal strength
No.
1
2
3
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Layer
Glass
1200 g/m2
Glass
1200 g/m2
Carbon
600 g/m2
Fiber
orientation
Along
the fibers
Across
the fibers
Shear strength
of the laminate
plane XY
[deg]
Xt
[MPa]
Xc
[MPa]
Yt
[MPa]
Yc
[MPa]
Sxy
[MPa]
± 45
520,0
320,0
520,0
320,0
60,0
0
855,0
537,0
44,0
84,0
25,0
0
1150,0
464,0
12,0
94,0
54,0
The mechanical properties of these laminates were determined with the aid of the
classical laminate theory, starting from the experimental characterisation of the unidirectional
lamina and the selected laminate stacking sequences (Barbero, 2011). To determine the
mechanical properties of a laminate tensile tests were performed using universal testing
machine. The results are presented in Table 6.2 and 6.3.
The girders were manufactured in the VARTM infusion process. Such a decision was
taken because this process, among the others used to manufacture FRP elements, is the most
indicated to produce large dimension elements (Siwowski et al. 2014). In fact, it is the process
commonly employed to produce FRP boat hulls. According to this process, once the steel mould
is properly prepared, the dry fabrics are piled up on it according to the stacking sequence defined
in the design process. The dry fabrics are then covered with a plastic sheet that is sealed against
the steel mould, forming what is known as a vacuum bag (Fig.6.9). The vacuum bag is provided
with two series of valves, one connected to a vacuum pump and the other connected to a resin
deposit. Once the vacuum is created in the bag, the resin valve is opened and the resin is sucked
in the bag, impregnating the dry fabrics. The process must be properly designed so that the resin
impregnates the fabrics uniformly, without leaving dry spots. The resin infusion allows placing
all the dry fabrics in the mould and impregnating them in one shot, so the cost of the FRP girders
is reduced compared to other manufacturing processes.
After the dry fabrics get properly impregnated, the mould is covered with rock-wool
blankets and the laminates are cured at 80ºC for 9 hours, with hot air blowing under the blankets
generated by industrial hot air generators. This curing process is recommended by the resin
supplier to obtain a glass transition temperature Tg of 77 - 80ºC. After being cured, the girders
are extracted from the mould and their edges trimmed to regularize them. A series of sandwich
panels are then glued on top of the webs of the girders, to work as stay-in-place formwork
during the casting process of the top slab. The infusion process of the girder is shown in Figure
6.10.
Figure 6.9. VARTM – vacuum assisted resin transfer moulding process scheme
The main objective of prototype girder testing was to evaluate its behaviour under static
load and to determine its real carrying capacity as well as modes of failure. The first step was
to check the girder performance under the standard service load according to the Polish code
for road bridges. The finite element method (FEM) model was validated for design purposes at
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this stage. The initial static tests were followed by a dynamic modal test of the girder. The main
results of these tests allowed to establish the main dynamic characteristics of the prototype
girder, i.e.: self-frequencies and corresponding modes, logarithmic decrement of damping and
the level of vertical accelerations. Finally, the girder was loaded with fatigue loading followed
by quasi-static loading until failure. The ultimate carrying capacity and modes of failure was
the most interested output expected at this stage of research. The output of the static test is
briefly presented below.
Figure 6.10. VARTM process of the FRP girder (courtesy: Mostostal Warszawa)
Figure 6.11. The FRP girder under testing in the laboratory
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The comprehensive laboratory test of the girder model took place at the RUT structural
laboratory (Fig.6.11).The 46 electrical resistance gauges were used for strain measurement: 11
strain gauges on the bottom flange, 6 on the top flange, 2 square rosettes on the web, 4 strain
gauges on the shear studs and 13 gauges on the concrete slab deck. Displacements were
recorded with 18 transducers along the length of the girder and at bearing pad location. Figure
6.12 shows the location of the strain gauges and displacement transducers in the middle of the
girder, on the deck slab and the bottom flange. Calibrated load cells were used to monitor the
load. Load, strain, and displacement data were recorded every second during the test with a
digital data acquisition system.
Figure 6.12. Instrumentation layout for the girder in the mid span (P – displacement
transducers, T – strain gauges)
The girder was tested in four-point bending using a 2 x 630 kN hydraulic actuators
mounted on a steel frame (Fig.6.13). The load was distributed on the contact surface 2,50 ×
0,30 m of the deck slab. The girder was loaded in five stages with several increasing load levels,
which corresponded to design loading. The fifth, final stage, was to load the girder until testing
machine limit equals 1260 kN.
Figure 6.13. Loading scheme
Figure 6.14 shows the displacement plot in the measuring point located on the deck slab
in the middle of the girder for the final load cycle (0 – 1260 kN). Figures 6.15 and 6.16 show
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the strain plots in the measuring points located on the deck slab and bottom flange in the middle
span of the girder for the final load cycle (0 – 1260 kN).
1400,0
1200,0
Load [kN]
1000,0
800,0
600,0
400,0
200,0
P7/3
0,0
0,0
20,0
40,0
60,0
80,0
100,0
120,0
140,0
160,0
180,0
200,0
Displacement [mm]
Figure 6.14. Mid-span displacement (mm)
1400,0
T10/1
T10/3
T10/5
1200,0
800,0
600,0
Load [kN]
1000,0
400,0
200,0
0,0
-1000,0
-900,0
-800,0
-700,0
-600,0
-500,0
-400,0
-300,0
-200,0
-100,0
0,0
Strain [μm/m]
Figure 6.15. Strain plots in the deck slab in mid-span
The girder behaved linearly until the applied total load of 1260 kN (2 × 630.0 kN) and
no residual displacement was observed after unloading. Under the maximum load the strain of
the bottom flange was equal 5,2‰, which corresponds to the tension stress of 603 MPa in the
carbon laminate and 219,5 MPa in glass laminate oriented 0 ̊ in the direction of the longitudinal
axis of the girder. The maximum stress of the carbon laminate was 52,6%, while the glass
laminate stress was 25,7% of the characteristic strength of individual laminates according to
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Table 6.2. The concrete deck slab was not crushed under the load of 1260 kN. It should be noted
that the girder reached the load-bearing capacity at the maximum range of actuators testing
machine without undergoing the global destruction. The maximum girder’s capacity established
in the test (in terms of bending moment) was Mn = 5922,0 kNm, which is 323% of the
characteristic bending moment for which the girder was designed.
1400,0
T7/1
T7/2
T7/3
1200,0
Load [kN]
1000,0
800,0
600,0
400,0
200,0
0,0
0,0
1000,0
2000,0
3000,0
4000,0
5000,0
6000,0
Strain [μm/m]
Figure 6.16. Strain plots in the bottom flange in mid-span
However, the girder under the maximum load of 1260,0 kN suffered several small local
damages in the form of:
 local inner and outer delamination in both flanges and within the transition zone between
web and upper flange (Fig.6.17, 6.18)
Figure 6.17. Local delamination of the upper flange in the middle of the girder (the location of
structural notch made in the production stage of the girder)

failure of the epoxy adhesive joining steel plates of shear studs with the upper flange
(Fig.6.19)
177

local plastic deformation at the transition zone between the bottom flange and the web
(Fig.6.19)
scratch of laminate in the bottom flange.

Figure 6.18. Local delamination of the bottom flange at a distance of 7 m from the support
(left) and scratch of laminate at a distance of 10 m from the support (right)
Figure 6.19. Plastic deformation at the transition zone between the bottom flange and the web
in the middle of the girder (left) and the failure of epoxy adhesive joining upper flange and
steel plate of shear stud
Since any global failure during the static test was not observed, the determination of
damage inside the girder was based on the acoustic emission data. However, the prototype FRP
girder exhibited satisfying structural behaviour during the test and reasonable agreement with
the predictions from the FEM analysis. Moreover, the total safety factor obtained in the test
(γ=2.23) seems to be sufficient to plan further implementation of the girder in the bridge under
consideration.
After the tests were completed, the manufacturer produced four identical composite
girders according to bridge design. The girders were transported on the construction site and
installed on the bridge supports with a light crane (Fig.6.20). The assembly lasted only one
hour. After the girders were installed on the elastomeric bearings, the formwork of the edge
crossbeams and the bridge deck was performed, two reinforcement meshes made of GFRP bars
were installed and the bridge deck was concreted. After installing the bridge equipment, the
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structure went under safety tests and was opened to public. The bridge will be structurally
monitored by RUT for three years (com-bridge.pl).
Figure 6.20. FRP girders installation on the bridge supports with a light crane
The research project gives a very promising future for FRP girder applications in road
bridge construction. The VARTM manufacturing technique can be successfully applied to
produce large-scale bridge girders, taking into account laminate thickness to about 20 mm. In
order to make that elaborated technology more competitive, further research and demonstration
in large scale projects are of a great importance. Moreover, it will also be necessary to test the
long-term behaviour of this kind of structures as well as its real behaviour under the service
load.
6.3. Glulam timber
Wood is one of the earliest building materials, and often its use has been based more on
tradition than on principles of engineering. However, the structural use of wood and woodbased materials has increased steadily recently, including a renewed interest in the use of timber
as a material to construct bridges. Supporting this renewed interest has been an evolution of our
understanding of wood as a structural material and our ability to analyse and design safe,
durable, and functional timber bridges. In the mid-20th century, glued-laminated timber (or
glulam) was introduced as a viable structural material for bridges. The use of glulam grew to
become the primary material for timber bridges and has continued to grow in popularity.
Glued-laminated timber is a type of structural timber product comprising a number of
layers of dimensioned timber bonded together with durable, moisture-resistant structural
adhesives. By laminating a number of smaller pieces of timber, a single large, strong, structural
member is manufactured from smaller pieces. These structural members are used as many
bridge elements: vertical columns or horizontal beams, as well as curved, arched shapes.
Glulam is readily produced in curved shapes and it is available in a range of species and
appearance characteristics to meet varied bridge requirements. The laminating process allows
timber to be used for much longer spans, heavier loads and complex shapes. Connections are
usually made with bolts or plain steel dowels and steel plates. Glulam provides the strength and
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versatility of large wood members without relying on the old growth-dependent solid-sawn
timbers. Glulam is two-thirds the weight of steel and one sixth the weight of concrete. The high
strength and stiffness of laminated timbers enable glulam beams and arches to span large
distances without intermediate columns, allowing more design flexibility than with traditional
timber construction. The size is limited only by transportation and handling constraints.
Pressure-treated glulam timbers or timbers manufactured from naturally durable wood
species are well suited for creating bridges. Wood’s ability to absorb impact forces created by
traffic and its natural resistance to chemicals, such as those used for deicing roadways, make it
ideal for these installations. Since glulam timber can be satisfactorily treated with preservatives,
they offer a durable and long-lasting structural element. When designed so that field cutting,
drilling, and boring are avoided, glulam bridges can provide a service life of at least 50 years.
Glulam has been successfully used for pedestrian, road and even railway bridges. Glulam
timber beam bridges are perhaps the most prevalent forms of timber bridges today. This
popularity is primarily due to the large variety of member sizes offered by glulam. Commonly
used for clear spans ranging from 6 to 24 m, glulam beam bridges have been used for clear
spans up to 45 m.
Figure 6.21. The footbridge over Dunajec River in Sromowce
A well-known example of the Polish glulam structure is a footbridge in Sromowce, built
in 2004-2005. Spanning the Dunajec River, this cable-stayed footbridge links the village of
Sromowce Niżne in Poland to the village of Červený Kláštor in Slovakia. The deck of the
footbridge is suspended from a pylon consisting of steel tubes. The deck itself is a glued
laminated timber structure. On completion in 2006, the bridge became the longest glulam bridge
in the world, with a span 90 metres long (Fig.6.21).
The most recent example is a cable-stayed bridge for pedestrians and cyclists over the
San River, which connects two waterfront streets in Przemyśl. The footbridge is made up of
two spans, 2 x 68,5 m. The total length of the structure is 138 m, and its usable width – 3,0 m
(Fig.6.22).
The footbridge deck is composed of two glulam girders, made of class GL32c timber,
braced with steel crossbeams (Fig.6.23). Wind bracing at the crossbeam level is X-shaped,
made of steel bars with φ 30 mm. The deck surface is made of timber composite, laid
180
perpendicularly to the footbridge axis. The barriers are glulam girders, rising 135 cm above the
deck level and protected from the outside with wooden composite. All the structure is supported
with the set of cables, 32 and 40 mm in diameter, in a fan pattern. The steel pylon is 38,0 m
high. As the structural depth of the deck had to be limited, in the bottom part of the timber
girders there are the holes in which crossbeams are placed.
Figure 6.22. The scheme of the footbridge over San River in Przemyśl (dimension in cm)
Figure 6.23. The cross-section of the footbridge over San River in Przemyśl
(dimension in cm)
The A-shaped pylon is made of steel, with its legs crossed in the upper part. The legs
have box cross-sections, 550x500 mm, which are filled with C40/50 concrete in the bottom
sections. Beneath the deck, the legs of the pylon are braced with the crossbeam with the box
cross-section, 450x450 mm. The pylon and one abutment are founded on the CFA piles, another
abutment is seated directly on limestone. Due to the limitations to build-up the San river-bed
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(the area is included in Natura 2000 ecologically protected zone), the chosen technology was
to assemble the whole structure along with the pylon on temporary supports, located along the
river bank, and after the installation of the cables – to rotate the footbridge by 90 degrees to
reach its final position (Fig.6.24).
Figure 6.24. The footbridge superstructure during rotation
6.4. Self-compacting concrete
Self-compacting concrete (SCC) is a highly flowable, non-segregating concrete that
spreads into place, fills formwork, and encapsulates even the most congested reinforcement, all
without any mechanical vibration. It is defined as a concrete mix that can be placed purely by
means of its own weight, with little or no vibration. As a high-performance concrete, SCC
delivers these attractive benefits while maintaining all of concrete's customary mechanical and
durability characteristics. Adjustments to traditional mix designs and the use of
superplasticizers creates flowing concrete that meets tough performance requirements. If
needed, low dosages of viscosity modifier can eliminate unwanted bleeding and segregation.
Since its inception in the 1980s, the use of SCC has grown tremendously. The
development of high performance polycarboxylate polymers and viscosity modifiers have made
it possible to create “flowing” concrete without compromising durability, cohesiveness, or
compressive strength. The flowability of SCC is measured in terms of spread when using the
slump test. The spread (slump flow) of SCC typically ranges from 455 to 810 mm depending
on the requirements for the project. The viscosity, as visually observed by the rate at which
concrete spreads, is an important characteristic of plastic SCC and can be controlled when
designing the mix to suit the type of application being constructed.
SCC's unique properties give it significant economic, constructability, aesthetic and
engineering advantages. SCC is an increasingly attractive choice for optimizing site manpower
(through reduction of labor and possibly skill level), lowering noise levels, and allowing for a
safer working environment. SCC allows easier pumping (even from bottom up), flows into
complex shapes, transitions through inaccessible spots, and minimizes voids around embedded
182
items to produce a high degree of homogeneity and uniformity. That's why SCC allows for
denser reinforcement, optimized concrete sections and shapes, and greater freedom of design
while producing superior surface finishes and textures.
The first significant Polish application of SCC to build a bridge took place in 2002,
when the Zamkowy Bridge over the Wisłok River was built in the city centre of Rzeszów. The
total length of the bridge is 173,0 m. The bridge is the continuous five-span structure with the
span lengths of 2 x 44,0 + 3 x 28,0 m (Fig.6.25). Two middle spans over the Wislok river are
supported on three RC fixed arches with the span lengths of 50,0 m each (Fig.6.26). Due to the
local area configuration arches are relatively flat (f/l = 1/5,8). Due to very dense reinforcement
mesh of RC arches they were made of the self-compacting concrete with the total volume of
about 900 m3 and with the standard compressive strength not less than 50 MPa. It was one of
the largest world structural applications of SCC for bridge structures. Moreover, anticorrosive
concrete with migrating corrosion inhibitors was applied in the safety barriers.
Figure 6.25. The longitudinal cross-section of the Zamkowy Bridge (dimension in cm)
Figure 6.26. The bridge cross-section (left) and flat RC arches in the middle spans (right)
(dimension in cm)
The superstructure is made up of three steel box-girders with the depth of 1,5 m and
axial spacing of 6,4 m, composite with RC deck slab 0,25 m thick. In the main span the
superstructure is supported with three RC arches with altering cross-section: from 2,5 m square
section on the base to the rectangular 1,7 x 2,5 m section in crown. The arches are founded on
diaphragm walls with the lengths of 16,0 m, creating together a rigid frame structure. The bridge
183
abutments are founded also on diaphragm walls, but pillars are founded directly on the ground.
The total width of the deck is 21,0 m. The 12,0 m four-lane carriageway is situated on the deck
along with two pedestrian walkways with the width of 1,5 m, two cycle lanes with the width of
1,5 m, separated from the carriageway with the concrete barriers. In the middle of the central
span two viewing galleries were designed by means of widening the deck slab.
Figure 6.27. Reinforcing and concreting of the arches of Zamkowy Bridge: assembling of
reinforcing mesh (top left), concrete pouring (bottom left), arch concrete treatment (right)
The calculations for the arches revealed really big bending moments, which, due to the
limitations in the arch cross-section, resulted in dense reinforcement (Fig.6.27). That was the
main reason to apply self-compacting concrete in arch performance. The formula of SCC to be
applied in the bridge arches was prepared and the studies on the properties of respective
mixtures were carried out at RUT (Radomski, 2003). After several tests were performed, the
following basic formula was accepted (the amount of each ingredient in 1 m3 self-compacting
mixture):
 cement: CEM I 42,5 HSR NA by Rejowiec - 450 kg
 additives: liquid silica fume Woerosil 500 S-P by Woermann – 70 kg
 admixtures: superplasticizer Visco-Crete 3 by Sika – 1.1 % of the cement mass, i.e. 4,95 kg
 air entraining admixture Sikanol A 12 by Sika – 0.055% of the cement mass, i.e. 0,248 kg
 sand aggregate 0/2 “Jagniówka” – 689 kg
 basalt aggregate 2/8 “Wilków” – 437 kg
184


basalt aggregate 8/16 “Wilków” – 619 kg
water – 175 kg, of which 136 kg is mixing water, 35 kg is absorption water and 4 kg is the
water included in admixtures.
The water-to-cement ratio was 0,389 while air entraining was 5 %.
The arches were concreted in a continuous way by means of standard concrete pumps,
simultaneously from the both river banks, starting from the bottom and finishing at the arch
key. Concreting of a single arch with the capacity of 290 m3 took from 18 to 22 hours. When
the operation was completed, the arch was covered with mats and traditionally cured for 7 days.
Removing the formwork from each arch took place 14 days after the concrete was poured.
During the preparation of the concrete mixture and after the concrete hardened, numerous tests
were carried out, with the results as follow:
 concrete consistency after 5 min: 610 - 620 mm and after 30 min: 660 – 670 mm
 air entrainment after 30 min – 4,8 %
 concrete compressive strength: after 3 days – 40,2 ÷ 40,5 MPa, after 7 days – 60,2 ÷ 61,5
MPa, after 28 days – 80,6 ÷ 81,6 MPa
 frost resistance – at least F150
 water resistance – at least W8
 water absorption – 3,6 %.
The specifications quoted above show that the applied concrete mix met the demands to
be used to construct the arches (and even exceeded the norms in terms of compression strength),
and the consistency corresponded the limits laid down for the types of concrete with the
maximum aggregate fraction 16 mm. The first Polish application of self-compacting concrete
let us gain the experience which turned out to be helpful in the following applications of that
kind of advanced material.
185
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